Full-Scale Field Tests of Different Types of Piles: Jialin Zhou Erwin Oh

Advanced Topics in Science
and Technology in China 62
Jialin Zhou
Erwin Oh
Full-Scale Field
Tests of Different
Types of Piles
Project-Based Study
Advanced Topics in Science and Technology
in China
Volume 62
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Jialin Zhou · Erwin Oh
Full-Scale Field Tests

of Different Types of Piles
Project-Based Study
Jialin Zhou Erwin Oh
School of Engineering and Built School of Engineering and Built
Environment Environment
Griffith University Griffith University
Gold Coast, QLD, Australia Gold Coast, QLD, Australia
ISSN 1995-6819 ISSN 1995-6827 (electronic)

Advanced Topics in Science and Technology in China
ISBN 978-981-33-6182-9 ISBN 978-981-33-6183-6 (eBook)
https://doi.org/10.1007/978-981-33-6183-6
Jointly published with Zhejiang University Press

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Preface
This book provides full-scale field tests of different types of pile foundations. For the
testing, the traditional static load tests which consider various loading orientations are
provided. In detail, the compressive, uplift and horizontal static load tests with project
case studies are demonstrated. For the test in which loading capacity is limited, the
improved static load test is introduced. This improved testing method can be used
for huge capacity determination of super-long and large diameter pile foundations.
Moreover, the dynamic load tests, inclinometer monitoring, and the tests that aim to
determine the load transfer mechanism of pile foundation are detailed in this book.
This book also covers the up-to-date popular topic with detailed project studies.
This includes the academic investigation of post-grouting technology effect on
drilled shaft piles, the research of displacement and non-displacement precast pile
foundation, the study of fiber-reinforced polymer (FRP) material used in the
geotechnical environment such as deep excavation pit in tunneling project, and the
research of super-long and large diameter pile foundations. The study of the
composite pile foundation provided in this book contains two types: the FRP bar
reinforced concrete piles and FRP laminar confined concrete piles; these new types
of composite piles provide practical information to the geotechnical engineer, and
they could be an alternative for the traditional pile in some specific environmental
conditions. Most importantly, this book provides pile failure tests considering
different failure conditions, the failure caused by inappropriate pile construction,
eccentric loading, plunging failure of piling system are illustrated. All these
investigations provide essential and academic information for researchers as well as
engineers in the role of Civil and Geotech.
v
vi Preface
Not only the different types of piles are studied, but also the relevant theory and
literatures are reviewed in this book. In addition, the diagrams are plotted in an easy
way and the explanation of the diagrams and tables are described in detail. The
research methods corresponding to the practical projects are illustrated in detail as
well, hence it can be a perfect book for the bachelor and master’s degree students
whose objectives are related to the civil and geotechnical engineering.
Gold Coast, Australia Jialin Zhou

Contents
1 Introduction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1
1.1 Background . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1
1.2 Objectives . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2
1.3 Book Organization . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 4
2 General Principles and Practices . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5
2.1 General Introduction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5
2.2 Soil Stratum Determination . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6
2.2.1 Laboratory Tests . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6
2.2.2 In-Situ Tests . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 22
2.3 Types of Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 28
2.3.1 General Pile Categorization . . . . . . . . . . . . . . . . . . . . . . . . . . . . 28
2.3.2 Cast-In-Situ Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 30
2.3.3 Precast Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 33
2.3.4 Composite Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 34
2.3.5 Other Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 41
2.4 Analytical Method for Bearing Capacity . . . . . . . . . . . . . . . . . . . . . . . 45
2.4.1 Meyerhof Method . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 45
2.4.2 Brown Method . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 46
2.4.3 Nordlund Method . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 47
2.4.4 Tomlinson Method . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 48
2.4.5 Effective Method . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 49
2.5 Field Tests . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 52
2.5.1 Static Load Tests . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 52
2.5.2 Dynamic Load Test . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 64
2.5.3 Osterberg Cell Test . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 68
2.5.4 Statnamic Test . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 70
2.6 Concluding Remarks . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 73
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 73
vii
viii Contents
3 Field Tests of Post Grouted Concrete Piles . . . . . . . . . . . . . . . . . . . . . . . . 81

3.2 Site Conditions . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 84
3.3 Pile Description . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 85
3.4 Static Load Test Setup . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 87
3.5 Dynamic Load Test Setup . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 89
3.6 Static Load Tests Results . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 90
3.6.1 Compressive Static Load Tests . . . . . . . . . . . . . . . . . . . . . . . . . 90
3.6.2 Uplift Load Static Load Tests . . . . . . . . . . . . . . . . . . . . . . . . . . 97
3.7 Dynamic Load Tests Results . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 101
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 105
4 Field Tests of Precast Concrete Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 107
4.2 Small and Large Displacement Concrete Piles . . . . . . . . . . . . . . . . . . 110
4.2.1 Project Introduction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 110
4.2.2 Subsurface Conditions . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 110
4.2.3 Pile Preparation . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 112
4.2.4 Test Setup . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 112
4.3 Non-displacement Square Concrete Piles . . . . . . . . . . . . . . . . . . . . . . 113
4.3.3 Pile Descriptions . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 118
4.3.4 Installation Process of Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . 119
4.3.5 Test Setup . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 120
4.4 Static Load Test Results of Precast Concrete Piles . . . . . . . . . . . . . . . 121
4.4.1 Results of Small Displacement Precast Piles . . . . . . . . . . . . . 121
4.4.2 Results of Large Displacement Precast Piles . . . . . . . . . . . . . 123
4.4.3 Results of Non-displacement Precast Piles . . . . . . . . . . . . . . . 124
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 136
5 Field Performance of Composite Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . 139
5.2 CFRP Laminar Concrete Composite Piles . . . . . . . . . . . . . . . . . . . . . . 143
5.2.1 Background Introduction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 143
5.2.4 Test Setup . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 146
5.2.5 Results of CFRP Laminar–Confined Concrete Piles . . . . . . . 147
5.3 GFRP Bar–Reinforced Concrete Composite Piles . . . . . . . . . . . . . . . 155
5.3.4 Test Setup and Procedures . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 162
Contents ix
5.3.5 Results of GFRP Bar–Reinforced Concrete Piles . . . . . . . . . 163

References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 170
6 Field Tests of Super-Long and Large Diameter Piles . . . . . . . . . . . . . . . 173
6.2 Compressive Static Load Tests . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 174
6.2.1 Geotechnical Conditions . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 174
6.2.2 Description of Tested Drilled Shaft Piles . . . . . . . . . . . . . . . . 174
6.2.3 Test Setup . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 177
6.2.4 Test Process . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 180
6.3 Test Results and Discussion . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 181
6.3.1 Settlement–lg Time and Load–Settlement Curves . . . . . . . . . 181
6.3.2 Load Transfer Mechanism . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 185
6.3.3 Shaft Resistance Development and Distribution . . . . . . . . . . 187
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 194
7 Piles Under Ultimate Loads . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 197
7.2 Subsurface Explorations . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 199
7.3 Designed Ultimate Bearing Capacity . . . . . . . . . . . . . . . . . . . . . . . . . . 200
7.4 Pile Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 204
7.5 Test Setup . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 204
7.6 Observation of Pile Under Failure Condition . . . . . . . . . . . . . . . . . . . 206
7.6.1 Piles with Inadequate Rigidity . . . . . . . . . . . . . . . . . . . . . . . . . 206
7.6.2 Piles Suffering from Eccentric Loads . . . . . . . . . . . . . . . . . . . 206
7.6.3 Failure from Inadequate Soil Rigidity . . . . . . . . . . . . . . . . . . . 208
7.7 Test Results . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 209
7.7.1 Pile with Achieved Design Requirements . . . . . . . . . . . . . . . . 209
7.7.2 Failure from Inadequate Concrete Strength . . . . . . . . . . . . . . 211
7.7.3 Failure from Eccentricity . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 214
7.7.4 Plunging Failure . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 215
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 223
8 Capacity and Settlement Analysis . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 225
8.2 Post Grouted Concrete Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 226
8.2.1 Capacity Discussion . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 226
8.2.2 Settlement Discussion . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 228
8.3 Precast Concrete Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 243
8.3.1 Small and Large Displacement Piles . . . . . . . . . . . . . . . . . . . . 243
8.3.2 Non-displacement Precast Piles . . . . . . . . . . . . . . . . . . . . . . . . 257
8.4 Concrete Composite Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 261
8.4.1 FRP Laminar–Confined Composite Piles . . . . . . . . . . . . . . . . 261
x Contents
8.4.2 FRP Bar–Reinforced Composite Piles . . . . . . . . . . . . . . . . . . . 263

8.5 Piles Under Ultimate Load . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 264
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 265
9 Conclusions and Recommendations . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 267
9.2 Post Grouted Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 267
9.3 Precast Concrete Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 268
9.4 Concrete Composite Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 269
9.5 Super-Long and Large Diameter Piles . . . . . . . . . . . . . . . . . . . . . . . . . 270
9.6 Piles Under Ultimate Load . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 271
Abbreviations
AASHTO American Association of State Highway and Transportation Officials

AFRP Aramid Fibre-reinforced Polymer
BFRP Basalt Fibre-reinforced Polymer
CAPWAP Case Pile Wave Analysis Program
CD Consolidated Drained Tri-axial Tests
CFRP Carbon Fibre-reinforced Polymer
CPT Cone Penetration Test
CU Consolidated Undrained Tri-axial Tests
DCM Deep Cement Mixing
ER Equipment Energy Ratio
FHWA Federal Highway Administration
FRP Fibre-reinforced Polymer
GFRP Glass Fibre-reinforced Polymer
LVDT Linear Variable Differential Transformer
O-cell Osterberg Cell
OCR Over-consolidation Ratio
PDA Pile Driving Analyser
PT(s) Proof Test(s)
PHC Pretensioned Spun High-strength Concrete
SPE Soil Plug Effect
ROV Remote-operated Vehicles
RSP Static Resistance on Pile during Driving
RTL Total Static and Dynamic Resistance during Driving
SLT Static Load Test
SMW Soil Mixing Wall
SPT Standard Penetration Test
xi
xii Abbreviations
SRP Structurally Reinforced Plastic

TBM Tunnel Boring Machine
TDM T-shaped Deep Mixed
UU Unconsolidated Undrained Tri-axial Tests
Notations
N-value Standard Penetration Test (SPT) Blow Count

N 60 Corrected SPT N-value
(N 1 )60 Normalised Blow Count
CN Overburden Correction Factor
ID Relative Density

σv0 Overburden Pressure
Cu Undrained Shear Strength
Pa Atmospheric Pressure
IP Plasticity Index
qu Unconfined Compressive Strength
γ Unit Weight
e Void Ratio
σ Vertical Effective Stress
Cc Compression Index
Ce Expansion Index
mv Coefficient of Volume Compressibility

E oed One-dimensional Elastic Modulus
w Moisture Content
ρ Material Density
Cz Coefficient of Curvature
IL Liquidity Index
Uv Degree of Consolidation
Cv Coefficient of Consolidation
c Cohesion of Soil
c Effective Cohesion of Soil
ϕ Effective Friction Angle of Soil
Tv Time Factor
σ Normal Stress of Soil
ϕ Friction Angle of Soil
μ Coefficient of Friction
NO Average Corrected SPT N-value for Stratum Overlying Bearing Stratum
xiii
xiv Notations
NB Average Corrected SPT N-value of Bearing Stratum

DB Pile Embedment Depth into Bearing Stratum
Kδ Lateral Earth Pressure Coefficient at Depth d
CF Correction Factor for K δ
αt Depth–Width Relationship Factor
Nq Capacity Factor
qc Cone Tip Resistance of Cone Penetration Test (CPT)
fs Shaft Resistance
qt Toe or Tip Resistance
B Bjerrum-Burland Beta Coefficient
Ks Earth Pressure Coefficient
Interface Friction Angle between Pile and Soil
Nt Toe Bearing Capacity Coefficient
pt Effective Overburden Pressure at Pile Toe
sf Offset Settlement
Qu Ultimate Bearing Capacity of Pile
Qb Base Resistance of Pile Foundation
Qs Shaft Resistance of Pile Foundation
Rs Compressive Settlement Ratio
Rd Uplift Displacement Ratio
Rh Horizontal Displacement Ratio
St Vertical Settlement at Certain Loading at Certain Time
Ht Horizontal Displacement at Certain Loading at Certain Time
S (Fi) Vertical Settlement at End of Certain Loading Stage
H (Fi) Horizontal Displacement at End of Certain Loading Stage
S (In) Settlement at Beginning of Certain Loading Stage
H (In) Horizontal Displacement at Beginning of Certain Loading Stage
Quk Ultimate Bearing Capacity of Precast Concrete Pile
Qsk Ultimate Shaft Capacity of Precast Concrete Pile
Qpk Ultimate End Capacity of Precast Concrete Pile
qsik Shaft Resistance of Precast Concrete Pile
qqk End Resistance of Precast Concrete Pile
Ap Total Area from Pipe Pile Toe
Aj Effective Area of Pipe Pile Toe
λp Plug Effect Coefficient
Ap1 Hollow Area of Pipe Pile Toe
hb Embedment Depth of Pile in Bearing Stratum
H cr Critical Load of Horizontal Static Load Tests (SLTs)
Hu Ultimate Load of Horizontal SLTs
RP-S Permanent Settlement Ratio
S max Settlement under Maximum Applied Load
Sp Permanent Settlement
QMax Maximum Applied Loads during SLTs
Qa Allowable Load of Pile
Notations xv
Ft Converted Force by Pile Driving Analyser (PDA)

εt Recorded Strain by Strain Transducers
Vt Converted Velocity by PDA
at Recorded Acceleration by Accelerometer
Fstn Measured Statnamic Force
Fv Damping Force from Soil
Fp Pore Water Pressure Resistance
List of Figures
Fig. 2.1 Direct shear apparatus-shear box . . . . . . . . . . . . . . . . . . . . . . . . . . 7

Fig. 2.2 Typical results from drained direct shear test (AS
1289.6.2.2, 1998) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8
Fig. 2.3 Determination of friction angle from direct shear test (AS
1289.6.2.2, 1998) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8
Fig. 2.4 Components of shear box . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9
Fig. 2.5 Direct shear apparatus . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10
Fig. 2.6 Horizontal movement versus shear stress . . . . . . . . . . . . . . . . . . . 11
Fig. 2.7 Horizontal movement versus vertical movement . . . . . . . . . . . . . 11
Fig. 2.8 Normal stress versus shear stress . . . . . . . . . . . . . . . . . . . . . . . . . . 12
Fig. 2.9 Basic theory of element under stress . . . . . . . . . . . . . . . . . . . . . . . 13
Fig. 2.10 Void ratio–effective stress relationship (Craig 1983) . . . . . . . . . . 13
Fig. 2.11 Apparatus of consolidation test . . . . . . . . . . . . . . . . . . . . . . . . . . . 14
Fig. 2.12 Typical compression versus time (AS 1289.6.6.1, 1998) . . . . . . . 15
Fig. 2.13 The log time method . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 16
Fig. 2.14 The root time method . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 17
Fig. 2.15 Void ratios versus vertical pressure of consolidation tests . . . . . . 17
Fig. 2.16 Theory of soil element at failure state . . . . . . . . . . . . . . . . . . . . . . 18
Fig. 2.17 Mohr–Coulomb failure criterion . . . . . . . . . . . . . . . . . . . . . . . . . . 19
Fig. 2.18 The triaxial apparatus . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 20
Fig. 2.19 Set of three triaxial tests . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 21
Fig. 2.20 Mohr–Coulomb circle of triaxial tests . . . . . . . . . . . . . . . . . . . . . . 22
Fig. 2.21 Standard Penetration Test schematic (Mayne et al. 2002) . . . . . . 23
Fig. 2.22 Overburden correction factor (Skempton 1986) . . . . . . . . . . . . . . 24
Fig. 2.23 Variation of overburden correction factor CN (Samtani
and Nowatzki 2006a) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 24
Fig. 2.24 Schematic of CPT truck (Samtani and Nowatzki 2006a) . . . . . . . 26
Fig. 2.25 Soil behaviour type classification chart (Robertson 1990) . . . . . . 26
Fig. 2.26 Piezocone measurement (CPTu ) (Samtani and Nowatzki
2006a) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 27
Fig. 2.27 Piezocone results for Apple Freeway Bridge (Samtani
and Nowatzki 2006a) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 28
xvii
xviii List of Figures
Fig. 2.28 Pile classification chart (Hannigan et al. 2006) . . . . . . . . . . . . . . . 29

Fig. 2.29 Situations in which deep pile foundations may be needed.
(modified from Vesic 1977) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 31
Fig. 2.30 Cast-in-situ concrete pile using slurry support technology . . . . . 32
Fig. 2.31 Concrete-filled pipe pile (Khaleghi et al. 2016) . . . . . . . . . . . . . . 35
Fig. 2.32 Deep soil mixing (Madhyannapu and Puppala 2015) . . . . . . . . . . 36
Fig. 2.33 Corrosion of steel, deterioration of timber and degradation
of concrete (Pando 2003) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 37
Fig. 2.34 Degraded concrete piles in Adelaide, South Australia,
Australia . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 37
Fig. 2.35 Types of FRP composite piles (Guades et al. 2012) . . . . . . . . . . . 39
Fig. 2.36 Tested composite piles (Juran and Komornik 2006) . . . . . . . . . . . 40
Fig. 2.37 Ribbed bored piles (McNamara and Gorasia 2016) . . . . . . . . . . . 42
Fig. 2.38 Extended end concrete pile (Kim et al. 2017) . . . . . . . . . . . . . . . . 42
Fig. 2.39 Structure of the two-layered composite pile (Li et al. 2015) . . . . 43
Fig. 2.40 Failure patterns of stiffened DCM piles (Wonglert
and Jongpradist 2015) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 44
Fig. 2.41 TDM column–supported embankment (Liu et al. 2011) . . . . . . . 44
Fig. 2.42 Nordlund’s general equation for ultimate pile capacity
(Nordlund 1979) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 47
Fig. 2.43 Adhesion factor determination scenario A (Samtani
and Nowatzki 2006b) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 49
Fig. 2.44 Adhesion factor determination scenario B (Samtani
and Nowatzki 2006b) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 49
Fig. 2.45 Adhesion factor determination scenario C (Samtani
and Nowatzki 2006b) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 50
Fig. 2.46 Chart for estimating β coefficient as a function of soil type
ϕ’ (Fellenius 1991) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 51
Fig. 2.47 Chart for N t coefficients as a function of soil type ϕ’ angle
(Fellenius 1991) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 51
Fig. 2.48 Basic mechanism of a compression pile load test (Kyfor
et al. 1992) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 53
Fig. 2.49 Schematic of anchored reaction frame (Hannigan et al.
2016) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 54
Fig. 2.50 Typical setup for tensile load test (Courtesy of WKG2 ) . . . . . . . . 55
Fig. 2.51 Setup for lateral static load test (Courtesy of WKG2 ) . . . . . . . . . 56
Fig. 2.52 Typical results of compressive SLT (Hannigan et al. 2006) . . . . . 57
Fig. 2.53 s-lgt result of compressive SLT . . . . . . . . . . . . . . . . . . . . . . . . . . . 57
Fig. 2.54 Interpretation based on Davisson’s method (US unit)
(Samtani and Nowatzki 2006b) . . . . . . . . . . . . . . . . . . . . . . . . . . . 58
Fig. 2.55 Interpretation based on double tangent method (Samtani
and Nowatzki 2006b) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 59
Fig. 2.56 Interpretation based on DeBeer’s method . . . . . . . . . . . . . . . . . . . 60
Fig. 2.57 Interpretation based on Chin’s method . . . . . . . . . . . . . . . . . . . . . 60
List of Figures xix
Fig. 2.58 Determination of shaft and end resistance based on Chin’s

method . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 61
Fig. 2.59 Typical results of uplift SLT . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 61
Fig. 2.60 Typical tension load test for load–movement curve
(Hannigan et al. 2006) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 62
Fig. 2.61 Modified Mazurkiewicz method . . . . . . . . . . . . . . . . . . . . . . . . . . 62
Fig. 2.62 Typical lateral load test for pile head load–deflection curve
(Hannigan et al. 2006) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 63
Fig. 2.63 Deflection behaviour versus depth (Kyfor et al. 1992) . . . . . . . . . 63
Fig. 2.64 a Schematic diagram of apparatus for dynamic test;
b Strain transducer and accelerometer installed on pile
surface (Hannigan et al. 2016) . . . . . . . . . . . . . . . . . . . . . . . . . . . . 65
Fig. 2.65 Typical force and velocity traces (Hannigan et al. 2016) . . . . . . . 65
Fig. 2.66 Wave mechanics of rod and pile . . . . . . . . . . . . . . . . . . . . . . . . . . . 66
Fig. 2.67 Soil resistance effects on force and velocity records
(Hannigan 1990) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 67
Fig. 2.68 Schematic of CAPWAP analysis method (Hannigan et al.
2016) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 68
Fig. 2.69 Schematic of test pile by O-cell testing. Source
www.loadtest.com . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 69
Fig. 2.70 Typical O-cell results—shaft failure. Source
www.loadtest.com . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 70
Fig. 2.71 Typical O-cell results—end failure. Source
www.loadtest.com . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 70
Fig. 2.72 Schematic of test pile by statnamic testing. Source
www.profound.nl . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 71
Fig. 2.73 Typical result presentation by statnamic test (Courtesy
of Berminghammer Foundation Equipment) . . . . . . . . . . . . . . . . . 72
Fig. 3.1 Grouting concrete pile with cement layer (Zhou et al. 2017b) . . . 82
Fig. 3.2 Grouting pipe system of concrete pile . . . . . . . . . . . . . . . . . . . . . . 86
Fig. 3.3 Mud slurry support . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 87
Fig. 3.4 Compressive SLTs setup . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 88
Fig. 3.5 Uplift SLTs setup . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 88
Fig. 3.6 Dynamic load tests setup . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 89
Fig. 3.7 Results of Q-s curvatures of compressive loaded piles . . . . . . . . . 90
Fig. 3.8 s-lgQ curves of tested piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 91
Fig. 3.9 s-lgt curves of P51 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 92
Fig. 3.10 s-lgt curves of P121 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 92
Fig. 3.11 s-lgt curves of P126 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 93
Fig. 3.12 Double tangent method of P51 . . . . . . . . . . . . . . . . . . . . . . . . . . . . 94
Fig. 3.13 Double tangent method of P121 . . . . . . . . . . . . . . . . . . . . . . . . . . . 94
Fig. 3.14 Double tangent method of P126 . . . . . . . . . . . . . . . . . . . . . . . . . . . 95
Fig. 3.15 DeBeer’s method of tested piles . . . . . . . . . . . . . . . . . . . . . . . . . . . 95
Fig. 3.16 Davisson’s offset method of tested piles . . . . . . . . . . . . . . . . . . . . 96
Fig. 3.17 Chin’s method of tests piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 96
xx List of Figures
Fig. 3.18 Results of Q-s curvatures of uplift loaded piles . . . . . . . . . . . . . . 97

Fig. 3.19 s-lgQ curves of tested piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 98
Fig. 3.22 Offset method of tested piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 100
Fig. 3.23 Mazurkiewicz method of P15 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 100
Fig. 3.24 Mazurkiewicz method of P16 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 101
Fig. 3.25 CAPWAP results of P121 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 101
Fig. 3.26 CAPWAP results of P126 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 102
Fig. 3.27 Simulated load–settlement curves, unit shaft resistance
and load transfer characteristic of P121 . . . . . . . . . . . . . . . . . . . . . 102
Fig. 3.28 Simulated load–settlement curves, unit shaft resistance
and load transfer characteristic of P126 . . . . . . . . . . . . . . . . . . . . . 102
Fig. 4.1 Location of tested piles (not to scale) . . . . . . . . . . . . . . . . . . . . . . 110
Fig. 4.2 Driven small displacement pipe pile . . . . . . . . . . . . . . . . . . . . . . . 111
Fig. 4.3 Weighted platform of SLT . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 112
Fig. 4.4 Location of general engineering construction site
(not to scale) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 117
Fig. 4.5 Tested non-displacement precast pile . . . . . . . . . . . . . . . . . . . . . . 118
Fig. 4.6 Precast piles in factory . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 120
Fig. 4.7 Auger drilling machine . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 121
Fig. 4.8 Drilled hole for precast pile . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 122
Fig. 4.9 Transferring precast piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 123
Fig. 4.10 After descending the first piece of pile . . . . . . . . . . . . . . . . . . . . . 124
Fig. 4.11 Lateral SLT setup . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 125
Fig. 4.12 Uplift SLT setup . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 126
Fig. 4.13 Load–settlement results of 21 m pipe pile . . . . . . . . . . . . . . . . . . . 126
Fig. 4.14 Load–settlement results of 22 m pipe pile . . . . . . . . . . . . . . . . . . . 127
Fig. 4.15 Load–settlement results of 24 m pipe piles . . . . . . . . . . . . . . . . . . 127
Fig. 4.16 Pipe piles at Area A with different lengths . . . . . . . . . . . . . . . . . . 128
Fig. 4.17 Pipe piles at Area B with different lengths . . . . . . . . . . . . . . . . . . 128
Fig. 4.18 Load–settlement results of 27 m rectangular piles . . . . . . . . . . . . 129
Fig. 4.19 Load–settlement results of 28 m rectangular piles . . . . . . . . . . . . 129
Fig. 4.20 Same pipe with different load increments . . . . . . . . . . . . . . . . . . . 130
Fig. 4.21 Rectangular piles at Area C with different lengths . . . . . . . . . . . . 130
Fig. 4.22 H–X curves of P15 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 131
Fig. 4.23 H–X curves of P163 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 131
Fig. 4.24 H–X curves of P152 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 132
Fig. 4.25 Horizontal SLT interpretation of P15 . . . . . . . . . . . . . . . . . . . . . . . 132
Fig. 4.26 Horizontal SLT interpretation of P163 . . . . . . . . . . . . . . . . . . . . . . 133
Fig. 4.27 Horizontal SLT interpretation of P152 . . . . . . . . . . . . . . . . . . . . . . 133
Fig. 4.28 Load–displacement results of uplift load piles . . . . . . . . . . . . . . . 134
Fig. 4.29 Modified Mazurkiewicz method of uplift loaded piles . . . . . . . . . 134
Fig. 4.30 Load transfer mechanism of P156 . . . . . . . . . . . . . . . . . . . . . . . . . 135
Fig. 4.31 Accumulated shaft resistance . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 135
List of Figures xxi
Fig. 4.32 Unit shaft resistance of P156 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 136

Fig. 5.1 Preparation of the concrete piles . . . . . . . . . . . . . . . . . . . . . . . . . . 144
Fig. 5.2 Preparation of the composite material . . . . . . . . . . . . . . . . . . . . . . 144
Fig. 5.3 CFRP-confined concrete piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . 145
Fig. 5.4 Installation of composite rectangular pile . . . . . . . . . . . . . . . . . . . 145
Fig. 5.5 Installation of composite pipe pile . . . . . . . . . . . . . . . . . . . . . . . . . 146
Fig. 5.6 Compressive SLT . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 147
Fig. 5.7 Equipment for SLT . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 147
Fig. 5.8 Q-s curves of tested rectangular piles . . . . . . . . . . . . . . . . . . . . . . 148
Fig. 5.9 Q-s curves of tested pipe piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . 148
Fig. 5.10 s-lgQ curves of four tested piles . . . . . . . . . . . . . . . . . . . . . . . . . . . 149
Fig. 5.11 P116 CFRP-confined concrete rectangular pile . . . . . . . . . . . . . . 150
Fig. 5.12 P115 concrete rectangular pile . . . . . . . . . . . . . . . . . . . . . . . . . . . . 150
Fig. 5.13 P2108 CFRP-confined concrete pipe pile . . . . . . . . . . . . . . . . . . . 151
Fig. 5.14 P2054 concrete pipe pile . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 151
Fig. 5.15 Double tangent method of rectangular piles . . . . . . . . . . . . . . . . . 152
Fig. 5.16 Double tangent method of pipe piles . . . . . . . . . . . . . . . . . . . . . . . 152
Fig. 5.17 DeBeer’s method of rectangular piles . . . . . . . . . . . . . . . . . . . . . . 153
Fig. 5.18 DeBeer’s method of pipe piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . 154
Fig. 5.19 Davisson’s offset method of rectangular piles . . . . . . . . . . . . . . . . 154
Fig. 5.20 Davisson’s offset method of pipe piles . . . . . . . . . . . . . . . . . . . . . 155
Fig. 5.21 Chin’s method of P116 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 155
Fig. 5.22 Chin’s method of P115 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 156
Fig. 5.23 Chin’s method of P2054 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 156
Fig. 5.24 Chin’s method of P2018 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 157
Fig. 5.25 Construction plan map (not to scale) . . . . . . . . . . . . . . . . . . . . . . . 157
Fig. 5.26 Soil layers adjacent to pile (not to scale) . . . . . . . . . . . . . . . . . . . . 159
Fig. 5.27 GFRP stirrups of piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 160
Fig. 5.28 GFRP reinforcements of piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . 160
Fig. 5.29 Assembled GFRP cage of pile . . . . . . . . . . . . . . . . . . . . . . . . . . . . 161
Fig. 5.30 Assembled steel cage of pile . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 161
Fig. 5.31 PVC tube on GFRP cage . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 162
Fig. 5.32 PVC tube on steel cage . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 162
Fig. 5.33 Readout probe for deflection measurement of pile . . . . . . . . . . . . 163
Fig. 5.34 Installation of steel-reinforced concrete pile . . . . . . . . . . . . . . . . . 164
Fig. 5.35 Excavation after first concrete support being installed . . . . . . . . . 164
Fig. 5.36 Deflection of CFRP pile at the beginning of excavation . . . . . . . . 165
Fig. 5.37 First steel support applications . . . . . . . . . . . . . . . . . . . . . . . . . . . . 166
Fig. 5.38 Second steel support applications . . . . . . . . . . . . . . . . . . . . . . . . . 166
Fig. 5.39 a Lateral deflection of GFRP pile during three stages;
b Total horizontal deflection of FRP pile . . . . . . . . . . . . . . . . . . . . 168
Fig. 5.40 Deflection of a GFRP bar–reinforced and b steel bar–
reinforced concrete bored piles . . . . . . . . . . . . . . . . . . . . . . . . . . . 169
Fig. 6.1 Subsurface conditions and tested piles (not to scale) . . . . . . . . . . 175
Fig. 6.2 Simplified diagram of the tested piles (not to scale) . . . . . . . . . . . 176
xxii List of Figures
Fig. 6.3 Reaction system design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 177

Fig. 6.4 Anchoring reaction system . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 178
Fig. 6.5 Construction process . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 180
Fig. 6.6 s–lgt results of P12 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 181
Fig. 6.7 s-lgt results of P66 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 182
Fig. 6.8 s-lgt results of P105 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 183
Fig. 6.9 Load–settlement curves of P12 . . . . . . . . . . . . . . . . . . . . . . . . . . . 183
Fig. 6.10 Load–settlement curves of P66 . . . . . . . . . . . . . . . . . . . . . . . . . . . 184
Fig. 6.11 Load–settlement curves of P105 . . . . . . . . . . . . . . . . . . . . . . . . . . 184
Fig. 6.12 Load–settlement curves of tested piles . . . . . . . . . . . . . . . . . . . . . 185
Fig. 6.13 Load transfer mechanism of P12 . . . . . . . . . . . . . . . . . . . . . . . . . . 186
Fig. 6.14 Load transfer mechanism of P66 . . . . . . . . . . . . . . . . . . . . . . . . . . 186
Fig. 6.15 Load transfer mechanism of P105 . . . . . . . . . . . . . . . . . . . . . . . . . 187
Fig. 6.16 Shaft resistance development of P12 . . . . . . . . . . . . . . . . . . . . . . . 188
Fig. 6.17 Shaft resistance development of P66 . . . . . . . . . . . . . . . . . . . . . . . 189
Fig. 6.18 Shaft resistance development of P105 . . . . . . . . . . . . . . . . . . . . . . 189
Fig. 6.19 Shaft resistance distribution along the pile for P12 . . . . . . . . . . . . 191
Fig. 6.20 Shaft resistance distribution along the pile for P66 . . . . . . . . . . . . 192
Fig. 6.21 Shaft resistance distribution along the pile for P105 . . . . . . . . . . 193
Fig. 7.1 Failure of tension bar system in reaction pile (Handley
et al. 2006) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 198
Fig. 7.2 Platform bearing failures under Kentledge test (Handley
et al. 2006) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 199
Fig. 7.3 Plan view of the tested piles (not to scale) . . . . . . . . . . . . . . . . . . 200
Fig. 7.4 Subsurface conditions of P80/P60/P40 . . . . . . . . . . . . . . . . . . . . . 201
Fig. 7.5 Subsurface condition of P12 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 202
Fig. 7.6 Percussion of pile head . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 205
Fig. 7.7 Compressive SLT . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 205
Fig. 7.8 Equipment of SLT . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 206
Fig. 7.9 Concrete crack from pile head . . . . . . . . . . . . . . . . . . . . . . . . . . . . 207
Fig. 7.10 Deformed reinforcement . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 207
Fig. 7.11 Pile suffering from eccentric loads. a Bent reinforcement
of left part b Pile suffering from eccentricity c Fractured
concrete of right part . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 208
Fig. 7.12 Plunging failure of pile . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 208
Fig. 7.13 Q-s curve of P80 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 209
Fig. 7.14 s-lgt curve P80 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 210
Fig. 7.15 s-lgQ curve P80 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 211
Fig. 7.16 Interpretations a Double tangent method b Davisson’s
offset method c Chin’s method . . . . . . . . . . . . . . . . . . . . . . . . . . . . 212
Fig. 7.17 Q-s curve of P60 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 213
Fig. 7.18 s-lgt curve of P60 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 214
Fig. 7.19 s-lgQ curve of P60 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 215
List of Figures xxiii
Fig. 7.21 Q-s curve of P40 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 217

Fig. 7.23 s-lgQ curve of P40 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 218
Fig. 7.25 Q-s curve of P12 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 220
Fig. 7.27 s-lgQ curves of P12 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 221
Fig. 8.1 Settlement curves versus maintained time of P51 . . . . . . . . . . . . . 229
Fig. 8.2 Settlement curves versus maintained time of P121 . . . . . . . . . . . . 230
Fig. 8.3 Settlement curves versus maintained time of P126 . . . . . . . . . . . . 230
Fig. 8.4 Non-linear regression of P51 (loading stage) . . . . . . . . . . . . . . . . 231
Fig. 8.5 Non-linear regression of P121 (loading stage) . . . . . . . . . . . . . . . 231
Fig. 8.7 Non-linear regression of P51 (unloading) . . . . . . . . . . . . . . . . . . . 233
Fig. 8.8 Non-linear regression of P121 (unloading) . . . . . . . . . . . . . . . . . . 234
Fig. 8.9 Non-linear regression of P126 (unloading) . . . . . . . . . . . . . . . . . . 234
Fig. 8.10 Non-linear regression of single pile (Yang et al. 2012) . . . . . . . . 235
Fig. 8.11 Test and computed settlements from loading stages of P51 . . . . . 236
Fig. 8.12 Test and computed settlements from unloading stages
of P51 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 236
Fig. 8.13 Test and computed settlements from loading stages of P121 . . . . 237
of P121 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 237
of P126 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 238
Fig. 8.17 Vertical displacement of P15 under maintained load . . . . . . . . . . 239
Fig. 8.18 Vertical displacement of P16 under maintained load . . . . . . . . . . 240
of P15 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 244
of P16 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 245
Fig. 8.27 Borehole log 162 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 251
Fig. 8.28 Designed capacity versus double tangent interpretation . . . . . . . . 253
Fig. 8.29 Designed capacity versus modified double tangent
interpretation . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 254
xxiv List of Figures
Fig. 8.30 Allowable load versus double tangent interpretation . . . . . . . . . . 254

Fig. 8.31 Designed capacity versus modified Chin’s interpretation . . . . . . . 255
Fig. 8.32 Soil profile of Area A based on borehole logs . . . . . . . . . . . . . . . 255
Fig. 8.33 Soil profile of Area B based on borehole logs . . . . . . . . . . . . . . . . 256
Fig. 8.34 Soil profile of Area C based on borehole logs . . . . . . . . . . . . . . . . 256
Fig. 8.36 Non-linear regression of P15 (unloading stage) . . . . . . . . . . . . . . 259
Fig. 8.38 Non-linear regression of P163 (unloading stage) . . . . . . . . . . . . . 260
of P15 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 261
List of Tables
Table 2.1 Ultimate and peak shear stresses and normal stresses . . . . . . . . . 12
Table 2.2 Empirical values for ϕ,γ & I D of granular soils based
on corrected N-value (Bowles 1977) . . . . . . . . . . . . . . . . . . . . . . . 25
Table 2.3 Empirical values for unconfined compressive strength (qu )
and consistency of cohesive soils based on uncorrected
N-value (Bowles 1977) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 25
Table 2.4 Input factors for Brown’s method (Hannigan et al. 2006) . . . . . . 46
Table 2.5 Approximate range of β and Nt coefficients (Fellenius
1991) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 51
Table 2.6 Recommended safety factor (Hannigan et al. 2006) . . . . . . . . . . 52
Table 3.1 Simplified soil layers . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 85
Table 3.2 Reinforcement properties of tested piles . . . . . . . . . . . . . . . . . . . . 85
Table 3.3 Description of test piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 86
Table 3.4 Results of piles under different interpretation methods . . . . . . . . 97
Table 3.5 Static and dynamic load tests of compressive loaded piles . . . . . 104
Table 3.6 Static and dynamic load tests of uplift loaded piles . . . . . . . . . . . 105
Table 4.1 Load–settlement results of open-ended pipe pile (1) . . . . . . . . . . 114
Table 4.2 Load–settlement results of open-ended pipe pile (2) . . . . . . . . . . 115
Table 4.3 Load–settlement results of solid rectangular concrete pile . . . . . 116
Table 4.4 Soil parameters of subsurface layers . . . . . . . . . . . . . . . . . . . . . . . 117
Table 4.5 Strain gauge information . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 118
Table 4.6 Summary of capacity and interpretation of horizontal
loaded piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 133
Table 4.7 Summary of capacity and interpretation of uplift loaded
piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 134
Table 5.1 Results of piles under different interpretation methods . . . . . . . . 157
Table 5.2 Support installation information and excavation duration . . . . . . 158
Table 5.3 Properties of simplified soil layers . . . . . . . . . . . . . . . . . . . . . . . . 159
Table 5.4 Parameters and prices of reinforcements . . . . . . . . . . . . . . . . . . . . 162
Table 5.5 Deflection behaviours of CFRP- and steel-reinforced
concrete piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 170
Table 6.1 Information of test process . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 180
xxv
xxvi List of Tables
Table 6.2 Proportion of shaft and toe resistance under working load . . . . . 188
Table 7.1 Ultimate bearing capacity of piles . . . . . . . . . . . . . . . . . . . . . . . . . 203
Table 7.2 Information of tested piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 204
Table 7.3 Summarisation of the pile capacity . . . . . . . . . . . . . . . . . . . . . . . . 223
Table 8.1 Settlement coefficients of compressive loaded bored pile . . . . . . 232
Table 8.2 Test and computed settlements of P126 from loading
8,100 kN . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 239
Table 8.3 Settlement coefficients of uplift loaded bored piles . . . . . . . . . . . 241
Table 8.4 Shaft resistance, qsik . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 247
Table 8.5 End resistance, qpk . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 249
Table 8.6 Calculation of pile adjacent to borehole 162 . . . . . . . . . . . . . . . . 251
Table 8.7 Summarization of tested precast piles . . . . . . . . . . . . . . . . . . . . . . 252
Chapter 1
Introduction
1.1 Background
With the soaring requirement for building space in metropolises, high-rise buildings
are becoming increasingly popular. Pile foundations can resist more loading through
end bearing and friction resistance than can shallow foundations; hence, the use of
pile foundations is more common. As a structural element that transfers the loads
from upper structures into the soil layers, piles can be generally categorized into
precast piles and cast-in-situ piles. As a result of the numerous advantages—such
as convenience of construction without considering the transfer of piles, cost and
schedule of construction—cast-in-situ piles are the most accepted type of piles in
construction projects.
For bored piles, the process of construction leads to soil deposits remaining at
the base area after drilling the hole, which consequently leads to a decrease of the
pile capacity and increase of settlement. Engineers currently use admixture, such as
bentonite or polymeric slurry, to avoid the collapse of the soil during the drilling
process. However, these admixtures create a layer between the concrete pile surface
and soil layers. This new layer, created by the reaction among bentonite, soil and
water, has a low friction coefficient, which results in decreased shaft resistance. To
solve the above problems, the post grouting technique is the most acceptable method.
This method involves pressurizing admixtures into soils through a device, such as
a pipe installed inside the pile. The highly pressurized materials compact the loose
soils, and the injected materials mix with the soils and reinforce the pile end or shaft.
For precast piles, piles are driven into the soil layers and the soils are forced to
compact. Based on the amount of soil compacted, these precast piles can be catego-
rized into small and large displacement piles. Normally, open-ended pipe piles are
© Zhejiang University Press 2021 1

J. Zhou and E. Oh, Full-Scale Field Tests of Different Types of Piles,
Advanced Topics in Science and Technology in China 62,
https://doi.org/10.1007/978-981-33-6183-6_1
2 1 Introduction
considered small displacement piles, and solid piles are identified as large displace-
ment piles. A non-displacement pile generally refers to a bored pile (or cast-in-
place pile) because the soil is removed during construction. There are many inves-
tigates in terms of the small and large displacement piles, but the research of the
non-displacement precast piles is very limited.
When based on the selected type of materials, the pile foundation can be catego-
rized into timber, steel, concrete or composite piles. Past studies investigated Fiber-
reinforced polymer (FRP) composite piles with consideration of their damage or
corrosive effect on steel, timber, fiber and concrete materials. Different to the previous
research on FRP piles that are made of fiber and polymer, this book provides the
investigation on concrete piles with inside FRP bar reinforcement and outside FRP
laminar reinforcement.
Within this context, there are different types of piles used under different construc-
tion technology. To investigate these piles under different effects, such as post
grouting or composite confinement, it is vital to determine the ultimate bearing
capacity and settlement behavior of the pile foundation. Compared with experi-
mental tests in lab, field tests can provide the most accurate information about pile
behavior. For compressively loaded piles, the test method can include Static Load
Tests (SLTs), dynamic load tests, Osterberg Cell (O-cell) tests and statnamic tests.
However, for tests that apply lateral and uplift loading, SLTs are the only option.
In practice, static field tests are mostly the Proof Tests (PTs), which aims to assure
the designed pile foundation possessing adequate capacity within required settle-
ment. Field tests for loading the pile to failure are quite limited. The obtained result
sometimes may provide limited information, and is hard to be used for capacity
determination, thus interpretations based on the data obtained from field tests shall
be used.
This book focuses on field tests of different types of piles associated with various
up-to-date technologies. Static and dynamic load tests are performed to determine the
ultimate compressive bearing capacity and settlement behaviors of pile foundations.
However, the traditional SLT can only provide limited loads especially when testing
the super-long and large diameter pile foundations. So, this book also introduces
an improved SLT that can provide large compressive loads and the load transfer
mechanism research is involved. As aforementioned above, the PTs are the mostly
performed tests, thus this book not only covers the PTs, but also includes failure field
tests.
1.2 Objectives
The objectives of this book are to investigate bored piles with different grouting
technology; small, large and no displacement precast piles; FRP composite piles,
super-long and large diameter piles and piles applied with ultimate loading. A full-
scale field tests are illustrated including static compressive, uplift and horizontal load
tests, dynamic load tests, inclinometer monitoring as well as tests for determination
of the stress in pile foundation.
1.2 Objectives 3
For cast-in-situ concrete piles, this book aims to:

1. Determine the compressive and uplift ultimate bearing capacity of non-grouted,
base grouted only, and base-and-shaft grouted piles.
2. Research the shaft resistance mechanism of grouted piles through conducting
static and dynamic load tests.
3. Propose empirical formula for vertical settlement and uplift displacement
prediction.
4. Investigate the behavior of pile foundation under different types of failure
conditions, and compare failure load test with proof test.
5. Study the capacity as well as the load transfer mechanism of the super-long and
large diameter pile foundations.
For precast concrete piles, this book aims to:

1. Determine the ultimate bearing capacity of small and large displacement piles
(concrete pipe and rectangular piles).
2. Research the geotechnical mechanism of concrete pipe and rectangular piles.
3. Investigate the behavior of non-displacement piles through performing uplift and
horizontal SLTs.
4. Propose empirical formula for horizontal displacement prediction.
For composite piles, this book aims to:

1. Determine the ultimate bearing capacity of precast pipe and rectangular piles
confined by Carbon FRP (CFRP), and determine the capacity improvement
compared with piles without FRP confinement effect.
2. Research the deflection behavior of concrete bored pile with inside Glass FRP
(GFRP) reinforcement as a replacement for traditional steel bar reinforcement.
For the super-long and large-diameter concrete piles, this book aims to:
1. Research the load transfer mechanism, shaft resistance development and shear
stress distribution of the super-long and large diameter pile foundations.
2. Introduce an improved static load test which can apply large amount of
compressive load.
For piles applied with ultimate loads, this book aims to:
1. Determine the concrete pile behavior under failure caused by inadequate concrete
strength.
2. Investigate the failure of concrete piles because of the application of eccentric
loading, via performing field tests.
3. Study the plunging failure of a pile foundation.
4 1 Introduction
1.3 Book Organization
After the background and objectives have been illustrated in this chapter, the related
general academic principles and practices is provided in Chap. 2. Chapter 2 first
reviews the methods to determine the soil parameters and subsurface exploration,
and then reviews the categorization of piles. This is followed by discussion of the
analytical methods used to design the pile capacity, and review of field tests to indicate
which test methods should be selected to obtain the real pile capacity onsite.
Based on the research methodology presented in Chap. 2, Chap. 3 focuses on
studying concrete bored piles with consideration of the construction process, and
provides one case study. The post grouting technique effect is researched through
comparing base-and-shaft grouted, base grouted and non-grouted piles. The ultimate
compressive and uplift bearing capacities of these piles are determined through inter-
pretations of the result obtained from SLTs. Further, the shaft resistances of these
piles are analyzed and compared with the results obtained from dynamic load tests.
Chapter 4 investigates small, large and non-displacement concrete precast piles
through two case studies. For the small and large displacement piles, this chapter
focuses on determining compressive behaviors by considering various pile lengths
and concrete strengths. For the non-displacement precast piles, this chapter focuses
on determining the piles’ uplift and horizontal ultimate bearing capacity. This chapter
also analyses the uplift load transfer mechanisms of these non-displacement precast
piles.
Different to the investigations of bored piles (Chap. 3) and precast piles (Chap. 4),
Chap. 5 focuses on the composite piles with use of FRP materials. Chapter 5 presents
two case studies. The first case study researches the ultimate bearing capacity of
concrete piles with CFRP laminar confinement, and the second case study examines
the deflection behavior of passive bored piles in which steel reinforcement is replaced
by GFRP bars during the construction process.
Chapter 6 illustrates the improved compressive static load test which is used
for testing the super-long and large dimeter piles. Detailed pile design as well as
installation of the steel rebar meter (also known as strain gauge or sister rebar) are
demonstrated based on a case study. The load transfer mechanism, shaft resistance
development and shear stress distribution are discussed via comparing the results
from three super-long and large diameter piles.
As a result of SLTs mostly being used as PTs, which requires extrapolation of the
results, previous research on SLTs until failure is limited, and data demonstrating
plunging failure is rare. Thus, in Chap. 7, SLTs under different failure conditions
are performed, with consideration of the aspects of rigidity of concrete, soils and
eccentric loading.
Chapter 8 provides the capacity and settlement discussion of the different types
of concrete pile foundations obtained from performed field tests. Also, the methods
to obtain the empirical formula that are used for prediction of vertical and lateral
displacement of soil-pile system are proposed. Finally, Chap. 9 presents the study
findings, conclusions and recommendations based on all the case studies undertaken
in Chaps. 3 to 7. It also discusses potential avenues for further research.
Chapter 2
General Principles and Practices
2.1 General Introduction
All types of engineers are required to have sophisticated understanding and knowl-
edge of subsurface conditions to undertake their projects. Soil analysis is more
complicated than analysis of other materials because of soil’s non-continuum char-
acteristic. Materials such as steel and concrete are relatively uniform solids, and their
materials analysis can be based with a high degree of confidence on the assumption
of the materials’ solid mechanics and strength (Samtani and Nowatzki 2006a). In
contrast, soil materials can widely differ over time and space. Further, many factors
can influence the behavior of soil, such as soil particles and climatic issues.
The constitutive behavior of soil, which indicates the strength and stiffness prop-
erties, can be obtained through conducting laboratory and in-situ tests. Commonly,
the laboratory tests include consolidation, direct shear and triaxial tests. Given that
the change of soil in volume is a time-dependent process, determining the degree
of consolidation, U v , and coefficient of consolidation, C v , are very important for
analysis of the total settlement of soil. Further, the shear strength of soils is another
significant aspect used for foundation design. This chapter first reviews the literature
referring to direct shear, oedometer and triaxial tests. The parameters of cohesion, c,
and friction angle ϕ, of the soil can be obtained from both direct shear and triaxial
tests. Direct shear tests are more usual and are easy to handle, yet triaxial tests
allow the soil to be consolidated or drained first; therefore, engineers must assess
the projects’ requirements to appropriately select and perform the tests. Note that,
for the laboratory index, this book does not review factors such as moisture content
(w), density (ρ), unit weight (γ ), void ratio (e), coefficient of uniformity (C u ) and
curvature (C z ), plasticity (I P ) and liquidity index (I L ).
Laboratory tests, sometimes however, are valueless for quantifying the mechanical
behavior of an element of soil. For example, it is difficult and expensive to determine
accurate data from an undisturbed deposit, such as sand and sensitive clays. Further,
the response of a small element of soil that contains features within the macro-fabric

https://doi.org/10.1007/978-981-33-6183-6_2
6 2 General Principles and Practices
cannot represent the complete soil behavior (Knappett and Craig 2012). The principal
in-situ tests which include the Standard Penetration Test (SPT), Cone Penetration
Test (CPT), field vane test and pressuremeter test could cover the shortage of the
lab tests. These tests are related to the corresponding response of soil under direct
loading to determine the soil properties onsite instead of soil transfer. Boreholes are
required to conduct SPTs, others such as CPT and field vane test that do not require
boreholes are considered as ‘direct-push’ technologies.
Analytical methods indicate the procedure to obtain the capacity of a pile for
design, based on the parameters acquired from laboratory and in-situ tests. Meyerhof
(1976), Nordlund (1963) and Brown et al. (2001) proposed methods for estimating of
the bearing capacity of a single pile in cohesionless soils. For a single pile in cohesive
soils, Tomlinson (1980) and Fellenius (1991) provided the total and effective methods
to determine the capacity. All of these methods are now widely accepted by engineers,
and are discussed follows by the review of subsurface explorations in Sect. 2.2.
It should be noted that analytical methods are used for primary design, and this
is the theoretical value of pile capacity. Static Load Test (SLT) is the most accurate
method to determine the load capacity of pile foundations. SLT involves loading from
the pile head with specific time intervals, and monitoring the displacement of the
pile head. These tests can be categorized into vertical compressive and uplift testing
and lateral testing. Detailed information is provided in the standards AS 2159 (Stan-
dards Australia 2009), ASTM-D-1143/D-07 (ASTM International 1994), ASTM-D-
3966/D-07 (ASTM International 1995a) and ASTM-D-3689/D-07 (ASTM Interna-
tional 1995b). However, performance of these compressive, uplift and lateral SLTs
until failure load is expensive and time consuming; hence, the most common method
is to twice apply the designed load and determine the pile head displacement, which
is called a ‘proof test’. The purpose of this test is to confirm that the foundation
movement is within the standard requirement to safely support the designed loading.
However, this test cannot directly provide the ultimate bearing capacity of a pile.
Various methods can be used for interpretation of SLTs to determine the ultimate
bearing capacity of piles. Section 2.5 reviews the methods for extrapolations of SLTs,
such as those by Davisson (1972), Butler and Hoy (1976), DeBeer (1970), Brinch-
Hansen (1963) and Chin (1970). Further, this same section provides the theory, test
setup, and data analysis methods of the dynamic load test, Osterberg cell test and
statnamic test that are also used to determine the capacity of single piles.
2.2 Soil Stratum Determination
2.2.1 Laboratory Tests
2.2.1.1 Direct Shear Tests
The shear strength parameters of soils are important for foundation design, and the
soil strength contains frictional strength and cohesive strength. Frictional strength is
2.2 Soil Stratum Determination 7
dependent on the stress state, such as overburden pressure and the friction angle
between soil particles. Based on the experiments, Coulomb (1776) proposed a
formula to determine the shear resistance of soil, which is related to normal stress,
σ and the internal friction angle, ϕ, as provided in Eq. 2.1. It can be seen that soil
strength is related to friction between soil particles because tan ϕ is equal to the
coefficient of friction, μ. However, soil strength also relates to drainage of water, as
Terzaghi (1944) indicated. So, the soil strength is also represented by the effective
stress and friction angle provided in Eq. 2.2:
τ f = c + σ tanϕ (2.1)

τ f = c + σ tanϕ (2.2)
where:
τ f = shear resistance of soil.

c and c = total and effective cohesion.

σ and σ = total and effective normal stress.

ϕ and ϕ = total and effective friction angle.
The parameters of shear strength can be determined by conducting laboratory
direct shear and triaxial tests. This section provides the setup, procedures and data
analysis of direct shear tests based on AS 1289.6.2.2 (Standards Australia 1998a) and
ASTM-D-3080–98 (ASTM International 2007a). The basic theory involves applying
horizontal shear force to the sample, which is vertically pressured and confined in
a metal box (Fig. 2.1). The horizontal and vertical displacement of the sample are
recorded under a certain vertical stress, and then the peak and ultimate shearing
strength can be determined through plotting horizontal shear stress versus displace-
ment, as illustrated in Fig. 2.2. These tests can also indicate the soil displacement
behavior until failure condition. Through changing the value of vertical normal stress
Fig. 2.1 Direct shear apparatus-shear box

Fig. 2.2 Typical results from drained direct shear test (AS 1289.6.2.2, 1998)
and repeating the test, the peak and ultimate friction angle can be obtained after plot-
ting normal stress versus shear stress, as shown in Fig. 2.3. It should be noted that,
for the pure saturated sand sample, the cohesion value is zero.
Fig. 2.3 Determination of

friction angle from direct
shear test (AS 1289.6.2.2,
1998)
Two perforated grid plates

Retaining plate
Loading pad
Bottom half of shear box Upper half of shear box

Two screws
Fig. 2.4 Components of shear box
To perform the tests, it is first necessary to prepare the sample in the shear box.
Figure 2.4 shows the components of this box. The upper half of the shear box is
placed on the lower half, ensuring that all the inscribed holes match. Then, the two
clamping screws are placed in their proper holes (clearance holes) to hold the two
shear box halves in place.
The retaining plate at the bottom of the shear box is fixed, and then the perforated
grid plate is fixed, noting that the serration orientation should be perpendicular to the
shear force. A spoon is used to transfer the sample into the shear box several times,
noting that several blows need to be applied with a tamping tool to mechanically
dense the sand (20 times should be adequate, keeping the tamping light and uniform).
The height clearance from the lid of the chamber to the final sand level should be
measured four times near each wall of the shear box. This measured height should
be more than 5 mm.
The perforated grid plate is fixed on top (minding the serration) with the loading
pad, and then the assembly is placed into the shear box container, as shown in Fig. 2.5,
slotting the end onto the push rod. Later, it is necessary to ensure that the lever beam
is horizontal at a free position, and the required surcharge weight is added to produce
the required normal stress, and then the jack screw is slowly released.
The horizontal and vertical Linear Variable Differential Transformers (LVDTs)
and load cell are fixed, ensuring that the equipment is alive. The connecting bolts
are removed to avoid the loading being resisted by bolts, and the lifting screws are
Impact data logger
LVDTs
Load cell
Direct digital shear machine
Shear box container Loading yoke
Lever beam
Load hanger
Jack screw
Lever hanger
Using weight to provide

normal stress
Fig. 2.5 Direct shear apparatus
wound down a ‘half a turn’ each to separate the upper and lower halves by about
0.5 mm. These are then backed off a few turns.
Trial direct shear tests were conducted at Griffith University. The vertical stresses
applied were 50, 150 and 250 kPa. The plotted horizontal movement versus shear
stress diagram, horizontal versus vertical displacement diagram, and normal stress
versus shear stress diagram are provided in Figs. 2.6, 2.7 and 2.8, respectively.
Table 2.1 summarizes the data obtained from these tests. The parameters of peak
and ultimate friction angles were determined as 42° and 38°, and the cohesion was
determined as 5.25 and 2.59 kN/m2 respectively.
2.2.1.2 Consolidation Tests
A fully saturated soil with a property of low permeability will gradually decrease its
volume under loading because of change in effective stress. This soil behavior may be
due to a reduction in pore water pressure or drainage of pore water pressure, and this
process is called ‘consolidation’. In contrast, the process of swelling is an increase
in soil volume under loading because of negative excess pore water pressure. This
section reviews the one-dimensional consolidation of soil, where loading increment
is applied vertically, and the lateral strain is restricted.
250
200
Shear Stress (kN/m/m)
150
50kPa
150kPa
100 250kPa
50
0
0 1 2 3 4 5 6 7
Horizontal Displacement (mm)
Fig. 2.6 Horizontal movement versus shear stress
0.8
Vertical Movement (mm)
0.6
50kPa
0.4 150kPa
250kPa
0.2
0
0 1 2 3 4 5 6 7
-0.2
Horizontal Movement (mm)
Fig. 2.7 Horizontal movement versus vertical movement

300
250
y = 0.9014x + 5.2546
Shear Stress (kN/m/m)
200 R² = 0.9978
150
UlƟmate Shear Strength

100
y = 0.6833x + 2.5926 Peak Shear Strengh
50 R² = 0.9921
Linear (UlƟmate Shear Strength)
Linear (Peak Shear Strengh)

0
0 50 100 150 200 250 300
Normal Stress (kPa)
Fig. 2.8 Normal stress versus shear stress
Table 2.1 Ultimate and peak

Ultimate shear stress Maximum shear stress Normal stress
shear stresses and normal
stresses 40.28 52.78 50
98.06 135.56 150
176.94 233.06 250
As shown in Fig. 2.9, the height of an element will decrease from initial height
of H0 to H1 when vertical stress is applied, and, at this stage, the void ratio, e, can be
determined. By plotting e under each load stage versus the vertical effective stress,
σ , and in logarithmic scales, the compressibility characteristics can be preliminarily
determined, as shown in Fig. 2.10.
The e-log σ curve illustrates a one-dimensional compression line (1DCL), as
shown in Fig. 2.10. The slope of this virgin compression line is called a ‘compression
index’, with a symbol of C c , and the slope of unload–reload is called the ‘expansion
index’, with a symbol of C e or C r (re-compression index). The compression index
can be determined as follows:
e0 − e1
Cc = (2.3)
σ1 − σ0

where e0 and σ0 are the initial void ratio and initial stress.
The coefficient of volume compressibility, mv , is defined as the volume change
per unit increase in effective stress per unit volume. It is related to e0 and e1 , as
illustrated in Eq. 2.4. The constrained modulus or one-dimensional elastic modulus,

E oed is the reciprocal of mv :
Fig. 2.9 Basic theory of element under stress
Fig. 2.10 Void ratio–effective stress relationship (Craig 1983)

1 e0 − e1
mv = (2.4)
1 + e0 σ1 − σ0
The consolidation of soil mass involves a time-dependent change in volume. The

settlement of soil under stress refers to the readjustment of soil particles resulting
from dissipation of water. This process is known as ‘primary consolidation’. It has

been reported that, the greater the initial void ratio, the more water will be squeezed
out, and the greater the primary consolidation will be (Samtani and Nowatzki 2006a).
After the primary consolidation process is finished, the soil particles reorient their
position or deform under constant load, which leads to secondary compression. This
process is known as the ‘creep behavior’ of soil. Given that the second compression
is also a time-dependent process, as with the primary consolidation process, it is vital
to plot the diagram with time or lg time to determine the soil properties.
Traditionally, consolidation tests which are also known as oedometer tests are
used to determine the soil characteristics. The consolidation tests are governed by
AS 1289.6.6.1 (Standards Australia 1998b) and ASTM-D-2435–06 (ASTM Inter-
national 2007b). The equipment used to perform oedometer tests contains a load
device (Fig. 2.11a) that can provide the required vertical loads; a consolidation cell
(Fig. 2.11b) used to hold the soil specimen that is confined by a ring; a monitoring
device, such as dial gauges or LVDTs to obtain the deformed height of the soil
element; and a stopwatch to record time. The configuration of a consolidation cell
may be different, yet normally contains two porous stones, a load cap and a confining
ring.
To obtain the plotted compression versus time diagram depicted in Fig. 2.12, an
oedometer test should be performed. The procedure for this test is as follows:
A cutting ring is used to prepare the soil sample and make the top and bottom
of the sample confined by the consolidation ring, also, avoid any compaction on the
sample. Later, filter paper is placed on the top and bottom as protection to stop the
drainage plates from being contaminated. The filter paper size must be appropriate,
and the paper should avoid contacting the ring.
(a) Loading device (b) Consolidation cell (Craig, 1983)
Fig. 2.11 Apparatus of consolidation test

Fig. 2.12 Typical compression versus time (AS 1289.6.6.1, 1998)
The porous disc is then located, which should be soaked in water for 24 h afore-
hand, and then the prepared samples (with ring and filter paper) are carefully placed
on this porous disc. Another porous disc is added on top of the samples, and the
cutting ring is clamped. Later, the top cap is positioned. The cell is placed on the
oedometer, and the loading yoke is placed onto the top cap. There must be contact
without any extra loading applied. Later, the cell is filled with water to allow the
sample to be fully saturated. The dial gauge or LVDT is set up to record the vertical
settlement of the sample under loading.
A one-dimensional consolidation trial test of fine clay was conducted at Griffith
University, and the results are provided in Fig. 2.13. Based on the Casagrande (1936)
method or log time method, the theoretical curve contains three parts. The first part
is approximately parabolic, the second part is closely linear, and the third part is an
asymptote referring to the horizontal axis at U v = 100%.
As depicted in Fig. 2.13, the dial gauge reading of a0 is related to the initial
reading, and as corresponds to U v = 0. Through finding the intersection of the two
tangent lines, as shown in Fig. 2.13, the a100 can be determined, and the gauge reading
that represents a50 and t 50 can be determined later. The compression between as to
a100 is known as primary consolidation. The time factor, T v , can be determined once
the t 50 value is determined, and the C v can then be determined via Eq. 2.5. Note
that the d can be taken as half of the average thickness of the specimen because the
oedometer test is a two-way drainage condition.
The theory curve that plots the diagram
√ of root time versus dial gauge reading is
linear up to 60% consolidation, and t90 is used for extrapolation of the linear part
Fig. 2.13 The log time method
of the curve (Taylor 1948). As shown in Fig. 2.14, similar to the Casagrande method,
the settlements between a0 and as represent initial compression, the settlements
between as and a100 give primary consolidation and the settlements between a100 and
af illustrate secondary compression. The coefficient of consolidation is governed by
Eq. 2.6:
0.196d 2
Cv = (2.5)
t50
0.848d 2
Cv = √ (2.6)
t90
Referring to these two methods, the initial compression ratio, primary consolida-
tion ratio and second consolidation ratio can be expressed by Eqs. 2.7–2.9. Using the
data from this test, the relationship between the void ratios versus vertical pressure
can also be found with Fig. 2.15:

Initial compression ratio: r 0 = (a0 − as )/ a0 − a f (2.7)

Primary compression ratio: r p = (as − a100 )/ a0 − a f (2.8)

Secondary compression ratio: rs = 1 − r0 + r p (2.9)
Fig. 2.14 The root time method
0.06
0.05
0.04
Void Ratio
0.03
0.02
0.01
0
0 200 400 600 800
Vertical Effective Stress (kPa)
Fig. 2.15 Void ratios versus vertical pressure of consolidation tests

2.2.1.3 Triaxial Tests
The disadvantage of direct shear tests is that the shear failure surface is predeter-
mined as a horizontal plane. Further, direct shear tests are mostly performed on a
drained specimen because the apparatus cannot control drainage of water. For clay
that contains water, triaxial tests are the traditional method to determine the strength
parameters.

With a soil element in failure state, the principal stresses σ1 and σ3 are applied
vertically and horizontally, as shown in Fig. 2.16. Shear stress and normal stress act
on the failure plane. Through geometrical relationship in triangle abc, and establish
force equations of equilibriums:

σ3 ds sinθ − σ f ds sinθ + τ f d s cosθ = 0 (2.10)

σ1 d s cosθ − σ f d s cosθ − ds τ f sinθ = 0 (2.11)

By moving Eq. 2.11 into Eq. 2.10, stresses σ f and τ f are provided in Eqs. 2.12
and 2.13:
Fig. 2.16 Theory of soil

element at failure state
1
1
σf = σ1 + σ3 + (σ1 − σ3 )cos2θ (2.12)
2 2
1
τf = (σ − σ3 )cos2θ (2.13)
2 1

The relationships between σ f and τ f and σ1 and σ3 can be provided by a Mohr–
Coulomb diagram, as shown in Fig. 2.17. The failure envelope represents the function,
as previously provided in Eq. 2.2. Through the geometrical relationship, it can be
found that:

1
1
AD = RDsinϕ = c cotϕ + σ1 + σ3 sinϕ = (σ1 − σ3 ) (2.14)
2 2
Equation 2.14 can be rearranged into Eqs. 2.15 and 2.16 for clay as follows:

ϕ ϕ
σ1 = σ3 tan2 (450 + ) + 2c tan(450 + ) (2.15)
2 2

ϕ ϕ
σ3 = σ1 tan2 (450 − ) − 2c tan(450 − ) (2.16)
2 2
For cohesionless soil, such as sand, c = 0, the principle stresses can be determined
as follows:

ϕ
σ1 = σ3 tan2 (450 + ) (2.17)
2
Fig. 2.17 Mohr–Coulomb failure criterion


ϕ
σ3 = σ1 tan2 (450 − ) (2.18)
2
As reviewed above, for one type of soil, there will be a point that represents
the failure condition under particular principle stresses from the Mohr–Coulomb
diagram (Point A in Fig. 2.17). By changing the state of stress, a series of points
can be determined, all of which are in the failure envelope function (when soil
fails). Triaxial tests can determine certain parameters (such as cohesion and friction

angle) through applying different stresses vertically (σ1 ) and horizontally (σ3 ) until
the test sample reaches failure and plotting Mohr–Coulomb diagrams. The primary
advantage of triaxial tests is that they are suitable for all types of soils because they
allow saturated soil with relatively low permeability to be consolidated. Compared
with direct shear tests, this test has a device to control the drainage of soil, and the
amount of water drained out of soil can be measured.
Triaxial tests are governed by AS 1289.6.4.1 (Standards Australia 2016a) and AS
1289.6.4.2 (Standards Australia 2016b), and the equipment is provided in Fig. 2.18.
The soil sample is first prepared in a cylinder and protected by a rubber membrane.
Later, hydraulic pressure is applied on this sample, and this pressure is simultaneously
applied horizontally and vertically. Then, the vertical deviation stress is applied
through the loading ram onto the loading cap until shear failure occurs.
Loading ram
Loading cap
Perspex Cylinder
Porous disc Soil sample (inside

rubber membrane)
All-round hydraulic
pressure supply
Drainage or pore water pressure measurement
Fig. 2.18 The triaxial apparatus

For the Unconsolidated Undrained (UU) test, the drainage and consolidation
process are not allowed during the application of confining or axial pressure. The
performance of a UU test can determine the undrained shear strength, and can model
the response of soil applied with a rapid axial load. Different to this test, the Consoli-
dated Drained (CD) test can allow the consolidation of soil while confining pressure
before the axial pressure is applied. During this test, the loading rate is slow enough
to ensure there is no excess pore water pressure built. The CD test simulates the
long-term drained condition thus the effective strength of soil can be determined.
This equipment can also perform the Consolidated Undrained (CU) test. Similar to
the CD test, the sample is consolidated first, yet the vertical pressure is applied rela-
tively quickly. The amount of pore pressure can be measured during CU tests, so
these tests can obtain the effective and total strength parameters of soil.
A UU trail test was conducted in a laboratory at Griffith University, with confine-
ment pressures of 150, 300 and 600 kPa. While the vertical deviation pressure was
applied, the strain data were determined, and the results are provided in Fig. 2.19.
The normal stresses that represented the failure of soil were determined as 789, 1,160
and 1,608 kPa. When these values were plotted into the Mohr–Coulomb diagram,
the function of the failure envelope could be determined (Fig. 2.20) and the param-
eters of soil strength could be obtained later. As shown in Fig. 2.20, the intersection
between this function (failure envelope) and y-axis gave the cohesion, and the slope
of this function was the friction angle.
1200
1000
Deviator Stress (kPa)
800
UU1
600
UU2
400 UU3
200
0
0 2 4 6 8 10 12
Strain (%)
Fig. 2.19 Set of three triaxial tests

800
600
400
Shear Stress (kPa)
200
0
0 500 1000 1500 2000
-200
-400
-600
-800
Vertical Stress (kPa)
Fig. 2.20 Mohr–Coulomb circle of triaxial tests
2.2.2 In-Situ Tests
2.2.2.1 Standard Penetration Test (SPT)
In 1902, the SPT was introduced by the Raymond Pile Company, and it is now one
of the most widely used in-situ tests because of its low cost, simplicity and ability
to be rapidly conducted. As shown in Fig. 2.21, in this test, a split barrel sampler
is driven into the soil by a drop hammer (weight of 63.5 kg) with falling height of
30 in (760 mm) to achieve to a 6 in (150 mm) penetration, this process is known as
‘seating’ process. After the second and third penetration finished (6 in, respectively),
the number of drops (penetration depth of 300 mm) of the hammer is recorded. This
is known as the uncorrected N-value or SPT blow count.
There are three common types of drop hammer (safety, donut and automatic
hammer) and four types of drill rods for performing the tests, which consequently
lead to the amount of energy being transferred to the soil during the SPT. The actual
energy transferred is less than the theoretical energy, which is influenced by numerous
factors, such as eccentric loading and frictional losses. The constitutive properties of
the soils cannot vary with the equipment used; thus, the N-value should be corrected
to N 60 , representing a standardized Energy Ratio (ER) of 60%. This corrected N-
value or N 60 can be determined through Eq. 2.19 with consideration of borehole
diameter, rod length and energy ratios.
Fig. 2.21 Standard Penetration Test schematic (Mayne et al. 2002)

ER
N60 = N (2.19)
60
The N-values of similar materials increase with increasing overburden pressure

(σvo ) at the test depth. For coarse-grained soils, such as sand and gravel, the N 60 is
conventionally normalized to (N 1 )60 , which refers to 1 atmosphere effective over-
burden stress. This is governed by Eq. 2.20, while the overburden correction factor
is provided in Eqs. 2.21 (Liao and Whitman 1986) and 2.22 (Peck et al. 1974), and
Figs. 2.22 (Skempton 1986) and 2.23 (Samtani and Nowatzki 2006a).
(N 1 )60 = C N N60 (2.20)

pa n
CN = (2.21)
σV 0

40
C N = 0.77log ≤2 (2.22)
σV 0
Fig. 2.22 Overburden correction factor (Skempton 1986)
Fig. 2.23 Variation of overburden correction factor CN (Samtani and Nowatzki 2006a)
Table 2.2 Empirical values for ϕ, γ & I D of granular soils based on corrected N-value (Bowles
1977)
Description Very loose Loose Medium Dense Very dense
Relative density I D 0–0.15 0.15–0.35 0.35–0.65 0.65–0.85 0.85–1.00
Corrected SPT N 0–4 4–10 10–30 30–50 50 +
Friction angle ϕ 25–30° 27–32° 30–35° 35–40° 38–43°
Unit weight γ (kN/m3 ) 11.0–15.7 14.1–18.1 17.3–20.4 17.3–22.0 20.4–23.6
Table 2.3 Empirical values for unconfined compressive strength (qu ) and consistency of cohesive
soils based on uncorrected N-value (Bowles 1977)
Consistency Very soft Soft Medium Stiff Very stiff Hard
qu (kPa) 0–24 24–48 48–96 96–192 192–384 384 +
SPT N-value 0–2 2–4 4–8 8–16 16–32 32 +
Unit weight γ 15.8–18.8 15.8–18.8 17.3–20.4 18.8–22.0 18.8–22.0 18.8–22.0
(saturated) kN/m3
where:
(N 1 )60 = SPT N-value corrected for energy and overburden pressure.
N60 = SPT N-value corrected for 60% energy transfer.
C N = overburden correction factor.
Pa = atmospheric pressure.

σV 0 = effective vertical stress at the sample depth.
n = exponent typically equal to 1 in clays and 0.5 in sandy materials.
The corrected and uncorrected N-values can be used to determine not only the
relative density, I D , and undrained shear strength, C u , but also the angle of internal
friction, ϕ; unit weight, γ ; and unconfined compression strength, qu , as provided
in Tables 2.2 and 2.3. It should be noted that, for soils that contain gravel-size
particles, Table 2.2 may be unreliable, and the correlations can only be used as a
rough estimation. Further, Table 2.3 should only be used for preliminary design
purposes because the unconfined compressive strength of cohesive soils is crude and
unreliable (Hannigan et al. 2006).
2.2.2.2 Cone Penetrometer Test (CPT and CPTu)
The cone penetrometer is one of the most versatile tools available for soil exploration
(Lunne et al. 1997). Tests are mostly performed using a CPT test rigs truck, as shown
in Fig. 2.24. This rig has a hydraulic jack inside the truck, and the weight of the truck
itself is used as a reaction mass to push the CPT rod into the ground. During this
test, a penetrometer consisting of a friction sleeve and cone is used to determine
the tip (qc ) and shaft (f s ) resistance during penetration. Performing CPT aims to
Fig. 2.24 Schematic of CPT truck (Samtani and Nowatzki 2006a)
Fig. 2.25 Soil behaviour type classification chart (Robertson 1990)
identify the stratigraphy and provide soil layer information. Based on the calculation
as illustrated in Fig. 2.25, the property of soil can be obtained. It can also be used for
deep foundation design because the CPT process is similar to driven pile installation.
Fig. 2.26 Piezocone measurement (CPTu ) (Samtani and Nowatzki 2006a)
The cone tip resist, qc , is determined by F c /Ac (force and area of the piezocone),
and the friction resistance, f s , obtained from the friction sleeve is determined by
F s /As (force and area of the sleeve). For a standard CPT, the soil tip resistance qt
is equal to cone tip resistance qc . As shown in Fig. 2.25 (left side of the diagram),

after determined the normalized cone resistance ((qt − σvo )/σvo ) and friction ratio
( f s /(qt − σvo )), the soil classification can be accomplished after finding the zone in
figure.
However, for fine-grained soils, the excess pore water pressure should be consid-
ered; thus, a more sophisticated device is required. As shown in Fig. 2.26, a piezocone
can be used because this device can measure the excess pore water pressure (CPTu ).
If this equipment is used, pore water ratio ((u − u o )/(qt − σvo )) is used for soil
classification as illustrated from the right side of Fig. 2.25. CPTu tests are faster and
cheaper than drilling and sampling. They also provide more consistent and reliable
results than SPT N-values due to the load condition is easier to control. The disad-
vantage of CPTu is that it requires skilled operator to run the test, and no soil samples
are obtained.
Typical result of CPTu includes the shaft and tip resistance as well as water
pressure shown in Fig. 2.27. Also, CPTu data can be used to interpret the peak angle

of shear resistance (ϕmax ), relative density (I D ), undrained shear strength (C u ) and
OCR based on proposed equations and diagrams (Jamiolkowski et al. 2003; Mayne
2007). There has been intensive research and tests performed to predict the capacity
of piles based on CPT data. Abu-Farsakh et al. (2017) indicated that the Schmertmann
(1978) method; De Kuiter and Beringen (1979) method; LCPC method (Bustamante
and Gianeselli 1982); and recently developed method by Lehane et al. (2013) can all
provide better results for pile capacity than can the other CPT methods.
Fig. 2.27 Piezocone results for Apple Freeway Bridge (Samtani and Nowatzki 2006a)
2.3 Types of Piles
2.3.1 General Pile Categorization
Pile foundation or deep foundation is the slender structural element that transfers
loading from the upper structure to the soil (Tomlinson 2001). Compared with shallow
foundations, there are situations in which the use of pile foundation is preferred, such
as when the pile is designed to resist large concentrated loads, when the near surface
of soil layers is not stiff, and when displacements should be kept small for settlement-
sensitive buildings. There are numerous types of pile foundation, and they can be
broadly categorized into concrete, steel, timber and composite piles, based on the
materials used. A detailed classification system of pile foundations is provided in
Fig. 2.28, referring to the installation methods and equipment used for installation.
Timber piles have been used for over a millennium. Timber is an ideal selection
for piling when used for cohesion pilling or for resisting light loads. The advantages
of timber include its high strength to weight ratio, ease of handling and ability to be
cut during construction (Tomlinson and Boorman 2001). Timber piles are suited for
a modest load because this material is vulnerable and can be easily damaged at the
pile head and toe when driven hard. In addition, when pile foundations are required
to be driven into dense sand or gravel or to be used as end bearing piles on rocks,
2.3 Types of Piles 29
Fig. 2.28 Pile classification chart (Hannigan et al. 2006)
the use of a timber pile can cause excessive bending, and cracking may occur. Under
such situations, a steel pile is preferred because of its properties of a relatively small
cross-section area and high strength. The configurations of the steel pile include H-
piles, pipe piles and screw piles. H-piles with high strength and improved corrosion
resistance in atmospheric and marine environments are detailed in ASTM A588 and
ASTM A690, respectively. However, these H-piles are difficult to obtain.
The most common pile foundations used currently are concrete piles. These piles
can be categorized into bored piles (cast-in-situ piles) and precast piles. Precast piles
can be classified into prestressed concrete piles and reinforced concrete piles. The
configurations of precast piles and reinforced concrete piles are similar. The differ-
ence between these two types of piles is that pretensioned or post-tensioned steel
reinforcements are used in prestressed piles, and, because of using these reinforce-
ments, prestressed concrete piles are more resistant to weathering and corrosion
under continued compression loads (Hannigan et al. 2006). From another perspec-
tive, prestressing or tensile stress results in damage to prestressed piles during driving
and handing. Compared with precast piles, the advantage of bored concrete piles is
that there is no need to consider the piles’ transfer to the construction site, and no
need to predetermine pile lengths. However, casting onsite may lead to dust pollution
when mixing concrete and welding the reinforcement cage.
The pile foundation can also be categorized into compressive, uplift and lateral
loaded piles based on the loading conditions. Figure 2.29 depicts the loading situa-
tions of pile foundations. Normally, a pile foundation can reach the bearing stratum
and loads can be vertical and horizontal (Fig. 2.29a, d). In addition to the individual
types of loading or combination loading, pile foundations can also be needed when
there is a scour around effect, liquefaction effect or seismic event (Fig. 2.29e, h).
Further, a pile foundation may be used as a fender system or retaining structure
to resist earth pressure caused by excavation and undesirable movement caused by
downdrag (Fig. 2.29i, k).
2.3.2 Cast-In-Situ Piles
Pile foundations perform well in high-rise buildings because of their high value of
bearing capacity. However, during the process of cast-in-situ construction, some soil
deposits may remain at the base area after drilling the hole. These soil deposits are
caused by a collapse of soil layers and are very difficult to remove from the drilled
hole. This leads to decreased end resistance capacity of the pile and increased pile
settlement. As a result of these remaining soil deposits from the base of the hole,
SLTs conducted in Taiyuan city, China, found that the capacity and settlement of
some piles from a mansion were much lower than the designed requirements (Shi
2005). Further, the drilling operation loosening the soil underneath the base of the
bored hole also leads to excessive working load settlement.
Another problem associated with the bored pile is the use of admixtures, such as
bentonite or polymeric slurry (slurry support), which are used to avoid the collapse of
soil layers into bored holes. The use of these materials may lead to capacity decrease
of the pile shaft because these admixtures combine with soil and water to create a
layer between the soil and pile, which consequently leads to a decrease in the friction
resistance of the pile. It has been reported that this admixture layer or composite layer
decreases 30 to 40% of the pile-bearing capacity (Li et al. 2000).
Such limitations of cast-in-situ concrete piles can be handled by grouting tech-
niques. Grouting equipment can be categorized into a flat jack system, which consists
of grout delivery pipes connected to a steel plate with a rubber membrane and sleeve-
port system that comprises two to four U-tubes installed at the bottom of the pile. This
Fig. 2.29 Situations in which deep pile foundations may be needed. (modified from Vesic 1977)
U-tube is covered by rubber and can be arranged in various configurations (Dapp

et al. 2006).
Cement is a widely used grouting material. It is blended into various ratios and
then injected into the pile toe under high pressure, transferred by a grouting tube.
Through this technology, the grouting application restores the original density of the
base soil and reduces the settlement of the pile when loading transfers from upper
structures (Tomlinson and Boorman 2001). Moreover, as shown in Fig. 2.30, the
grouted cement can force out the soil-admixture layer that will lead to a decrease of
the shaft resistance, and create a new solid cement layer, this layer bonds the soil
as well as the concrete surface, and consequently improves the friction between soil
layers and pile surface (Zhou et al. 2017a).
Fig. 2.30 Cast-in-situ concrete pile using slurry support technology
Numerous studies have been conducted to determine the effectiveness of this tech-
nology. Field SLTs were conducted to compare the behaviors of one base grouted
pile with two conventional slurry stabilized piles (no grouting being applied). These
tests provided the load transfer characteristics and clarification of the mechanism of
the base grouted concrete pile. They also provided the correlation among cement
consumption, the number of grouting stages and the volume of a pile shaft (Ho
2003). The diameter of piles with grouting techniques ranges from 0.4 m to approx-
imately 2.5 m. Tests of two piles were conducted at Paksey Bridge over the Padma
(Ganges) River in western Bangladesh, with tested piles of 1.5 m in diameter (Wilkins
and Castelli 2004). The construction of large diameter grouted piles (diameters of
2.4 m) commenced with The Pinnacle—a 290 m high skyscraper in London, England
(Patelet al. 2015).
Besides analytical methods, the finite element method has also been used to
determine the bearing capacity for grouting treated piles. Recent tests and numerical
simulation have shown that post grouted concrete piles can double the ultimate
capacity of a defected pile, and increase about 20% of a traditional pile’s capacity
(Nguyen et al. 2012). Several case history projects with application of base grouting
techniques were provided by Sinnreich and Simpson (2013). In their paper, they
used the O-cell test method which can provide a bidirectional axial compressive
load for the determination of the shaft-and-end grouted pile capacity. However, the
results were inconclusive because some projects illustrated an increase of grouted
pile capacity and some projects did not.
Pressure grouting piles or post grouting piles have been successfully employed
around the world for about 40 years (Lai et al. 2000). However, most previous grouting
pile investigations concerned base grouting; therefore, research into shaft grouting is
very limited. Moreover, research has seldom considered the ultimate uplift bearing
capacity of grouted bored piles under static and dynamic load tests. As Sinnreich
and Simpson (2013) stated, further research into the mechanics of grouting bored
piles is needed, as some results of ultimate bearing capacity between grouted and
non-grouted piles have been paradoxical.
2.3.3 Precast Piles
Precast piles can be categorized into timber, steel and concrete piles based on the
materials, or into H, circular, square and polygon piles based on the configura-
tions of pile sections. Generally, they can also be categorized into small, large
and non-displacement piles based on the amount of soil squeezed during a driven
process. Normally, solid concrete piles and end-closed pipe piles are considered large
displacement piles, and piles such as H-steel piles with a small cross-sectional area
are considered small displacement piles.
For precast concrete rectangular piles (large displacement piles), the sizes can
range from 200 to 700 mm in diameter and 10 to 25 m in length, and the working
loads that can be resisted vary from 200 to 1,200 kN. These piles can be a normally
reinforced structure or a prestressed structure. Past studies focused on the pile tests
illustration (Mansur and Hunter 1970), pile capacity under different types of soil
profiles (Gregersen et al. 1975; Balasubramaniam et al. 2009), drivability (Ashford
and Jakrapiyanun 2001; Fellenius and Samson 1976; Hussein et al. 2006), load
transfer mechanism (Fellenius 2002; Hsu, 2014), configurations of piles and strength
of concrete (Liew et al. 2004). Recently, there have been intensive investigations
aiming to research materials’ effect on concrete piles. For example, because of a lack
of data using Illinois pulverised coal combustion bottom ash in concrete structures,
full-size tests were performed on concrete piles with this utilization of bottom ash,
and then the results were compared with the traditional reinforced concrete pile with
fly ash admixture (Kumar et al. 2004).
For pipe piles, open-ended piles possess an anticipated greater efficiency
(capacity to weight ratio) than do round solid piles. Other advantages of these small
displacement piles include that they are easy to drive, and the soil compaction
effect is small. However, the behavior of open-ended piles is more complicated
because the soil plug effect created inside the pile should be considered when
investigating these piles. Recently, a new CPT-based HKU method was proposed for
base capacity estimation of open-ended pipe piles with mechanical consideration
of annulus resistance and plug resistance (Yu and Yang 2012). Detailed reviews of
these precast piles were provided with consideration of soil profiles, methods of
driven technique and configurations of precast piles (Marcos et al. 2013). Further,
another aspect associated with precast piles in clay is the setup or ‘freeze’ of the
foundation, which is a normal phenomenon whereby pile capacity increases with

the increment of time. Recently, research has commenced to determine the amount
of setup time of precast piles by performing dynamic load tests, statnamic load tests
and SLTs (Abu-Farsakh et al. 2017).
A screw pile is also a small displacement precast pile which is mostly made of
steel. This precast pile consists of a helical plate and circular hollow section. AS 2159
(Standards Australia 2009) classifies the steel screw pile as a ‘preformed displace-
ment’ pile; however, this screw pile does not compact the ground as does a driven
pile. The methods based on CPT value to predict the capacity of pile are inaccurate.
Yttrup and Abramsson (2003) proposed a method for determining the base-bearing
capacity of steel screw piles in sand. The interpretation used to determine the ultimate
bearing capacity of this screw pile is based on the modified Davisson’s offset method
and is governed by a settlement equal to 10% of the helix diameter (Perko 2009).
Recent research has found that Chin’s method and Decourt’s method overestimate
the ultimate bearing capacity of this screw pile; therefore, the reduction factor has
proposed (Malik et al. 2016).
Non-displacement piles are mostly bored and cast-in-place. As illustrated in
Sect. 2.3.2, these piles are mostly manufactured by removing the soil through a
drilling process by machine such as auger, and then by placing reinforcements inside
the drilled hole. Finally, this hole is filled with concrete. As a result of this construc-
tion process, where no soil is compacted, these piles are considered non-displacement
piles. However, there is very limited research on the process of removing the soil first
and then descending the precast pile into drilled hole, which can also be considered
a non-displacement pile.
2.3.4 Composite Piles
2.3.4.1 Traditional Composite Piles
Traditional piles refer to piles made of traditional construction materials. These are
steel, concrete and timber piles. Piles made of more than one material are recognized
as traditional composite piles. For example, piles made of steel pipes with a concrete
core are considered composite piles, as shown in Fig. 2.31. In 1898, Considere
found that confined cylinders exhibit an extreme increase in longitudinal compressive
strength, and the first use of concrete-filled steel tubes occurred in 1901 by Sewell
(Peterson 1999). The inner diameter of concrete-filled tubes or piles can range from
0.2 to 2.8 m, with wall thickness varying from 8 to 20 mm. These piles are mainly
used in marine engineering and bridge engineering because of their high strength and
good ductility. In particular, for bridge construction, the use of concrete-filled pipe
piles is more economic because they account for 15 to 20% of the total construction
cost of conventional steel bridges (Nakamura 1998).
When these piles are used in the marine condition, seawater corrosion leads to
decreased tube wall thickness and delamination between the tube and concrete core.
Fig. 2.31 Concrete-filled pipe pile (Khaleghi et al. 2016)
In this context, numerous studies have been performed to investigate ways to inspect
these composite piles. Compared with remote-operated vehicles (ROVs), acoustic
methods—including sonic, ultrasonic and acoustic emission methods—are less time
consuming. However, ROVs cannot be used to examine internal defects. Thus, many
wave modes have been researched, and many studies have been conducted to investi-
gate and detect corrosion and cracks in tubes using cylindrical Lamb waves (Alleyne
and Cawley 1995; Cheng and Cheng 1999; Kundu and Ryu 2002; Rose et al. 1994;
Yang et al. 2014).
Other research aspects of these concrete-filled pipe piles vary. For example, field
tests on piles were performed in consideration of the fact that earthquakes lead
to the soil liquefaction. Three tested piles were applied with lateral vertical and
moment loadings. The tested piles included two concrete shaft piles and one concrete-
filled pipe pile. The results indicated that stronger piles may tolerate larger lateral
displacement than weaker piles (Naesgaard 1992). Moreover, some investigations
have focused on pile behavior under special soil conditions. For example, a study
of the concrete-filled pipe pile was conducted focusing on the effect of frozen soils
on lateral behavior. It found that frozen soils drastically changed the ground motion
characteristics and lateral behavior (Xu 2009). Other research has focused on the
behavior of these piles under different loadings (Chen et al. 2011; Lao et al. 2004).
Further research on the inclined steel pipe pile was conducted recently and found
that p-y curves increased linearly when the lateral displacement at the pile head was
smaller than 20 mm (Huo et al. 2015).
Piles with relatively high strength can be recognized as rigid piles. These piles
are mostly made of cement, aggregates and reinforcement. In contrast, Deep Cement
Mixing (DCM) piles, which is a product created by DCM technology that is used
for soft ground treatment, can be determined as soft piles. The strength of these piles
is comparatively low because there are no aggregates used. As shown in Fig. 2.32,
the machine drills the hole and simultaneously sprays cement solidifying agent and
Fig. 2.32 Deep soil mixing (Madhyannapu and Puppala 2015)
water until the required depth, and then agitates the ground by reverse rotation. This
process is finished when the ground is sufficiently agitated.
These soft piles can improve the stability of road embankments (Taesiri and Chan-
taranimi 2001) and can decrease vertical settlements (Wonglert and Jongpradist
2015). Intensive research focused on the slope stabilization of DCM piles. These
previous works considered strength and deformation parameters, creep behavior
and safety factors for designing DCM piles (Bergado et al. 1999; Buathong and
Mairaing 2010; Laiet al. 2006; Miura et al.2001; Mokhlesur et al. 2011). A recent
study referring to the spacing, depth, elastic modulus and volume of a row of DCM
piles was conducted with results comparison to FEM modelling. The researchers
presented recommendations of parameter determination for design to minimize the
lateral displacement of soil (Boathong et al. 2014). For settlement prediction, a recent
study proposed an empirical settlement formula through non-linear regression of data
obtained from a field test (Yang et al. 2012); however, a limitation was that the test
loading was relatively small and the settlement predictions from the unloading stages
were not presented.
2.3.4.2 Fiber-Reinforced Polymer (FRP) Composite Piles
There are many uses for piles, and issues arise when piles are located in harsh envi-
ronments, especially in marine or coastal conditions. Piles with traditional materials
can be destroyed because of corrosion of steel, deterioration of timber and degra-
dation of concrete, as shown in Fig. 2.33. The deterioration of timber, concrete and
steel piling systems costs the United States nearly US$2 billion per year for repair
Fig. 2.33 Corrosion of steel, deterioration of timber and degradation of concrete (Pando 2003)
and replacement (Hassan and Iskander 1998). In 1998, the Federal Highway Admin-
istration (FHWA) initiated a project on the use of FRP pilings as the replacement of
traditional piles in the context of a waterfront rehabilitation project, whose one goal
was to replace up to 100,000 bearing piles for lightweight structures.
The development of concrete piles has a long history. Engineers are now facing
a problem relating to piles because, even for piles made of concrete and steel,
which possess good rigidity and high strength, damage still occurs, as shown in
Fig. 2.34. These figures depict damage to concrete piles photographed from Adelaide
in Australia.
FRP piles have been employed for approximately three decades. Dating back
to April 1987, the first prototype recycled pile was driven at the Port of Los
Angeles (Juran and Komornik 2006). This replacement of creosote-treated timber
piles successfully avoided the threat of marine borers. As early as 1998, the Empire
State Development Corporation undertook a waterfront rehabilitation project, known
as Hudson River Park, which involved replacing up to 100,000 bearing piles for
lightweight structures (Pando 2003). Moreover, concrete-filled FRP composite piles
were employed by the Virginia Department of Transportation in 2000 for an entire
3 cm
Fig. 2.34 Degraded concrete piles in Adelaide, South Australia, Australia

bend of the new Route 40 bridge over the Nottoway River in Sussex County, Virginia
(Pando 2003; Pando et al. 2006).
Although this FRP material is expensive, the overall cost in the long term is
more economical than for the traditional materials of concrete and steel because
of the advantages of low management costs and a long service life. Ballinger also
stated that, although the cost of FRP composite materials may be higher, the cost
of labor and use of equipment for construction work may be lower because of their
lighter weight (Pando et al. 2002). Moreover, aside from the cost, there is growing
concern about the environmental and health effects of using treated timber and steel
painted with solvent or heavy-metal coatings (Guades et al. 2012). These poisonous
materials threaten marine borers, which led to engineers starting to replace timber
piles with FRP piles made of fiber and plastics or resins. Robinson and Iskander
(2008) stated that, by using recycled plastics to manufacture FRP piles, materials are
being used that may have otherwise been landfill. Therefore, this approach can be
more economical in aggressive environments when lifecycle costs are considered.
According to data from the EPA (2006), less than 10% of the 13.7 million tons (equal
to loading of 122 GN) of plastic containers and packaging produced annually in the
United States is recovered by recycling (Robinson and Iskander 2008).
Many researchers have started to examine FRP piles. Most research has devoted
attention to the drivability, driving efficiency, durability and surface friction between
FRP and soils, the effect of the hammer, and the resistance of the soil. Research on
the driving hammer effect was conducted using wave equation analysis that consid-
ered the effects of the weight and velocity of the hammer and pile property. This
research indicated that a single-acting steam hammer was more efficient than a diesel
hammer because it could drive the composite deeper with the same number of blows
(Hassan and Iskander 1998). A modelling study simulated via the software program
Microwave indicated that both hollow and concrete-filled FRP piles could be driven
by a heavier hammer (Ashford and Jakrapiyanun 2001). Moreover, soil resistance to
driving—including side friction and end bearing resistance—was also investigated
using the wave equation analysis program. The results from the entire spectrum
of the study demonstrated that, for drivability, there was more substantial differ-
ence in friction and end bearing conditions for concrete-filled FRP and concrete
piles (Mirmiran and Shahawy 1996). The outcome of a study that researched the
interface behavior between sand and FRP concluded that FRP exhibited similar rela-
tionships between peak interface friction coefficients and relative roughness for a
given granular material (Fam 2000; Pando et al. 2002).
A review on the driving performance of FRP composite piles indicated that the
types of composite piles are steel pipe core piles, structurally reinforced plastic (SRP)
piles, concrete-filled FRP piles, fiberglass pultruded piles, fiberglass-reinforced
plastic piles, hollow FRP piles and FRP sheet piles, as shown in Fig. 2.35 (Guades
et al. 2012). A further study by Pando (2003) emphasized that the composite prod-
ucts available in the market are steel pipe core piles, SRP matrix piles, concrete-filled
FRP pipe piles, fiberglass pultruded piles, plastic lumber piles, hollow FRP piles and
FRP sheet piles.
Fig. 2.35 Types of FRP composite piles (Guades et al. 2012)
Based on the results obtained from previous investigations, Robinson and Iskander
(2008) conducted notable research to examine the behaviors of FRP composite piles
under vertical loads to determine their stiffness, flexibility, settlement and bearing
capacities. Through this research, in-situ static and dynamic tests were conducted
with onsite SPTs and CPTs. Four types of FRP pile were tested in Elizabeth, New
Jersey, involving concrete-filled fiberglass shell piles (Lancaster pile, Fig. 2.36a),
polyethylene piles reinforced with steel bars (PPI pile, Fig. 2.36b), polyethylene piles
reinforced by fiberglass bars (SEAPILES, Fig. 2.36c) and solid polyethylene piles
(American Ecoboard pile, Fig. 2.36d). These tests indicated the possible applicability
of plastic piles to traditional axial loading applications, and highlighted the need
for further work on the long-term creep performance and durability of these piles
(Robinson and Iskander 2008).
Concrete-filled FRP piles came into notice after these tests were completed and
were considered the best FRP piles to resist upper loading. Though the other types of
FRP piles showed inadequate capacity, using these piles to replace the traditional pile
is acceptable. The reason is that the steel pipe pile is mostly used under conditions
of being exposed to water, which means the FRP is actually only used for protection
from corrosion. In another example, SRP, is mainly used in fendering applications and
regarded as a potential load-bearing pile (Guades et al. 2012). The FHWA proposed
that FRP composite piles can be used effectively as vertical load-bearing piles and
represent an alternative for deep foundation construction, especially in waterfront
environments and aggressive soils (Juran and Komornik 2006).
A large amount of research has focused on investigating reinforced concrete struc-
tures with externally bonded FRP applications for the purpose of strengthening
and repairing. However, one of the limitations of this technology is the propen-
sity of the FRP to prematurely de-bond at strains well below its rupture strain
(a) Lancaster pile (b) PPI pile
(c) SEAPILE pile (d) American Ecoboard pile
Fig. 2.36 Tested composite piles (Juran and Komornik 2006)
(Zhang et al. 2012a). Some researchers conducted tests to design FRP anchors, and
provided the criterion for optimal assessment. Luo (2014) reported on other appli-
cation projects used for slope reinforcement. For example, the technology of GFRP
reinforcement anchor was adopted in the project of the Changji Expressway, built in
2008, for the purpose of red sandstone slope reinforcement. This proposed reinforce-
ment system used GFRP bolts with 28 mm diameter, and the results demonstrated
that the slope was stabilized overall. Luo (2014) also discussed another GFRP anchor
application used in an underground retaining project, in which GFRP replaced the
Chinese traditional HPB325, and the results were successful after the pull-out tests.
Other potential options for anchorage include applying FRP bolts or FRP anchor
bolts, which are used for slope treatment in expansive soils. The reason for using FRP
bolts is that the traditional main governance method for an expansive soil slope is to
use a steel anchor bolt and frame beam or grid; however, these are easily corroded
and have poor durability (Liu 2014). Through comparison and analysis of the FRP
anchor bolt and steel bolt, the author pointed out that, although the maximum crack
widths were different, the pulling resistances were almost equal.
A Tunnel Boring Machine (TBM) is one of the most efficient machines used
for tunneling; however, this machine cannot easily break the walls or piles that are
used to retain the earth pressure (structures with steel reinforcement). The traditional
way to solve this problem is with manual work; however, this is dangerous to the
workers and is time consuming. For example, the project of Yuantong subway station
in Nanjing city, China, suffered human losses when breaking retaining structures in
2007, as the soil layers collapsed. In this accident, the ground cracked and extended
to 150 m (Liu et al. 2014).
Application of GFRP bars is the most commonly used form of reinforcement
during shield tunneling construction. The replacement of traditional steel reinforce-
ment with GFRP reinforcement in concrete structures (pile or retaining wall) success-
fully solves the problems mentioned above. This is because GFRP displays brittle
behavior when loading reaches and surpasses failure conditions, and the TBM can
directly cut the retaining structures. This technique is widely accepted in China,
and numerous projects have applied GFRP to fully or partially replace traditional
reinforcement. Some projects with GFRP geotechnical application (retaining walls)
are the Shanmei subway station R2 in Dongguan city (Ming 2011), the Shenzhen
Subway R5 in Shenzhen city (Zhang et al. 2011) and the Wuhan Changjiang tunnel
in Wuhan city (Jiao 2007). Note that most of the project designs are based on ACI
440.1R-06s (2006) and GB 50010–2002 (2002). A recent review indicated that there
is limited research on the deflection behavior of FRP bar–reinforced concrete piles
considering supports installation and soil excavation; and no investigation referring
to the FRP confinement on concrete piles (Zhou et al. 2016).
2.3.5 Other Piles
As a result of their small range of use, piles with unusual configurations and piles that
are a combination of a traditional pile and other elements can be considered ‘other
piles’. Such piles include ribbed piles, T-shaped piles, H-steel-reinforced concrete
pipe piles and stiffened DCM piles.
To achieve the requirement of piles with high capacity and low settlement,
geotechnical engineers have designed ribbed piles, which can increase shaft resis-
tance. Field trials indicated that these ribbed piles could increase capacity by 30 to
40% (Ground Engineering 2003). Recent research explored the influence of ribs on
the ultimate bearing capacity of bored piles. As shown in Fig. 2.37, three different
types of ribbed piles were tested and compared with a plain pile. It was found that
the ribbed piles were more environmentally friendly. Further, a plastic failure enve-
lope for the pile toe was proposed, and an ultimate design solution was presented
(McNamara and Gorasia 2016).
In some countries where pretensioned spun High-strength Concrete (PHC) piles
are commonly used, there have been new types of piles created to overcome the
shortcomings of these PHC piles. These new piles are called ‘extended end’ piles
(or ‘Ext piles’), as shown in Fig. 2.38. Some research has focused on these piles’
Fig. 2.37 Ribbed bored piles (McNamara and Gorasia 2016)
Fig. 2.38 Extended end concrete pile (Kim et al. 2017)
material properties and bearing capacity, based on laboratory trail tests (Lee and
Song 2010; Shin et al. 2014). Recently, a further study investigated Ext piles with
respect to time, cost and workability. Based on the field test, it was found that the
bearing capacity of Ext piles is better than PHC piles by about 35–50%, and the use
Fig. 2.39 Structure of the

two-layered composite pile
(Li et al. 2015)
of Ext piles will decrease the work duration and project cost by about 25% and 14%,
respectively (Kim et al. 2017).
Composite piles that use monolithic piezoelectric materials have some short-
comings because this material exhibits brittle behavior and is easy to crack. In this
context, some researches have commenced referring to the 1–3 composite mate-
rials that possess good flexibility. Previous work focused on the properties of the
composite material and the manufacture procedures (Manh et al. 2013; Park et al.
2003; Zhou et al. 2012a). A recent study investigated the 1–3 composite piles, as
shown in Fig. 2.39. In this study, PZT-5 was used as the piezoelectric material, while
polyurethane with low stiffness was selected as the passive polymer matrix (Li et al.
2015). The limitation of this research was that there were no field tests conducted
because the application of these piles is very rare.
For soft piles, the failure mode depends on the relative strength between piles
and soils. The strength of DCM piles (soft piles) is relatively low; thus, the pile
foundation system mostly fails because of failure of the pile head. To solve this
problem, previous studies have been undertaken referring to the cement content,
which can increase the strength of the pile by adding extra cement. However, these
studies found that, with cement content increasing, the cement-treated clay did not
linearly increase, and the efficiency was low (Jongpradist et al. 2010; Uddinet al.
1997). Further, in the manufacturing process of DCM piles of mixing cement with
soil underneath, it is uneconomic to add extra cement content along the length of the
DCM pile because only the pile head requires strength increase.
One method to increase the capacity of DCM piles is inserting a stiffer core into the
piles, as shown in Fig. 2.40. The field tests indicated that DCM piles with an inserted
core are an effective and economic approach for settlement control and capacity
improvement (Dong et al. 2004). The configuration of the core can vary, though the
Fig. 2.40 Failure patterns of stiffened DCM piles (Wonglert and Jongpradist 2015)
common cores are circular or rectangular concrete columns, and sometimes H-steels
(Jamsawang et al. 2011; Werasak and Meng 2013). Recent research focused on the
shape of the core, such as the ratio between the length of the core and pile, as well
as simulation (Voottipruex et al. 2011; Wonglert and Jongpradist 2015). Another
method to increase the capacity of DCM piles is changing their configurations. As
shown in Fig. 2.41, field tests have been conducted on the T-shaped deep mixed
(TDM) soil piles. These tests found that, compared with the normal DCM piles, the
T-shaped piles could save cement and time (Liu et al. 2011).
Fig. 2.41 TDM column–supported embankment (Liu et al. 2011)

2.4 Analytical Method for Bearing Capacity 45
2.4 Analytical Method for Bearing Capacity
2.4.1 Meyerhof Method
It seems that Meyerhof (1956) was the first person to develop a method of pile
capacity estimation based on the N-value obtained from standard penetration resis-
tance (Alansari 1999). The empirical correlations proposed by the Meyerhof were
based on analysis of numerous pile load tests in a variety of cohesionless soil deposits.
The report published by the FHWA highlighted that this method is quick and easy
to use; however, it should only be used for preliminary estimates and not for final
design because the SPT test data can be influenced by various factors (Hannigan
et al. 2006).
Meyerhof (1976) reported empirical equations for the average unit shaft resistance
(f s ) of displacement and non-displacement driven piles using the average corrected
−
SPT resistance value N (blows per 300 mm) as shown in Eqs. 2.23 and 2.24, respec-
tively. The recommended equation of base resistance (qt ) for driven piles in sands and
gravels is provided in Eq. 2.25. Note that for piles driven in a uniform cohesionless
stratum, the base resistance can be determined in Eq. 2.26. By correlating the SPT
N-value into the average SPT N-value, the shaft and base resistance can be obtained.
By multiplying the shaft and base areas respectively, the shaft and base capacity can
be calculated, and then, the total ultimate bearing capacity can be evaluated.
f s = 2N ≤ 100k Pa (2.23)
f s = N ≤ 100k Pa (2.24)
(40N B − 40N O )D B
qt = 400N O + ≤ 400N B (2.25)
b
40N B D B
qt = ≤ 400N B (2.26)
b
where:
N O = average corrected SPT N-value for the stratum overlying the bearing
stratum.
N B = average corrected SPT N-value of the bearing stratum.
D B = pile embedment depth into the bearing stratum in meters.
b = pile diameter in meters.
2.4.2 Brown Method
Instead of using the average corrected N-value, the Brown (2001) method uses the
value of N 60 to calculate the shaft and base resistances. The Brown method is a
simple empirical method that is based on capacity correlations from 71 SLTs. These
tests are preformed from Caltrans projects, and there is a wide variety of soil types
(Hannigan et al. 2006).
The recommended shaft resistance for impact piles is provided in Eq. 2.27, and
base resistance in Eq. 2.28. Note that the F VS is a reduction factor for vibratory
installed piles, and parameters of Ab and Bb are obtained from the regression analyses.
The detailed values are provided in Table 2.4. After the shaft and tip resistances are
determined, they are multiplied with the shaft and toe area, and then the total capacity
of a single pile can be determined. It should be noted that the shaft area of a H-pile
is the ‘box’ area, and, for the shaft area of an open-ended pile, the external surface
area is recommended.
f s = FV S (Ab + Bb N60 ) (2.27)
qt = 0.17N60 (2.28)
Table 2.4 Input factors for Brown’s method (Hannigan et al. 2006)
Loading Installation Soil type F vs Ab Bb
condition method kPa (ksf) kPa (ksf)
Compression Impact Clay to sand 1.00 26.60 0.555 1.92 0.04
Gravelly sand 1.00 42.60 0.888 42.60 0.888
to boulders
Rock 1.00 138.00 2.89 138.00 2.89
Tension Impact Clay to sand 1.00 25.00 0.522 1.80 0.0376
Gravelly sand 1.00 40.00 0.835 0.00 0.000
to boulders
Rock 1.00 130.00 2.71 0.00 0.000
Vibratory Clay to sand 0.68 25.00 0.522 1.80 0.0376
Gravelly sand 0.68 40.00 0.835 0.00 0.00
to boulders
Rock 0.68 130.00 2.71 0.00 0.00
2.4.3 Nordlund Method
Based on field observations involving considerations of pile taper and displacement

of soil during several load test programs in cohesionless soils, Nordlund developed
this method in 1963 and updated it in 1979 (Eq. 2.29), as shown in Fig. 2.42. Nordlund
(1979) suggested that the shaft resistance is a function of the variables of the friction
angle of the soil, friction angle on the sliding surface, taper of the pile, effective unit
weight of the soil, pile length, minimum pile perimeter and volume of soil displaced.
The Nordlund method tends to over-predict pile capacity for piles with widths larger
than 600 mm. In addition, the effective overburden pressure, which is used to compute
the pile base resistance, is limited to 150 kPa. For a pile with embedded length d, if it
has a uniform cross-section, the angle of the pile taper from vertical (ω) is zero, and
the Nordlund equation becomes Eq. 2.30. Diagrams and tables of the relationships
between δ/ϕ and pile soil displacement (V ), evaluation of K δ with a friction angle

ranging from 25 to 40°, αt and Nq can be found in FHWA NHI-16–009.
d=D
sin(δ + ω)
Qu = K δ C F Pd Cd d + αt Nq At pt (2.29)
d=0
cos ω

Q u = (K δ C F Pd sinδCd d) + αt Nq At pt (2.30)
Fig. 2.42 Nordlund’s general equation for ultimate pile capacity (Nordlund 1979)
where:
D = depth of embedment.
K δ = lateral earth pressure coefficient at depth of d.
C F = correction factor for K δ .
Pd = effective overburden pressure.
δ = friction angle between soil and pile.
Cd = pile diameter at depth of d.
d = pile segment length.
αt = depth–width relationship factor (dimensionless).

Nq = capacity factor.
At = pile tip area.
pt = effective overburden pressure at pile tip.
2.4.4 Tomlinson Method
The static analysis for piles in cohesive soils includes the Tomlinson method (α
method), effective stress method (β method) and a method based on CPT. The
Tomlinson method is used to calculate the short-term load capacity of piles in cohe-
sive soils, and the effective stress method can be used to calculate the short- and
long-term load capacity of piles in both cohesive and cohesionless soils (Helwany
and Wiley 2007).
The Tomlinson method—also known as α method—is a total stress method that
uses the parameters acquired from undrained soil shear tests to calculate the capacity
of the pile in cohesive soil. This approach assumes that the shaft resistance is inde-
pendent of the effective overburden pressure and that the unit shaft resistance can be
expressed in terms of an empirical adhesion factor times the undrained shear strength
(Samtani and Nowatzki 2006b). Two factors are used to calculate the shaft resistance
and toe resistance via this method. Equations for the shaft and toe resistance are
provided in Eqs. 2.31 and 2.32. The capacity factor, N c , used to calculate the toe
resistance in clay is usually taken as 9 for deep foundations:
f s = ca = αcu1 (2.31)
qt = Nc cu2 (2.32)
where:
α = empirical (adhesion) factor.
cu1 = average undrained shear strength.
cu2 = undrained shear strength below the toe of a pile.
Nc = dimensionless bearing capacity factor.
Fig. 2.43 Adhesion factor determination scenario A (Samtani and Nowatzki 2006b)
Fig. 2.44 Adhesion factor determination scenario B (Samtani and Nowatzki 2006b)
The shaft resistance is equal to the adhesion between the pile and soil in failure
conditions, and the adhesion can be determined by using the undrained shear strength,
cu of clay along the pile multiplied with the empirical factor. This coefficient depends
on the clay strength, load magnitude, pile dimension, installation method and time
effect. Further, it is a function of the pile embedment and soil stratigraphy.
The adhesion factor, α, can be obtained from Figs. 2.43, 2.44 and 2.45 for piles
driven through a sand or gravel layer and into an underlying stiff clay stratum;
through soft clay and overlying stiff clay; and through stiff clays without any different
overlying strata, respectively.
2.4.5 Effective Method
Besides the Tomlinson method, the effective stress method is an alternative method
used to calculate static pile capacity in cohesive soil. Compared with the α method,
the β method uses drained soil strength parameters for capacity determination, instead
of parameters obtained from undrained soil tests. As such, the effective friction angle

of soil, ϕ , should be used. Thus, the β method can be used to calculate the capacity
Fig. 2.45 Adhesion factor determination scenario C (Samtani and Nowatzki 2006b)
in both cohesive and cohesionless soil. The formulas of the unit shaft resistance and
toe resistance are provided in Eqs. 2.33 and 2.34:
f s = β po (2.33)
q t = N t pt (2.34)
where:
β = Bjerrum-Burland beta coefficient = K s tan δ.
po = average effective overburden pressure along the pile shaft.
K s = earth pressure coefficient.
δ = interface friction angle between pile and soil.
Nt = toe bearing capacity coefficient.
pt = effective overburden pressure at the pile toe.
To determine the designed capacity of a pile using β method, the effective over-
burden pressure should be calculated in each layer. Moreover, the effective friction

angle, ϕ , should be determined from laboratory or in-situ tests in the soil profile.

The ϕ can also be determined via the SPT N-value or corrected N-value if laboratory
tests are unavailable.
Figure 2.46 is used to determine the parameter of β to compute the unit shaft
resistance of each soil layer. Table 2.5 and Fig. 2.47 can be used to determine the toe
bearing capacity factor Nt . The total shaft resistance can be determined by summing
up the values obtained from multiplying the shaft stress with the contact area (between
each layer and pile surface) based on in Eq. 2.35. The ultimate bearing capacity of
a pile based on the effective method is equal to the sum of the total shaft and end
resistance in kN, and the allowable designed load is governed by Eq. 2.36. Based on
the construction methods, the safety factor can be obtained, as illustrated in Table
2.6.
Fig. 2.46 Chart for estimating β coefficient as a function of soil type ϕ’ (Fellenius 1991)
Table 2.5 Approximate

Soil type ϕ β Nt
range of β and Nt coefficients
(Fellenius 1991) Clay 25–30 0.23–0.40 3–30
Silt 28–34 0.27–0.50 20–40
Sand 32–40 0.30–0.60 30–150
Gravel 35–45 0.35–0.80 60–300
Fig. 2.47 Chart for N t

coefficients as a function of
soil type ϕ’ angle (Fellenius
1991)
Rs = f s As (2.35)
Qu
Qa = (2.36)
Factor o f Sa f et y
Table 2.6 Recommended

Construction control method Factor of safety
safety factor (Hannigan et al.
2006) SLT with wave equation analysis 2.00
Dynamic testing with wave equation analysis 2.25
Indicator piles with wave equation analysis 2.50
Wave equation analysis 2.75
Gates dynamic formula 3.50
where:
Rs = total shaft resistance.
f s = shaft resistance between soil layers with concrete.
As = shaft area.
Q u = ultimate bearing capacity of pile foundation.
Q a = allowable pile capacity.
2.5 Field Tests
2.5.1 Static Load Tests
2.5.1.1 Compressive Static Load Tests
The basic scheduled mechanism is provided in Fig. 2.48 for a vertically loaded
pile test. Increment loads are applied from the pile head, and the settlements are
monitored (Kyfor et al. 1992). A load–movement curve of the pile head can be
plotted. In addition, the settlement and transferred loads anywhere along the pile can
be determined if instrumenting telltales or stain gauges inside the pile.
ASTM-D-1143/D-07 (1994) provides recommendations for several alternative
systems to perform SLTs under compressive loading. Commonly, the compressive
loads are provided by hydraulic jacks against a reaction beam (as demonstrated in
Fig. 2.49) that can be anchored by reaction piles or ground anchors or against a
weighted platform. The settlement of a pile can be recorded through installation of a
monitoring system, such as LVDTs or dial gauges. It should be noted that determining
the ultimate capacity of a pile based on gross movements is not recommended because
there is no consideration of elastic deformation of the pile shaft, which leads to
underestimation of long piles and overestimation of short piles. The presentations
and interpretation of typical SLTs are provided at the end of the Sect. 2.5.
There are seven types of methods to perform SLTs: quick tests, maintained tests,
loading in excess of maintained tests, constant time interval tests, constant rate of
penetration tests, constant movement increment test and cyclic loading tests. In this
book, all compressive SLTs performed were based on maintained tests. This test
2.5 Field Tests 53
Fig. 2.48 Basic mechanism of a compression pile load test (Kyfor et al. 1992)
maintains loads up to twice the value of the original designed loads (ASTM Interna-
tional 1994; Standards Australia 2009). Some national codes or local standards may
require maximum loading that varies from 1.5 to four times the design load. The test
loads in increments of 10 to 25% of the design load, and each of the loads should be
maintained until the recorded rate of axial movement does not exceed 0.1 to 0.25 mm
per hour. During each applied load, the settlement needs to be recorded with incre-
menting intervals of five, 15, 20 and 45 min. One hour may also be required if the
rate of axial movement does not meet the standard requirement. Similarly, the incre-
ment time intervals may also vary with different codes. It is recommended that local
standards should be used with consideration of the geotechnical situation, instead of
selecting one code.
2.5.1.2 Uplift Static Load Tests
The pile foundation can not only be used to resist the downwards loads from the upper
structure, but can also be used as a structure to resist uplift loads—such as for the
piles under a wind turbine. Compared with compressive SLTs, the basic mechanics
of this test are similar, except for the loading direction, which moves upwards. The
Fig. 2.49 Schematic of anchored reaction frame (Hannigan et al. 2016)
typical setup of uplift SLTs is provided in Fig. 2.50, which also shows the hydraulic
jacks used for load provision that are located at the top of the test beam. If reaction
piles are cast and connects with reaction beam which is also connected with pile
reinforcement, the hydraulic jacks can be placed on the leveled ground and pushing
reaction beam upward. The reason to this set-up is that more than two jacks can be
used, which will increase the testing capacity. Detailed information can be found in
Chap. 3. For the data acquisition, LVDTs and dial gauges are the commonly used
equipment to monitor displacement of the pile head.
The test procedures consist of quick tests, maintained tests, loading in excess of
maintained tests, constant time interval tests, constant rate of uplift tests and cyclic
loading tests (ASTM International 1995a). For determining driven piles, the quick
loading procedure is recommended (Hannigan et al. 2006). Increments of 10 to 15%
of the design load are required for this test, with a constant time interval of 2.5 min to
perform the uplift SLTs. Further, before the additional load increment is applied, the
time, load value and total upwards displacement should be recorded immediately.
After the final load is applied, unloading is required in five load decrements, and
all five loads should be held for five minutes. Compared with other methods, the
maintained test is comparatively time consuming, yet allows the soil to be stable.
2.5 Field Tests 55
Fig. 2.50 Typical setup for tensile load test (Courtesy of WKG2 )
This procedure requires maximum loads of up to 200% of the anticipated design

load, and each load increment can only be applied when the rate of displacement of
the pile head does not exceed 0.25 mm per hour.
2.5.1.3 Lateral Static Load Tests
In recent years, pile foundations have been used to resist lateral load, particularly
with regard to special design event such as seismic and vessel effects. For example,
structures that suffer from large wind loads, such as large overhead signs, will result
in lateral loads on the foundation. The primary purpose of lateral SLTs is to obtain
the p-y curves and capacity of a single pile. The instrumentation and observation
equipment of lateral SLTs are similar to compressive and uplift SLTs. The reaction
system used can involve reaction piles, weighted platforms or a deadman. In some
projects, reaction systems may be different. For the convenience of performing lateral
SLTs, sometimes, two piles are tested simultaneously because each can be considered
the reaction pile to the other as shown in Fig. 2.51.
Lateral SLTs contain standard loading, excess loading, cyclic loading, surge
loading, reverse loading, reciprocal loading, specified lateral movement and
combined loading tests (ASTM International 1995b). The most common lateral
SLTs are standard load testing. These tests apply a maximum load of 200% of the
design load, similar to the compressive and uplift maintained procedure; however,
the time intervals are different. At times, for cast-in-situ piles, a PVC tube may be
installed inside the pile foundation, and the deflection data along the pile shaft can
be determined by an inclinometer during this test for lateral behavior investigation.
Fig. 2.51 Setup for lateral static load test (Courtesy of WKG2 )
2.5.1.4 Extrapolations of Field Tests
The most reliable method to determine the actual compressive, uplift and lateral
capacities of piles involves conducting SLTs until failure of pile or yield of soil.
These methods refer to physical loading on piles at required time intervals, and then
monitoring the vertical or horizontal displacement of piles. Under consideration of
cost and time, loading until the pile experiences excessive displacement (failure of
soil-pile system) is conducted infrequently. In practice, the proof tests are the most
common tests instead of failure tests, which means that the maximum applied load
does not represent the ultimate bearing capacity of pile. The interpretation of these
field tests’ results is required to determine the ultimate bearing capacity.
For compressive SLTs, the typical presentation is plotting vertical settlement
versus loads, as illustrated in Fig. 2.52. To facilitate the interpretation, an imple-
mented scale considering elastic deformation of single pile was proposed, and the
elastic compression line was kept inclined at an angle of 20° (Vesic 1977). The
elastic deformation is governed by Eq. 2.37. In some Asian countries, the standards
also require diagrams of settlement-lg (time) curves (s-lgt) and settlement-lg (load)
curves (s-lgQ). As shown in Fig. 2.53, s-lgt curves can provide information of the
pile under failure load; however, for the proof test, the failure load curve cannot be
easily determined. Note that the s-lgQ curves are similar to DeBeer’s (1970) method.
QL
= (2.37)
EA
where:
2.5 Field Tests 57
Fig. 2.52 Typical results of compressive SLT (Hannigan et al. 2006)
Fig. 2.53 s-lgt result of lg Time (min)

compressive SLT 1 10 100 1000
0
10
20
30
Settlement (mm)
40
50 1680kN
2520kN
60
3360kN
70 4200kN
5040kN
80 5880kN
6720kN
90 7560kN
8400kN
100
= elastic deformation in mm.

Q = test loads in kN.
L = pile length in mm.
E = elastic modulus of pile in kPa.
A = pile cross-section area in m2 .
Fig. 2.54 Interpretation based on Davisson’s method (US unit) (Samtani and Nowatzki 2006b)
Davisson (1972) proposed a method by offsetting the elastic compression line to

determine the failure load and the corresponding settlement. As shown in Fig. 2.54,
the offset settlement value, sf , is equal to 0.15 inches plus the pile diameters in
inches over 120. Equation 2.38 provides the sf in SI units. Based on Davisson’s
offset method, the intersection of offset line with load–settlement curve (obtained
from SLTs) provides the ultimate capacity of a single pile. At times, this offset
line does not intersect with the load–settlement curve from loading phases, which
illustrates that the ultimate bearing capacity of this pile exceeds the maximum loads
applied from the SLTs. For large diameter piles, which are usually over 610 mm,
additional pile toe movement must be considered, and the offset value needs to be
replaced, as given in Eq. 2.39:
sf = + (4.0 + 0.008b) (2.38)
sf = + (b/30) (2.39)
where:
s f = offset displacement of elastic compression line at failure condition.
= elastic deformation of total pile length.
b = pile diameter or width.
Compared with Davisson’s criteria, the double tangent method is more commonly
used to determine cast-in-situ piles. As shown in Fig. 2.55, two tangent lines were
first determined, and the intersection of these two lines represented the failure load of
the pile with corresponding maximum failure settlement. Mind that when proof test
performed or the applied loads being relatively small, the second tangent line will
be very difficult to obtain. Under this condition, result handling based on experience
and local data required.
2.5 Field Tests 59
Fig. 2.55 Interpretation based on double tangent method (Samtani and Nowatzki 2006b)
DeBeer’s method—also known as DeBeer’s log–log method—is one of the other

extrapolations of SLTs’ results. As shown in Fig. 2.56, by plotting the load–settlement
data into logarithmic scales, DeBeer (1970) defined the failure load as the load
corresponding to the intersection of two distinct slopes. For plugging failure piles,
these two slopes can be visible; for example, as shown in the Fig. 2.56, the ‘turning
point’ of 7560 kN is discovered as the ultimate bearing capacity of the tested pile.
However, for piles that experience local failure, the capacity may be a range of values
(Paikowsky and Tolosko 1999).
Chin’s method operates under the assumption that the load–settlement relationship
is hyperbolic as illustrated in Eq. 2.40, and that the inverse slope, C1 , obtained from
the plotted the settlement/load versus settlement gives the capacity of a single pile
as provided in Eq. 2.41 (Chin 1970). An example is provided in Fig. 2.57, where the
slope for pile with label P80 is determined as 0.0002 from the plotted settlement/load
versus settlement diagram. The ultimate capacity of the pile is: Qu = 1/0.0002 =
5,000 kN. Given that this method is based on a mathematical relationship, the capacity
value of a single pile can pass beyond the maximum applied load from the SLTs.
This interpretation may provide overestimated values.
s
= c1 s + c2 (2.40)
Q
1
Qu = (2.41)
c1
where:
s = settlements from SLTs.
lg Load (kN)
500 5000 50000
0.1
1
lg Displacement (mm)
10
100
1000
Fig. 2.56 Interpretation based on DeBeer’s method
0.0045
0.004 y = 0.0003x + 0.0001

R² = 0.996
Settlement/Load (mm/kN)
0.0035
0.003 y = 0.0002x + 0.0002

R² = 0.9645
0.0025 P80
0.002 P40
0.0015 P12
y = 0.0001x + 0.0003
0.001
R² = 0.9133
0.0005
0
0 5 10 15 20
Settlement (mm)
Fig. 2.57 Interpretation based on Chin’s method

2.5 Field Tests 61
Q = loads applied from SLTs.

Qu = ultimate bearing capacity of single pile.
Yang and Xiao (2011) pointed out that, theoretically, Chin’s method illustrates
two lines, as depicted in Fig. 2.58. The reciprocal of these first gradient is the shaft
capacity before the end bearing capacity is mobilized, and the reciprocal of second
gradient represents the total bearing capacity.
For the results of uplift SLTs, the typical presentation is similar to compressive
SLTs, as shown in Fig. 2.59. However, a widely accepted method for ultimate capacity
in uplift load testing has not been published. Fuller (1983) pointed out that the criteria
for uplift capacity is related to elastic lengthening, as illustrated in Fig. 2.60. The
offset line of this elastic lengthening will intersect with the displacement-load curve,
0.009
0.007 y = 0.0005x + 0.0036

Qu= 1/0.0005
0.005 Qs
Qu
0.003 y = 0.0014x + 0.0013

Qs = 1/0.0014
0.001
0 2 4 6 8 10 12 14
Settlement (mm)
Fig. 2.58 Determination of shaft and end resistance based on Chin’s method
Fig. 2.59 Typical results of 25

uplift SLT
20
Movement (mm)
15
10
0
0 1000 2000 3000 4000 5000 6000
Uplift Load (kN)
Fig. 2.60 Typical tension load test for load–movement curve (Hannigan et al. 2006)
which represents the failure load and corresponding failure settlement. FHWA NHI-
16–009 (2006) recommended the offset value is to be 4.0 mm, and the uplift design
load can be equal to half to two-thirds of the failure load.
Another method used to determine the ultimate uplift bearing capacity of
non-failure uplift SLTs is the modified Mazurkiewicz method (Chim-oye and
Marumdee 2013). The assumption in this method is that the nominate settlement
(load/settlement) value is equal to 0 when the loading is small, and the settlement is
very high. Therefore, as shown in Fig. 2.61, a line goes through the y-axis, and the
intersection of two lines illustrates the uplift ultimate bearing capacity of the pile.
8000
7000
Ultimate Uplift Bearing Capacity
6000
5000
Load (kN)
4000
3000
2000
y = -17.347x + 6298.6
1000 R² = 0.9431
0
0 100 200 300 400 500 600
Load/Displacement (kN/mm)
Fig. 2.61 Modified Mazurkiewicz method

2.5 Field Tests 63
Referring to Reese (1984) and Wang and Reese (1993), a report for lateral SLTs
should contain a presentation such as pile head’s deflection-load curve (Fig. 2.62)
and deflection along the pile (Fig. 2.63). Compared with the compressive and uplift
SLTs, the interpretation and analysis of horizontal SLTs are much more complicated.
The ordinary results presentation (Fig. 2.62) can illustrate the lateral movement under
increment horizontal loads; however, it is better to get the ultimate bearing capacity
when cyclic loads being applied. Further, when researching the deflection behavior
of pile foundation, the lateral loads are commonly distributed along the pile, results in
Fig. 2.63 provides limited information because the applied load is a kind of point load
Fig. 2.62 Typical lateral load test for pile head load–deflection curve (Hannigan et al. 2006)
Fig. 2.63 Deflection

behaviour versus depth
(Kyfor et al. 1992)
acting on pile head. Note that the case study referring to cyclic horizontal load test
is presented in Chap. 4, the result presentation and interpretation is also included.
Another case study referring to the deflection behavior of pile by performing the
inclinometer monitoring is provided in the Chap. 5.
2.5.2 Dynamic Load Test
SLTs provide the most actual results for pile capacity and are accepted worldwide;
however, these tests cannot be used to determine the long-term settlement and down-
drag from consolidation and setting of soils. Further, these tests are costly and time
consuming. For example, if performing a compressive SLT, it is necessary to transfer
a reference beam or weighted platform (may exceed 900 tons) and cast reaction piles
(two to four piles). In addition, the duration of maintained load tests may exceed
to 24 h during some loading stages. In contrast, dynamic testing requires much less
time and money (Cheney and Chassie 2000).
The development of dynamic load test techniques started with a thesis project at
Case Institute Technology—known as the ‘case method’. During 12 years of research,
the dynamic load analysis was improved, including field solutions and a numerical
method called the ‘case pile wave analysis program’ (CAPWAP). These previous
projects and results can be found in Goble and Rausche (1970) and Goble et al.
(1975). Recent updated research information can be found in Rausche et al. (2004).
The application of dynamic testing facilitated the determination of pile soil
behavior, such as determining the soil resistance distribution by CAPWAP, eval-
uating the performance of pile foundations during the driving process, obtaining the
installation stresses in piles and determining the pile integrity. As shown in Fig. 2.64,
at least two strain transducers and two accelerometers need to be installed onto the
surface of the tested pile, and need to be bolted to diametrically opposite sides of the
pile (Hannigan et al. 2016). These transducers and accelerometers are used to record
the strain and acceleration during pile dynamic tests. The data are then transferred
to a data acquisition system, called the Pile Driving Analyzer (PDA), which finally
converts strain (ε) and acceleration (a) into force (F) and velocity (V ). Based on
Eqs. 2.42 and 2.43, PDA can provide a diagram of force and velocity curves versus
time, as illustrated in Fig. 2.65.
Ft = E Aεt (2.42)
Vt = at dt (2.43)
where:
Ft = converted force by PDA.
E = elastic modulus of pile.
2.5 Field Tests 65
Fig. 2.64 a Schematic diagram of apparatus for dynamic test; b Strain transducer and accelerometer
installed on pile surface (Hannigan et al. 2016)
Fig. 2.65 Typical force and velocity traces (Hannigan et al. 2016)
A = cross-section of pile.
εt = recorded strain by strain transducers.
Vt = converted velocity by PDA.
at = recorded acceleration by accelerometer.
The dynamic load test was developed based on the theory of wave mechanics. A
wave speed, C, can be created by a striking from a mass onto a uniform elastic rod,
in which the cross-section area is A and the elastic modulus is E. The force, F, is
generated at the surface of the rod after the strike, and this force will compress the
other part of the rod with an acceleration, a, where the particles of this rod attain a
velocity of V. If there is no resistance effect on this rod, the force wave (equal to
EA/C) and material wave (C) will travel down to the end of rod at a time of L/C
(length of rod is L), and these waves will reflect back to the rod head at a time of
2L/C.
If there is no resistance at the end of the rod, as illustrated in Fig. 2.66a, tensile
wave refection occurs, the force becomes zero and velocity doubles. If there is a
fixed end, compression wave reflection occurs, the velocity becomes zero and the
force magnitude doubles, as shown in Fig. 2.66b. Considering the rod as a pile with
small and large soil resistances at a depth of A and B (Fig. 2.66c), where the strike
is impacted on the pile head, the recorded force and velocity data versus time are
provided in Fig. 2.67.
Before time 2A/C (A is the depth in meters, not the cross-section area), the force
and velocity are proportional since there is no resistance. After 2A/C, the force wave
will increase and the velocity wave will decrease slightly because of the small soil
resistance. Similarly, the reflection wave will be affected by large soil resistance at
the time 2B/C, the force wave will increase, and the velocity wave will decrease
(a) Free end of rod (b) Fixed end of rod (c) Free end of pile
Fig. 2.66 Wave mechanics of rod and pile

2.5 Field Tests 67
Fig. 2.67 Soil resistance effects on force and velocity records (Hannigan 1990)
by a large magnitude. When the wave passes to the pile end, which is a free end
condition, the force wave will decrease and the velocity wave will increase (tensile
wave reflection).
The CASE method investigation was conducted at Case Western Research Univer-
sity. Based on one-dimensional wave propagation, the assumption of closed form
solution and pile material is linear elastic. The total static and dynamic resis-
tance during driving, RTL, is governed by Eq. 2.44. Based on the findings of the
dynamic resistance component function proposed by Goble et al. (1975), the static
pile capacity, RSP, can be determined based on Eq. 2.45:
1 EA
RT L = + /C (2.44)
2 Ft1 + Ft2 2 Vt1 + Vt2

Vt EA
RS P = RT L − J 1 + Ft1 − RTL (2.45)
C
where:
F = force obtained from gauge location.
V = velocity obtained from gauge location.
E = elastic modulus of pile.
A = cross-section area.
C = wave speed of pile material.
L = length of pile below gauge location.
t1 = time of initial impact.
t2 = time of reflection of initial impact from pile toe.
J = dimensionless damping factor referring to soil type near pile toe.
The CAPWAP method can provide more rigorous static load capacity of piles and
soil resistance along the pile as shown in Fig. 2.68. Similar to the CASE method,
measured force versus time should first be plotted. A computed force is then used
to match the obtained signal with the assumptions that: (i) the pile is simulated by a
series of pile segments, and (ii) soil resistance is simulated by elasto-plastic springs
Fig. 2.68 Schematic of CAPWAP analysis method (Hannigan et al. 2016)
and dynamic resistance. Adjustments and a repeat process are required for the soil
model, and this model is refined until no further agreement can be determined. Finally,
this process can be terminated and the soil model can be used to evaluate the capacity
of the tested pile. A detailed case study referring to dynamic load tests as well as
CAPWAP method will be provided in Chap. 3.
2.5.3 Osterberg Cell Test
The Osterberg cell test, also known as the O-cell test, is a proprietary method to
determine the capacity of driven piles and cast-in-place piles. Different to static and
dynamic load tests, O-cell tests do not need a reaction frame and anchor system. The
axial loading of these tests is provided by an expendable jack and load cell cast inside
the shaft. As shown in Fig. 2.69, the O-cell, which is welded with bearing plates,
is used to apply loading upwards and downwards. The vertical displacement of the
tested pile can be recorded by displacement transducers. The example referring to
the typical presentation (displacement versus loading) is provided in Figs. 2.70 and
2.71.
As shown in Fig. 2.70, after the maximum loading of 2,400 kN being applied,
the settlement of the pile end is 0.5 mm and the vertical upward movement of the
2.5 Field Tests 69
Fig. 2.69 Schematic of test pile by O-cell testing. Source www.loadtest.com
pile shaft is 2.7 mm. The shaft resistance of soil layers is fully developed. As shown
in Fig. 2.71, the ultimate end bearing is reached because the shaft displacement is
small and end displacement is large.
This test has many advantages compared with the traditional tests. First, O-cell
equipment can apply large loading of up to 27 MN (Samtani and Nowatzki 2006b).
Second, this test saves time because the testing can be conducted once the concrete
has reached suitable strength. Third, this test is economic and can provide accurate
data. Other advantages include convenient performance for offshore projects, high
safety (no need to use heavy reaction system).
There are also some disadvantages for O-cell test performance. First, the O-cell
device must be installed prior to construction or driving. Second, once the shaft has
reached the ultimate condition, the ultimate end resistance cannot be obtained, and
vice versa. Third, if the capacity of the O-cell is inadequate, the test fails because the
Fig. 2.70 Typical O-cell results—shaft failure. Source www.loadtest.com
Fig. 2.71 Typical O-cell results—end failure. Source www.loadtest.com
device has already been cast inside the pile. Fourth, this device is not reusable after
the test is finished. Finally, the tests on H-piles and sheet piles are not applicable.
2.5.4 Statnamic Test
The statnamic test method is also a proprietary method that was developed by the
Berminghammer Foundation Corporation. This test contains four phases, as shown
in Fig. 2.72. Phase I involves the test setup. The reaction mass is placed onto the test
shaft. A small volume of propellant is placed inside the pressure chamber. After the
2.5 Field Tests 71
A = pile to be tested
B = load cell
C = cylinder pressure chamber
D = piston
E = platform
F = silencer
G = reaction mass
H = gravel container
I = gravel chamber
J = optical measuring system
Fig. 2.72 Schematic of test pile by statnamic testing. Source www.profound.nl
propellant is ignited, the generated high-pressure gas accelerates the reaction mass
upwards, and the downwards reaction force occurs in Phase II. In Phase III, the mass
is forced upwards and the gravel fills the gravel container. Finally, in Phase IV, the
gravel flows over the pile head as a layer. This layer then catches the reaction mass
and transfers the impact forces. During Phases I to IV, the laser beam records the
movement of the pile head, and a diagram of load versus time and displacement can
be obtained, as shown in Fig. 2.73a. The typical result of load versus displacement
can be determined by Profound’s Foundation Pile Diagnostic System, as illustrated
in Fig. 2.73b.
The theory of this statnamic test is based on Newton’s three laws of motion. The
measured statnamic force, F stn , is the sum of the inertia force, F a , and soil resistance,
F soil , which is governed by Eq. 2.46. The soil resistance is the sum of the static soil
resistance, F u ; damping force from soil, F v ; and pore water pressure resistance, F p ,
so the measured statnamic force is governed by Eq. 2.47. Under the assumption of
pore water pressure resistance is included as part of the damping force, and pore water
pressure can be ignored (less than 5%). The static soil resistance is then governed by
Eq. 2.48:
Fstn (t) = Fsoil (t) + Fa (t) (2.46)
Fstn (t) = Fu (t) + Fv (t) + F p (t) + Fa (t) (2.47)

Fig. 2.73 Typical result presentation by statnamic test (Courtesy of Berminghammer Foundation
Equipment)
Fu (t) = Fstn (t) − k × u(t) − c × v(t) (2.48)
where:
Fstn = measured statnamic force.
Fa = inertia force.
Fsoil = soil resistance.
Fu = static soil resistance.
Fv = damping force from soil.
F p = pore water pressure resistance.
k = spring stiffness in N/m.
u(t) = measured displacement in m.
c = damping factor in Ns/m.
v(t) = velocity du/dt in m/s.
Compared with SLTs, which are time consuming and cumbersome, the statnamic
test can provide the same loading using a mass that is equal to only 5 to 10% of an
equivalent SLT. Compared with O-cell testing, the statnamic load test does not need to
cast the device inside the pile, and the vertical load is applied for a duration of 120 ms.
This testing method can determine the behavior of high-capacity piles with capacity
over 5.0 MN. The commercial test equipment nowadays is capable of producing
loads of up to 40 MN. Different to the dynamic load test, piles tested by the statnamic
load test are not dominated by stress wave propagation, and the created acceleration
is relatively low. Moreover, this statnamic test can directly provide measurements,
instead of processing the data by experienced engineers, as does the dynamic load
test.
2.6 Concluding Remarks 73
2.6 Concluding Remarks
This chapter reviewed subsurface explorations firstly, which included laboratory and
in-situ tests. It presented the test setup, test procedures and case studies for performing
the tests to provide a more comprehensive geotechnical understanding. This was
followed by a discussion of the methods used to determine the soil parameters that
were applied in previous projects’ designs, as well as the literature review of the
different types of piles, considering configuration, construction methods and inves-
tigations undertaken in the past. This chapter also reviewed the analytical methods
used to determine pile capacity based on the parameters acquired from laboratory
and in-situ tests.
Additionally, this chapter reviewed the SLTs associated with compressive, uplift
and lateral loads, including test setup, procedures based on the codes’ requirements,
presentation of typical results and extrapolations of SLT results. Finally, this chapter
reviewed the theories and methods used to determine the ultimate bearing capacity
of a single pile based on dynamic load tests, Osterberg cell test, and statnamic test.
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Chapter 3
Field Tests of Post Grouted Concrete Piles
With the soaring requirement of building space in metropolises, high rise building
development is indispensable and thus deep piles are imperative. Pile length can range
from 20 to 60 m due to large loads from upper structures needing to be transferred into
soils. Apart from designing long pile foundations, which can provide more friction
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tested in western Bangladesh. The tests results showing increments of end bearing
and shaft bearing capacity and decrements of base settlement (Castelli and Wilkins
2004). Another example is a skyscraper project in London, where the diameter of
piles was 2.4 m with a length of 63 m (Patel et al. 2015).
Compared to bored piles, precast concrete piles are frequently more costly
(Tomlinson and Woodward 2007). However, during the process of pile foundation
construction, some soil deposits can remain at the base area after drilling the hole.
The remaining deposit, which is accumulated by collapse during drilling, will lead
to a decrease in the end resistance capacity of the pile, and an increase of pile settle-
ment. This potential for reduced pile capacity was affirmed during construction and
performance of static load tests in Taiyuan City, China where capacities were much
lower than the designed requirement (Shi 2005). Furthermore, the drilling operation
can loosen the soil underneath the base of the bored hole, which can also lead to
excessive working load settlement. Engineers nowadays use admixtures like poly-
meric slurry or bentonite as a support to avoid the collapse of soils during drilling
construction; this, however, creates other issues. The use of these materials may lead
to capacity decrease of pile shafts because these admixtures create a layer between
soil and pile when combined with soil and water, which consequently leads to a
decrease of friction resistance of the pile. Li et al. (2000) reported that this admixture
layer or composite layer could decrease the pile bearing capacity by 30–40%.

https://doi.org/10.1007/978-981-33-6183-6_3
82 3 Field Tests of Post Grouted Concrete Piles
Such limitations, however, can be addressed with a base and shaft grouting tech-
nique (or post grouting technique). This construction technology of cast-in situ piles
started in the mid of 1970s. It can generally be categorized into a flat jack system
which consists of grout delivery pipes connected to a steel plate with a rubber
membrane, and a sleeve-port system that consists of 2–4 U-tubes installed at the
bottom of the pile. This U-tube is covered by rubber and can be arranged in various
configurations (Dapp et al. 2006). Cement admixture which is blended into various
ratios is the popular material for grouting. It is injected to the pile toe through U-
tubes under high pressure, consequently restoring the original density of the base soil
and reducing the settlement of the pile when loading transfers from the upper struc-
ture (Tomlinson and Woodward 2007). It is worth noting that due to high pressure,
partial cement admixture will force the admixture layers upwards, which leads to a
recovering and improvement of shaft resistance around the pile as shown in Fig. 3.1.
Field geotechnical testing generally includes dynamic load tests, compressive and
uplift static load tests and O-Cell tests (Zhussupbekov and Omarov 2016). Several
compressive static load tests of base grouted piles have been conducted. Due to
grouting being more popular in granular soils, some researchers have focused on
the soil type effect and load transfer mechanism (Ho 2003; Rollins et al. 2008;
Thasnanipan et al. 1998). Other researchers have concentrated on the construction
process and capacity improvements (Wang et al. 2006, 2015).
Recent tests and numerical simulations have shown that the post grouted concrete
piles can increase twice the ultimate capacity of defected piles, and increase about
20% compared to normal piles (Nguyen et al. 2012). Compared to grouted piles and
Cement layer created by grouting
Fig. 3.1 Grouting concrete pile with cement layer (Zhou et al. 2017b)
3.1 General Introduction 83
non-grouted piles, results have shown grouting could increase 18–19% of shaft resis-
tance and 63% base resistance (Wang and Zhang 2013). Through tests on seven piles,
Zhang et al. (2012c) have pointed out that post technology has had an enhancing effect
on shaft and base resistance but the increase in pile capacity was affected by slender-
ness ratio. For the research of grouting techniques on precast piles, a recent exper-
imental and FEM study has proposed the capacity estimation and method of grout
pressure (Thiyyakkandi et al. 2013). Several other projects with application of base
grouting techniques have been provided by Sinnreich and Simpson (2013); however,
these results are ambivalent because some projects illustrate increased capacity of
pile by grouting and some projects do not.
A large number of investigations have been carried out to obtain the behaviors
of piles under uplift loading. Some researchers have focused on the behaviors of
different types of piles such as precast piles, cast-in situ bored piles and steel pipe piles
etc. (Huo et al. 2013; Kusakabe et al. 1994; Shelke and Patra 2011). Other researchers
have concentrated on the uplift behavior of piles in cohesionless soil (Gaaver 2013;
Thiyyakkandi et al. 2014; Verma and Joshi 2010). Through experimental tests, a
study revealed that pile behavior under uplift force depends mainly on pile embed-
ment depth-to-diameter ratio and soil properties (Gaaver 2013). A further study
investigated pile behavior under combined uplift and lateral loading (Madhusudan
Reddy and Ayothiraman 2015). Recent research on belled and multi-belled piles in
dry loose sand has also been conducted, with uplift resistance being found to increase
up to 60% in comparison with straight piles (Moayedi and Mosallanezhad 2017).
The investigations of base-and-shaft grouted uplift field tests, capacity improvement,
as well as load transfer mechanism, however, remain limited.
To determine the load transfer behavior of piles under compressive and uplift static
load tests, wire vibration strain reinforced bars are commonly used. The disadvantage
of these strain gauge applications is that labor is required and it is time consuming.
In order to avoid the damage of strain gauges, the transfer of the steel cages is always
slow, which may lead to project being delayed. With technology developing, dynamic
load test appears which can provide good results with load transfer information.
Compared to static load tests, Pile Driven Analyzer (PDA) tests are relatively cheaper
and time saving (Budi et al. 2015). A popular program is the Case Pile Wave Analysis
Program or CAPWAP. The bearing capacity and the stress distribution along the pile
shaft and toe as well as simulated static load tests can be determined by matching
signals obtained from the dynamic load tests (Likins et al. 1996). The strain sensors
and accelerometers are used to measure the force and velocity inside the pile after
applying load provided by a hammer. The load transfer mechanism can be obtained
when the matched wave approaches the acquired wave onsite. Investigations of driven
polymeric piles using dynamic load tests were conducted in Elizabeth, New Jersey,
providing load testing installation and comparison between static and dynamic load
tests. That paper also focused on the possible application of plastic piles under axial
loading (Robinson and Iskander 2008).
As mentioned above, there is limited research on compressive and uplift loaded
grouted piles. This Chapter, therefore, aims to investigate the ultimate capacity of
post grouted concrete piles under compressive and uplift static load and dynamic
load tests. By comparing the compressive capacity among piles with shaft-and-base
grouting, base grouting only and no grouting, the increment of pile capacity by
grouting technique is determined. The increment of uplift capacity by grouting is also
observed by uplift static load tests. In addition, the interpretation of the static load test
result including the double tangent, DeBeer’s, Chin’s and Mazurkiewicz’s methods
are provided. This Chapter also provides dynamic load tests for load transfer mecha-
nism analysis. It should be noted that this Chapter investigated the piles behavior and
shaft mechanisms under different grouting techniques, which has not been reported
based on results’ comparison between static load tests and dynamic load tests.
3.2 Site Conditions
The examined project aimed to build a 22-level office building at a height of 82.95 m,
with a construction area of 50.8 × 42.2 m2 . The construction site was located in Jinan
city, China. The subsurface exploration was determined through laboratory and in situ
tests. In situ SPTs and CPTs and laboratory consolidation tests, direct shear tests and
triaxial tests were conducted based on the Chinese Standard for Soil Test Method
(GB/T 50123-1999 1999) and Code for Investigation of Geotechnical Engineering
(GB 50021-2001 2001), respectively.
A hammer with a weight of 63.5 kg was selected for the SPTs, and counts of
300 mm penetration were recorded as the N-value or N 63.5 -value. The soil classi-
fication was determined based on these N-values and soil sampling: Clause 3.3 of
Soil Classification (GB 50021-2001 2001). Laboratory tests including consolidation
test, shear test and triaxle tests (CU and UU) were conducted for soil characteris-
tics of specific gravity, relative density, porosity, void ratio, saturation ratio, water
content, plastic limit, liquid limit, plasticity index, liquidity index, cohesion and fric-
tion angle. These tests were conducted based on Clauses 5, 8, 9, 13, 14, 16 and 18
of Soil Test Method respectively (GB/T 50123 1999). Based on the borehole logs,
the Simplified soil layers were then discovered as follows, with the properties of soil
layers illustrated in Table 3.1.
1. miscellaneous fill: loose, containing gravel and rubble, diameter ranging from
2.0 to 7.0 mm, average thickness of layer of 2.61 m; 2. silty clay: medium stiff,
yellowish with high plasticity, average thickness of 3.0 m; 3. gravel: medium dense,
average thickness of 3.5 m; 4. silty clay: medium stiff, low plasticity, average depth
of 3.5 m; 5. gravel: medium dense with dark clay, diameter of gravel ranging from
5.5 m; 6. residual soil: dense, layer containing iron and manganese oxides, average
thickness of 5.0 m; 7. weathered diorite: medium to dense, average depth of 4.0 m;
8. highly weathered diorite: medium to dense, average thickness of 1.7 m; 9. bearing
stratum: very dense, rock-quality designation = 40, average strength of 10.65 MPa.
This project required underground carparks, so the deep excavation construction
is required (excavation depth of 9 m). After the soil was removed, the cast-in situ
piles were cast and surrounded by four layers. The compressive and uplift loaded
piles with labels of P51, P121, P126, P15 and P16 were adjacent with each other, so
3.2 Site Conditions 85
Table 3.1 Simplified soil layers

Soil layers Depth (m) Average N-value γ (kN/m3 ) C (kPa) ϕ (o ) F ak (kPa)
Miscellaneous fill 2.61 5 19.0 5.0 10.0
Silty clay 3 7 18.9 16.6 19.7 150
Gravel 3.5 21 20.0 5.0 30.0 260
Silty clay 3.5 21 19.3 18.2 16.8 200
Gravel 5.5 22 20.0 8.0 30.0 280
Residual soil 5 23 18.0 5.0 28.0 200
Weathered diorite 4 41 18.5 10.0 30.0 240
Highly weathered 1.7 43 25.5 12.0 48.0 450
diorite
Bearing stratum N/A >50 28.8 12.0 54.0 2000
the soil condition was similar. The soil layers were residual soil with a thickness of
5 m, weathered diorite with an average thickness of 4 m, highly weathered diorite
with an average thickness of 1.7 m and bearing stratum.
3.3 Pile Description
Three and two concrete bored piles with labels of P51, P121, P126, P15 and P16
were cast for conducting compressive and uplift SLTs, respectively. All these piles
were made of the same concrete and reinforcement, with the same diameter and
length. Detailed information is provided in Tables 3.2 and 3.3. However, the used
grouting technology was different. In contrast to the base grouting, the base-and-
shaft grouting system contained a circular tube (circular diameter R = 800 mm) with
tube diameter of 5 mm and an additional shaft grouting pipe, as shown in Fig. 3.2,
which was welded to steel cages. This grouting tube was drilled with small holes
with spacing of 150 mm, and covered by a membrane that was used to avoid the fine
sand blocking the grouting.
As depicted in Table 3.3, during the construction process, P126 was cast using
base-and-shaft grouting technology and P121 was cast using base grouting only. To
Table 3.2 Reinforcement properties of tested piles

Steel bars Diameter (mm) Type Spacing (mm) f y (MPa) Es (MPa) v
Longitudinal 28 HRB400 N/A 360 200,000 0.3
Stirrups top 4 m 16 HRB400 100 360 200,000 0.3
Stirrups of other 16 HRB400 200 360 200,000 0.3
part
Stiffening ring 14 HRB400 2000 360 200,000 0.3
Table 3.3 Description of test piles

Pile Technology used Types of tests Length Diameter Concrete Compressive
label (m) (mm) type strength (kPa)
P51 No grouting Compressive 19.5 800 C50 50
SLTs
P121 Base grouting Compressive 19.5 800 C50 50
SLTs
Dynamic load
test
P126 Base-and-shaft Compressive 19.5 800 C50 50
grouting SLTs
Dynamic load
test
P15 Base-and-shaft Uplift SLTs 19.5 800 C50 50
grouting
P16 No grouting Uplift SLTs 19.5 800 C50 50
Base Grouting Pipe
Shaft Grouting Pipe
Fig. 3.2 Grouting pipe system of concrete pile
obtain results of a trial comparison, P51 was cast without any grouting, these three
piles were designed to apply the compressive force. Previously, all these three piles
were designed to be uplift loaded after compressive SLTs being completed, but with
consideration of loading influence by compressive loads, the team determined to cast
other two piles with labels of P15 and P16 for the uplift loading tests. The difference
3.3 Pile Description 87
Fig. 3.3 Mud slurry support
between these two piles was that P15 was applied with base-and-shaft grouting
technology. Further, the dynamic load tests were conducted on P121 and P126 to
determine load transfer behaviors. All piles were cast with application of mud slurry
supporting, as shown in Fig. 3.3. Note that the mud slurry replaced the admixture
slurry during drilling process, which was only for economic considerations.
3.4 Static Load Test Setup
SLTs were conducted based on the Chinese Technical Code for Testing of Building
Foundation Piles (JGJ 106-2014 2014). The maintained load SLT procedure was
used. According to the code, the maximum applied loads should be twice the designed
loads, with increment value of 1/10 QMax . i.e., for compressive SLTs: QMax = 2 ×
4,500 kN; increment loading = 900 kN; for uplift SLTs: QMax = 2 × 1,500 kN;
increment loading = 300 kN.
As shown in Figs. 3.4 and 3.5, four and two hydraulic jacks (QF630T-20) were
used to provide loads to piles for compressive and uplift SLTs, respectively. For
the compressive SLTs (P51, P121, P126), the loading started at 1,800 kN, with
increments of 900 kN, and released with decrements of 1,800 kN back to 0 kN. For
the uplift load tests (P15, P16), the loading started at 600 kN, with increments of 300
kN, and released with decrements of 600 kN back to 0 kN.
Following Chinese codes GB 50202-2002 (2002) and JGJ 106-2014 (2014), for
the compressive SLTs, the settlement was recorded with intervals of five, 15, 30,
45 and 60 min (and 1 h later if required) under the maintained loads. The followed
loading could only be applied when the variable quantity of the recorded movement
Fig. 3.4 Compressive SLTs setup
Fig. 3.5 Uplift SLTs setup
was less than 0.1 mm. The test needed to stop when the recorded settlement was five-
times of the previous settlement or the recorded settlement was observed over 40 mm.
Same with the compressive SLTs, for the uplift SLTs, low-speed maintenance tests
were conducted. The applied load was maintained until the rate of axial movement
did not exceed to 0.1 mm. After each load was applied, the vertical movement of
the piles was recorded with intervals of five, 10 and 15 min, and 30 min. If the
accumulated time exceeded 1 h, record movement of pile with interval of 30 min.
The loading could be terminated in the case of the following conditions:
3.4 Static Load Test Setup 89
1. The applied loading was equal to the value of 0.9 times the ultimate strength of
reinforcement;
2. Quintuple movement change between two loading stages was discovered;
3. The pile head movement increased up to 100 mm.
3.5 Dynamic Load Test Setup
The setup of the dynamic load tests is illustrated in Fig. 3.6. For this test, the dynamic
penetration analyzer (PDA) (RS-1616 K[S]) was used for signal analysis. Two accel-
erators (SY-2) and two strain transducers (CYB-YB-FIKA) were installed symmet-
rically onto the pile surface with a distance equal to or over the value of 1.5 D
(D—diameter of pile) from the pile head. The wave signal was recorded by acceler-
ators and strain transducers after a heavy hammer dropped onto the pile head, and
the PDA matched the signal based on the CAPWAP.
Fig. 3.6 Dynamic load tests

setup
3.6 Static Load Tests Results
3.6.1 Compressive Static Load Tests
3.6.1.1 Load Settlement Curves
Figure 3.7 provides the load-settlement (Q-s) curves of three compressive loaded
bored piles. After the maximum loading of 9,000 kN was applied, the maximum
pile head settlement of the non-grouted (P51), based-and-shaft grouted (P126), and
base grouted (P121) piles were 11.68, 13.37 and 11.74 mm, respectively. From the
curves, it represents approximately linear from the loading stage. The final settle-
ments of P51 and P121 were discovered to be 8.11 and 7.36 mm, respectively. This
was likely because the bearing stratum was diorite, and the base grouting does not
effectively improve the settlement. Also, as shown in this Figure, P126 showed a
smaller permanent settlement of 2.8 mm. This demonstrated that base-and-shaft
grouting will decrease the final settlement of a pile foundation. Also, this illustrates
that the base-and-shaft grouting has changed characteristic of the pile-soil system
from semi-plastic state to a relatively elastic rigid body, and the maximum settlement
is primarily caused by the pile-soil elastic compression.
Fig. 3.7 Results of Q-s Compressive Load (kN)

curvatures of compressive
0 2000 4000 6000 8000 10000
loaded piles
0
4
Settlement (mm)
10
12
P51
14
P121
P126
16
3.6 Static Load Tests Results 91
Fig. 3.8 s-lgQ curves of Compressive Load-lgQ (kN)

tested piles
1 10 100 1000 10000
0
6
Settlement (mm)
8
10
12
P51
14 P121
P126
16
3.6.1.2 Settlement lg (Load) Curves
Through the settlement-lg load (s-lgQ) curvatures shown in Fig. 3.8, the ultimate
compressive bearing capacities of the piles were determined, albeit with difficulty, to
be 6,800, 7,100 and 8,000 kN, respectively. This was because the tests were ‘proof
tests’, which aimed to ensure the pile settlement being acceptable under the maximum
loading (equal to twice the design load). The non-failure settlements led to difficulty
in determining the capacity from the s-lgQ curves. The determination of ultimate
pile capacity needed more comprehensive analysis.
3.6.1.3 Settlement lg (Time) Curves
The settlement-lg time (s-lgt) curves of P51, P121 and P126 are provided in Figs. 3.9,
3.10 and 3.11, respectively. There were no conditions indicating an extreme settle-
ment decreasing trend under a specific loading stage, and the change of settlements
between the adjacent two loading stages was relatively close. Further, it was discov-
ered that, in each loading, the settlement did not change much as time went by, and
this demonstrated that all three piles were stable after the loading was applied. The
ultimate bearing capacity of the pile was unknown via analyzing these diagrams;
thus, further comprehensive analysis was required.
lg Time (min)
5 50 500
0
4 1800kN
Settllement (mm)
2700kN
6 3600kN
4500kN
8
5400kN
10 6300kN
7200kN
12 8100kN
9000kN
14
Fig. 3.9 s-lgt curves of P51
lg Time (min)
5 50 500
0
4
1800kN
Settllement (mm)
6 2700kN
3600kN
8
4500kN
10 5400kN
6300kN
12
7200kN
14 8100kN
9000kN
16

lg Time (min)
5 50 500
0
4 1800kN
Settllement (mm)
2700kN
6 3600kN
4500kN
8
5400kN
10 6300kN
7200kN
12 8100kN
9000kN
14
3.6.1.4 Interpretations of Static Load Tests
Double Tangent Method
As illustrated in the General Principles and Practice in Chap. 2, the double tangent
method determines two tangent lines from the load–settlement curves, and finds
the interaction between these two tangent lines. The point shows the corresponding
ultimate capacity and ultimate settlement. In this study, the capacities of the three
compressive loaded piles with labels of P51, P121 and P126 were determined to
be 7,560, 7,700 and 8,250 kN, with corresponding ultimate settlements of 6.0, 6.2
and 8.0 mm, respectively, as shown in Figs. 3.12, 3.13 and 3.14. It should be noted
that these values are conservative because of these tests are the proof test instead of
failure test. Under this condition, the double tangent method can be appropriately
adjusted based on the local experience.
DeBeer’s Method
The curves based on DeBeer’s method are provided in Fig. 3.15. As shown in the lg
settlement-lg load diagram, finding the intersection of two distinct slopes of each pile
is difficult. Similar to the settlement to lg load (s-lgQ) results, the ultimate bearing
capacities of those piles were determined with difficulty to be 7,600, 7,900 and 8,300
kN, respectively.
Compressive Load (kN)

0 2000 4000 6000 8000 10000
0
4
Settlement (mm)
10
12
14
Fig. 3.12 Double tangent method of P51

0 2000 4000 6000 8000 10000
0
2
4
Settlement (mm)
6
8
10
12
14
16
Davisson’s Offset Method
By calculation of the QL/EA (Eq. 2.37), the elastic deformation line can be obtained.
As shown in Fig. 3.16, the elastic deformation line and the parallel line based on
Davisson’s offset formula (Eq. 2.39) is plotted in the Q-s curves. It can be seen that
there is no intersection between the Q-s curves and the offset line. This is because
the total settlement of these piles was small under maximum applied loads. Hence,
this method did not fit this case to determine the ultimate bearing capacity. However,
this provided information that the ultimate bearing capacity of the pile was greater
than the maximum applied loads from the SLTs. This result also illustrates that for

0 2000 4000 6000 8000 10000
0
4
Settlement (mm)
10
12
14
Fig. 3.15 DeBeer’s method lg Load (kN)

of tested piles 100 1000 10000
0.1
1
lg Settllement (mm)
10
P51
P121
P126
100
the proof test result, the Davisson’s offset method may not be used for interpretation
to determine the ultimate bearing capacity.

0 2000 4000 6000 8000 10000
0
4
Settlement (mm)
8
P51
10
P121
12 P126
14 Elastic Line
Offset Line
16
Fig. 3.16 Davisson’s offset method of tested piles
0.0018
0.0016
0.0014
0.0012
0.001
0.0008
0.0006
0.0004
P51
0.0002 P121
P126
0
0 2 4 6 8 10 12 14 16
Settlement (mm)
Fig. 3.17 Chin’s method of tests piles
Chin’s Method
As discussed in the literature review in Chap. 2, the inverse slopes of /P results in
the failure value. In this study, the results of the three piles are provided in Fig. 3.17.
It was found that the ultimate bearing capacities of P51, P121 and P126 were 10,000,
Table 3.4 Results of piles under different interpretation methods

Pile Label Double tangent DeBeer’s Chin’s
Load (kN) Settlement Load (kN) Settlement Load (kN) Settlement
(mm) (mm) (mm)
P51 7,560 6.0 7,600 7.9 10,000 N/A
P121 7,700 6.2 7,900 8.3 11,111 N/A
P126 8,250 8.0 8,300 8.2 12,500 N/A
11,111 and 12,500 kN, respectively. Given that the determined ultimate loads were
beyond the maximum applied loading, the corresponding ultimate settlements were
unknown.
Based on the different interpretations, the ultimate bearing capacities of piles
suffering from compressive loading were determined. The ultimate bearing capacity
with corresponding ultimate settlement of these compressive loaded piles are
summarized in Table 3.4.
3.6.2 Uplift Load Static Load Tests
To determine if the piles achieved the design requirement when suffering from uplift
force and to determine the improvement of the base-and-shaft grouting technique,
two piles were selected for uplift SLTs—P15 and P16. As shown in Fig. 3.18, when
the maximum loading of 3,000 kN was applied, the maximum vertical movements of
18
16
14
Displacement (mm)
12
10
2 P15
P16
0
0 1000 2000 3000 4000
Uplift Load (kN)
Fig. 3.18 Results of Q-s curvatures of uplift loaded piles

18
16
14
Displacement (mm)
12
10
8
6
4
2 P15
P16
0
1 10 100 1000 10000
Uplift Load-lgQ (kN)
Fig. 3.19 s-lgQ curves of tested piles
P15 and P16 were 15.1 and 15.38 mm, respectively. These two tests also illustrated
that, when loading from 0 to 3,000 kN, the load line was approximately linear. This
result illustrated that these two piles could resist more loading. Further analysis was
required to determine the ultimate uplift capacity of the pile.
3.6.2.2 Settlement lg (Load) Curves
The analysis of the s-lgQ curves of these two uplifted piles indicated no plunging
points (the movement increase upward dramatically), as shown in Fig. 3.19. Also,
it can be found that the maximum upward displacements of both piles were very
small (less than 16 mm). This illustrated that the pile achieved the designed load, but
the maximum loading was not the ultimate bearing loading. Further research was
required to determine the ultimate bearing.
3.6.2.3 Settlement lg (Time) Curves
The settlement-lg time curves of the two tested piles that suffered from uplift loading
are provided in Figs. 3.20 and 3.21. The settlements of these two piles were stable
during each loading stages, and there was no dramatic displacement change between
two adjacent loading stages. Also, the maximum uplift movement for P15 and P15
is very small as shown in these figures. Further analysis was required to determine
the ultimate bearing capacity of these piles.
16
14
Displacement (mm) 12
600kN
10 900kN
8 1200kN
1500kN
6 1800kN
2100kN
4
2400kN
2 2700kN
3000kN
0
5 50 500
lg Time (min)
18
16
14
Displacement (mm)
12 600kN
10 900kN
1200kN
8
1500kN
6 1800kN
2100kN
4
2400kN
2 2700kN
0 3000kN
5 50 500
lg Time (min)
3.6.2.4 Interpretations of Static Load Tests
As illustrated in the literature review in Chap. 2, for uplift SLTs, an offset method
similar to Davisson’s offset method can be employed. The offset value is recom-
mended to be 4.0 mm, and the intersection of the uplift load–settlement curves
and offset line represents the ultimate load. Based on this method, the ultimate
bearing capacities of P15 and P16 were determined to be 1,500 and 1,700 kN, with
corresponding vertical displacements of 5.5 and 5.6 mm, respectively, as shown in
Fig. 3.22.
18
P15
16 P16
Elastic Line
Displacement (mm) 14
Offset Line
12
10
8
6
4
2
0
0 1000 2000 3000
Uplift Load (kN)
Fig. 3.22 Offset method of tested piles
The ultimate uplift bearing capacity of non-failure uplift SLTs can be determined
by the modified Mazurkiewicz method (Thanadol 1998). The assumption is that,
when the nominate settlement (load/settlement) value is equal to 0 (the loading is
small and the settlement is very high), then a line intersects with the y-axis, which
illustrates the uplift ultimate bearing capacity of pile. As shown in Figs. 3.23 and
3.24, two functions are determined, and, when x = 0, the ultimate uplift bearing
capacities of P15 and P16 were 6,299 and 5,442 kN, respectively.
8000
7000
6000
5000
Load (kN)
4000
3000
2000
y = -17.347x + 6298.6
1000
0
0 200 400 600
Fig. 3.23 Mazurkiewicz method of P15

3.7 Dynamic Load Tests Results 101
8000
7000
6000
5000
Load (kN)
4000
3000
2000 y=-13.136x+5442.1
1000
0
0 500 1000 1500 2000 2500
Fig. 3.24 Mazurkiewicz method of P16
3.7 Dynamic Load Tests Results
Two dynamic load tests were conducted on piles P121 (base grouting) and P126
(base-and-shaft grouting). Through CAPWAP, via many trials, the final matched
signals are provided in Figs. 3.25 and 3.26. Based on these two matched signals,
the simulated load–settlement curves, unit shaft resistance along the pile length and
load transfer characteristics are provided in Figs. 3.27 and 3.28. As indicated by
the dynamic load tests results, the ultimate bearing capacities of P121 and P126
were 9,071 and 9,851.5 kN, with corresponding maximum settlements of 26.37 and
18.93 mm, respectively.
As shown in Fig. 3.27, the shaft resistance of the base grouted pile was not
uniformly distributed along the pile, and, compared with the results illustrated in
Fig. 3.28, shaft grouting technology changed the shaft resistance and showed a rela-
tively uniform distribution. This phenomenon occurred because the shaft grouting
Force onsite
Matched Signal
Fig. 3.25 CAPWAP results of P121

Force onsite
Matched Signal
Fig. 3.26 CAPWAP results of P126
Fig. 3.27 Simulated load–settlement curves, unit shaft resistance and load transfer characteristic
of P121
Fig. 3.28 Simulated load–settlement curves, unit shaft resistance and load transfer characteristic
of P126
3.7 Dynamic Load Tests Results 103
changed the property of the soil from the pile shaft. It can also be seen that the shaft
resistances of P121 and P126 were 4,217.3 and 4,850.0 kN, respectively. From the
load transfer characteristic diagrams, it can be determined that the load decreased
along the pile length, yet increased with increasing applied loads. The comparison
of the load transfer characteristic diagrams in Figs. 3.27 and 3.28 also indicated that
the shaft grouting increased the shaft resistance of the pile.
All tested piles’ results are summarized in Tables 3.5 and 3.6. As shown in
Table 3.5, all these results obtained from different methods demonstrated that the base
grouting and base-and-shaft grouting improved the ultimate bearing capacity of the
piles. The ultimate load and settlement acquired from the double tangent method was
close to the outcome obtained from DeBeer’s method. Further, the results obtained
from Chin’s method and the dynamic method were close to each other.
For the friction resistance analysis of static uplift load tests and compressive
dynamic load tests, Table 3.6 demonstrates that the base-and-shaft grouted pile
possessed better capacity than did the base grouted pile and pile without grouting.
It can also be seen that the grouting technique improved the shaft capacity and end
bearing capacity of the piles. Further, compared with the shaft resistance results
obtained from the dynamic load tests, Mazurkiewicz’s method overestimated the
ultimate capacity of the uplift loaded piles.
This chapter provided the set-up of static and dynamic load tests, and various method-
ologies that were used to determine the ultimate compressive and uplift bearing
capacity under non-plunging failure conditions. Three types of pile foundations were
considered in this chapter, which were non-grouted pile, based grouted pile, and the
base-and-shaft grouted pile. By test result analysis, it illustrated that the results from
double tangent method and DeBeer’s method were close to each other, and Chin’s
method and dynamic method provided a close result.
This chapter also provided the load transfer characteristic of piles obtained from
dynamic load tests. Because the base grouting and shaft grouting techniques were
the same, these tests were conducted without consideration of grouting materials and
pressure. Further study should be initiated focusing on the influence of various mate-
rials, grouting pressure as well as proportion of the grouting material to determine
the best way to increase pile capacity.
104
Table 3.5 Static and dynamic load tests of compressive loaded piles
Pile label Technology used Double tangent DeBeer’s Chin’s Dynamic load test
Load (kN) Settl. (mm) Load (kN) Settl. (mm) Load (kN) Settl. (mm) Load (kN) Settl. (mm)
P51 No grouting 7,560 6.0 7,600 7.9 10,000 N/A N/A N/A
P121 Base grouting 7,700 6.2 7,900 8.3 11,111 N/A 9,071 26.37
P126 Base and shaft 8,250 8.0 8,300 8.2 12,500 N/A 9,851.5 18.93
3 Field Tests of Post Grouted Concrete Piles
References 105
Table 3.6 Static and dynamic load tests of uplift loaded piles
Technology used Dynamic load test (kN) Mazurkiewicz’s method (kN)
Ultimate bearing Shaft resistance Ultimate bearing Shaft resistance
No grouting N/A N/A 5,442 5,442
Base grouting 9,071 4,217.3 N/A N/A
Base and shaft 9,851.5 4,850.0 6,299 6,299
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Chapter 4
Field Tests of Precast Concrete Piles
Pile foundation is a long structural element that cooperates with soil to resist the
loading transferred from the upper structure. It can be primarily categorized into
cast-in-place and precast pile based on the piling construction condition. Precast pile
can resist more loading comparing to shallow foundation, and sometimes cost less
since it does not need to pre-drilling as the bored pile does. Mostly, the precast pile
foundation is considered as displacement pile because the precast pile will compact
soil layers during driving. Based on the configuration, it can be categorized into small
or large displacement piles.
For the precast concrete square piles (large displacement piles), these sizes can be
ranging from 200 to 700 mm in diameter and 12 to 25 m in length, and the working
loads that can be resisted vary from 200 to 1,200 kN. These piles can be a normally
reinforced structure or be a prestressed structure. Past studies were focused on the pile
tests illustration (Mansur and Hunter 1970), tests in different soil profiles (Balasub-
ramaniam et al. 2009; Gregersen et al. 1975), drivability (Ashford and Jakrapiyanun
2001; Fellenius and Samson 1976; Hussein et al. 2006), load transfer mechanism
(Fellenius 2002; Hsu 2014) and configurations of piles and strength of concrete (Liew
et al. 2004). Recently, there has been intensive investigations aimed to research the
materials effect on concrete pile. For example, full-size tests were performed on
concrete piles with Illinois PCC bottom ash, and these piles were compared to the
traditional reinforced concrete pile with fly ash admixture (Kumar et al. 2004).
The behavior of open-ended piles is more complicated because the Soil Plug Effect
(SPE) which created inside the pile should be considered when investigating the open-
ended piles. Recently, a new CPT-based HKU method was proposed for base capacity
estimation of the open-ended pipe piles with mechanical consideration of annulus
resistance and plug resistance (Yu and Yang 2012). Detailed reviews of these precast
piles were provided with consideration of soil profiles, methods of driven technique
and configurations of precast piles (Marcos et al. 2013). Furthermore, another aspect

https://doi.org/10.1007/978-981-33-6183-6_4
108 4 Field Tests of Precast Concrete Piles
that associated with precast pile in clay is the ‘setup’ or ‘freeze’ of the foundation,
which is a normal phenomenon of pile capacity increases with the increment of time.
Recent research has commenced to determine the amount of setup time of precast
piles by performing dynamic load tests, stanamic load tests and static load tests
(Gwizdala and Wieclawski 2013; Abu-Farsakh et al. 2017).
Because of the process of driving pile into soil layer is relatively similar to some
tests like standard penetration test, dynamic load tests etc., numerous studies have
been focused on the relationships between bearing capacity with penetration tests’
parameters (Lehane et al. 2013; Zhusupbekov et al. 2014). The real capacity is
literally accepted after the field tests being conducted and hence there are a lot
of field tests research performed. Practically, if static load tests are selected to be
conducted, the Proof Test (PT) is preferred because it only apply twice the allowable
load instead of failing it. Then, there are many extrapolation methods proposed
to determine the ultimate bearing pile capacity. Marcos et al. (2013) reviewed 8
interpretation criteria based on the database of 72 sites with 152 field compression
load tests. These tested piles included round and square shape, and the soil profile was
categorized into drained and undrained condition. It is concluded the Davisson and
slope tangent methods (or double tangent method) give a lower interpreted failure
load by 10–15%, and Chin’s method gives the upper bound solution.
The utilization of bored piles or driven piles is accepted worldwide in deep excava-
tion engineering. Different to a Soil Mixing Wall (SMW) in which the inside H-steel
pile can be pulled out after the underground construction is finished, the driven pile
or cast-in-situ pile can be used as part of the building. The use of these passive piles
as a retaining structure can improve the slope stability and hence allow much deeper
excavation to be performed. These piles protect the buildings including underground
structures near the construction site, and to some extent, they can also resist water
permeation.
Compared to driven piles which need to be transferred from the manufacturing
factory, cast-in-situ piles are used more commonly in excavation construction as
these bored piles are cast once the holes are drilled and soils are removed, which
improves the construction process. However, the disadvantage of this time-saving
construction is that the welding of the reinforcement cage and blending the concrete at
the construction site can cause severe environmental issues. Since 1995, the discharge
of carbon dioxide in China has been 5.5 times higher than in America and 13.8 times
higher than in Japan (Ma and Li 2006). It has been reported in the Chinese news that
the ejection of industrial dust increased to 11.75 million tons in 2011 (Environtology
2011a) and that construction waste accounted for 30–40% of total municipal solid
waste in 2011 (Environtology 2011b). This environmental problem can be controlled
when substituting bored piles with driven piles. However, because of the proximity
between passive piles in excavation construction and their immense size, driving
these piles can squeeze the soils, consequently leading to a huge soil compaction
effect. The best way to handle this issue is to excavate the soil first and then lower the
driven pile into the drilled hole; it is worth noting that this method also diminishes
noise pollution. The diameter of this hole cannot be larger than the diameter of the
pile because once there is a gap between the concrete surface of the pile and soil,
there will be no friction resistance which will result in decreased pile capacity. The
construction method that removes an amount of soil first and later lowers the driven
pile raises questions concerning the mechanism of friction resistance of the pile.
There is much research related to static load tests for determining the bearing
capacity of the pile. By comparing grouted bored piles with traditional slurry stabi-
lized bored piles under static load tests, a significant increase in skin friction resistance
has been observed (Chu 2004). Most researchers have considered vertical compres-
sive static load tests and interpretation of the test results for pile capacity determi-
nation; others have investigated the load transfer characteristic of piles. However,
studies of the load transfer mechanism of the non-displacement precast pile under
vertical uplift loads are very limited.
With technological developments, the statnamic method and dynamic methods
which contain CASE and CAPWAP approach, are also used for capacity deter-
mination under vertical loading and are replacing the conventional static load test
progressively. For the determination of ultimate lateral bearing capacity of a non-
displacement driven pile which is used as a retaining structure, a lateral static load
test still needs to be conducted. Different to the vertical static load test, the regular
lateral load tests are conducted under five cyclic loads. In detail, after the first hori-
zontal loads are applied, the loads will be released and later, the pile will be lateral
loaded and released again. This process will be repeated five times, and then the
second loading stage (increase the loads) can be performed.
This chapter examines the behaviors of small, large and non-displacement piles
through performing compressive, uplift and lateral SLTs. Two case studies are
provided. The first case study researched a total number of 40 piles tested by
compressive SLTs to evaluate the pile behavior. The pile foundations included 25
pipe concrete piles (small displacement precast piles) and 15 rectangular concrete
piles (large displacement precast piles). The first case study focused on researching
piles’ behavior under compressive loads. The piles lengths varied, and all reached
the angular gravel which was the bearing stratum. It should be noted that the
capacity comparison between interpretations’ outcome as well as the designed
bearing capacity, and the example with detailed calculation steps are provided in
Chapter 8. The second case study consisted of three piles tested by uplift loading and
three piles with horizontal loading (non-displacement precast concrete piles). This
second study aimed to determine the pile behavior under uplift and lateral loads and
the load transfer mechanism. Further, it included interpretation of non-displacement
precast piles under uplift and lateral loads.
4.2 Small and Large Displacement Concrete Piles
4.2.1 Project Introduction
The first examined construction project aimed to construct a 28-level apartment with
an underground garage. The driven location of the tested piles was in Shandong
Province, China. The precast piles were driven in the construction site, which is
shown in Fig. 4.1. Generally, the piles from three areas were tested by the compressive
SLTs. Areas A and B were the locations of the tested concrete pipe piles, and Area
C was the location of the tested solid rectangular piles. The solid rectangular pile
driven in Area C was used to resist the loads transferred from the upper building.
4.2.2 Subsurface Conditions
The subsurface exploration was achieved through laboratory and in-situ tests. The
in-situ SPTs and CPTs and laboratory tests of consolidation, direct shear and triaxial
tests were conducted based on the Chinese Standard for Soil Test Method (GB/T
50123-1999 1999b), Standard for Test Methods of Engineering Rock Mass (GB/T
50266-2013 2013b) and Code for Investigation of Geotechnical Engineering (GB
50021-2001 2001), respectively. Based on the borehole logs, the soil layers were
discovered as follows.
1. planting soil (Q4ml ): yellowish brown, loose and wet, contains clay, partially
contains plant roots, average thickness of 0.75 m; a. miscellaneous fill (Q4ml ): loose,
partially contains gravel and rubble, diameter ranging from 2.0 to 7.0 mm, average
Fig. 4.1 Location of tested piles (not to scale)

4.2 Small and Large Displacement Concrete Piles 111
thickness of 1.22 m; 2. silty clay (Q4al+pl ): medium stiff, yellowish with low plas-
ticity, average thickness of 3.91 m; 3. silty clay (Q4m ): grey to ash black, liquid-
plastic state, contains fine sand, partially contains clayey silt, average thickness of
4.92 m; a. fine sand (Q4m ): grey and cinereous, loose, saturated, average thickness
of 2.01 m; 4. silty clay (Q4al+pl ): grey to yellowish grey, medium stiff, low plas-
ticity, average depth of 2.57 m; 5. silty clay (Q4al+pl ): yellowish brown, plastic state,
partially contains medium to dense sand, average thickness of 2.57 m; a. medium
sand (Q4al+pl ): yellowish brown, medium dense, loose and saturated, average thick-
ness of 4.08 m; b. coarse gravel sand (Q4al+pl ): yellowish brown, medium to dense,
saturated, average thickness of 4.19 m; 6. angular gravel (Q4p1+d1 ): yellowish brown,
medium dense, saturated, particle size ranges from 2 to 50 mm, maximum to 100 mm,
highly weathered, average thickness of 5.66 mm; 7. highly weathered mica schist
(Pt ): yellowish brown to grey, highly weathered, thickness of 4.52 m; 8. medium
weathered mica schist (Pt ): grey, medium hard.
Fig. 4.2 Driven small displacement pipe pile

4.2.3 Pile Preparation
For precast open-ended pipe piles, the diameter was 400 mm and the length varied
from 20 to 26 m. In Area A, one 20 m pile, seven 21 m piles, seven 22 m piles, two
23 m piles and five 24 m piles were tested. In Area B, one 23 m pile and two 26 m
piles were tested. For precast solid rectangular piles, the cross-section was 400 ×
400 mm2 . All rectangular piles were tested in Area C. The tested piles consisted
of two 22 m piles, two 25 m piles, one 26 m pile, four 27 m piles and six 28 m piles.
The concrete strength of these tested piles was 30, 50 and 40 MPa from Areas A, B
and C, respectively.
4.2.4 Test Setup
Different to the case study illustrated in Chap. 3, all the pile foundation was driven
underground (Fig. 4.2), and the weighted platform was selected for SLT instead of
the reaction beams and piles, as demonstrated in Fig. 4.3. As a result of the designed
capacity of these piles being relatively small, compared with the bored piles depicted
in Chap. 3, the reaction system only needed to provide small reaction loads. Further,
Fig. 4.3 Weighted platform of SLT

4.2 Small and Large Displacement Concrete Piles 113
as numerous piles needed testing, it would have been very expensive to cast reaction
piles. This weighted platform is more practical to select when a large number of tests
is required because the transfer of the platform is much cheaper.
To perform compressive SLTs, two hydraulic jacks were used to provide loads.
Four dial gauges were symmetrically installed on a reference bar to measure the
vertical displacement of the pile head. For Area A, maintained loads with an incre-
ment of 130 kN were applied on the precast pipe pile until a maximum load of 1,300
kN being applied. For the pipe pile in Area B, an increment of 314 kN was applied
until a maximum load of 3,140 kN. For the rectangular concrete pile, maintained
loads with increments of 488 kN were applied on most piles (to maximum load of
4,880 kN), except for three piles that had increment loads of 360 kN (to maximum
load of 3,240 kN) for data comparison. All the information about these tested piles
is summarized in Tables 4.1, 4.2 and 4.3.
4.3 Non-displacement Square Concrete Piles
This project was conducted in western Jinan city, China—coordinates of E116°54 –

E117°02 , N36°35 –N36°40 (Fig. 4.4). It aimed to construct a subway station with
total construction area of 356.6 × 19.7 m2 . The excavation depth was 16.8 m for
this project and this project was based on the following local standards:
GB 50308-2008 (2008a) (Code for Urban Rail Transit Engineering Survey),
GB/T 18314-2009 (2009a) (Specifications for Global Positioning System [GPS]
Surveys), GB 50299-1999 (1999a) (Code for Construction and Acceptance of Metro
Engineering), JGJ 8-2007 (2007) (Code for Deformation Measurement of Building
and Structure), CJJ 76-2012 (2012) (Specification for Dynamic Observation of
Groundwater in Urban Area), JGJ 120-2012 (2012) (Technical Specification for
Retaining and Protection of Building Foundation Excavations), GB 50497-2009
(2009b) (Technical Code for Monitoring of Building Excavation Engineering), GB
50157-2013 (2013a) (Code for Design of Metro), GB 50111-2006 (2006) (Code
for Seismic Design of Railway Engineering), QCR 9218-200-15 (2015) (Code for
Tunneling Monitoring Technology Procedures) and GB 50911-2013 (2013b) (Code
for Monitoring Measurement of Urban Rail Transit Engineering).
Before the projects started, boreholes were driven to depths ranging from 31.0 to
52.0 m to determine the subsurface. In total, 401 SPTs and dynamic penetration
tests were conducted, with the latter reaching the bottom of gravel and medium
sand. In addition, based on the Chinese code GB/T 50123-1999 (1999b), laboratory
114
Table 4.1 Load–settlement results of open-ended pipe pile (1)

No 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15
Label P1172 P1470 P976 P991 P995 P1163 P1165 P1175 P630 P1473 P643 P1509 P1485 P1475 P1580
Length 20 21 21 21 21 21 21 21 22 22 22 22 22 22 22
(m)
Load Settl. Settl. Settl. Settl. Settl. Settl. Settl. Settl. Settl. Settl. Settl. Settl. Settl. Settl. Settl.
(kN) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (mm)
0 0 0 0 0 0 0 0 0 0 0 0 0 0 0 0
260 0.56 0.6 1 0.62 0.67 0.93 0.62 0.55 0.73 0.54 0.61 0.72 0.7 0.75 0.89
390 1.74 1.61 2.43 1.97 1.86 2.23 1.79 1.9 1.66 1.84 1.26 1.64 1.62 1.94 1.87
520 4.04 3.04 4.28 3.74 2.8 3.62 2.97 3.54 2.72 3.24 2.33 3.01 3.41 2.95 3.17
650 5.45 4.51 5.77 5.97 4.5 5.81 4.72 5.78 4.13 4.84 3.47 4.63 5.27 4.34 5.08
780 7.43 6.45 8.32 7.84 5.98 8.19 6.12 8.24 5.78 6.51 4.83 6.77 7.2 5.56 7.29
910 9.54 8.47 11.18 10.58 8.33 11.12 8.24 10.83 7.32 8.68 6.45 8.67 9.21 7.46 10.41
1,040 12 10.9 13.4 13.36 11.91 14.78 10.35 13.21 9.11 11.58 7.92 11.33 11.39 9.4 13.35
1,170 15.14 13.26 17.22 16.47 16.33 18.5 13.67 16.24 11.37 13.99 9.34 14.9 14.63 11.63 16.84
1,300 18.59 15.8 21.8 19.58 20.97 22.64 17.48 19.56 14.08 16.55 11.48 19.19 18.01 14.53 21
1,040 17.13 14.57 20.74 18.29 19.91 21.45 16.48 18.18 13.34 15.19 10.78 17.99 16.44 13.73 19.82
780 15.2 13.24 19.03 16.71 18.34 19.85 14.98 16.67 12.42 13.81 10.06 16.56 14.84 12.5 18.36
520 13.01 11.39 16.85 14.54 16.26 17.69 12.58 14.53 10.91 11.93 8.75 14.88 12.93 10.83 16.47
260 10.81 9.52 14.08 12.15 13.5 15.06 10.37 12.39 8.74 9.75 6.88 12.53 10.89 8.83 14.19
0 8.34 7.19 10.41 9.26 10.53 11.8 8.09 9.66 6.4 7.57 5.36 9.74 8.67 6.66 11.58
R P−S 55.14 54.49 52.25 52.71 49.79 47.88 53.72 50.61 54.55 54.26 53.31 49.24 51.86 54.16 44.86
4 Field Tests of Precast Concrete Piles
Table 4.2 Load–settlement results of open-ended pipe pile (2)
No 16 17 18 19 20 21 22 No 23 24 25
Label P1481 P639 P1275 P1287 P1299 P1453 P1465 Label S73 S103 S126
Length (m) 23 23 24 24 24 24 24 Length (m) 26 26 23
Load (kN) Settl. (mm) Settl. (mm) Settl. (mm) Settl. (mm) Settl. (mm) Settl. (mm) Settl. (mm) Load (kN) Settl. (mm) Settl. (mm) Settl. (mm)
0 0 0 0 0 0 0 0 0 0 0 0
260 0.89 0.66 0.91 0.96 0.7 0.47 0.66 628 0.8 0.91 0.93
390 1.6 1.14 2.06 2.06 1.67 1 1.53 942 1.38 1.56 1.72
520 3.53 2.28 3.5 3.36 3.34 1.84 2.64 1,256 2.71 2.41 3.1
650 5.27 3.11 5.05 5.18 4.85 3.1 3.95 1,570 4.65 3.69 5.01
4.3 Non-displacement Square Concrete Piles
780 7.67 4.51 7.02 7.92 6.63 4.41 5.12 1,884 7.17 5.29 7.38
910 9.6 5.79 9.41 10.72 8.77 6.2 6.92 2,198 9.79 7.48 10.06
1,040 12.7 7.56 11.8 14.3 10.93 7.84 8.8 2,512 13.15 9.9 13.55
1,170 15.6 9.26 14.58 18.66 13.22 10.46 11.13 2,826 16.58 13.09 17.65
1,300 20.1 12.08 18.2 23.58 15.63 13.2 13.73 3,140 20.95 16.6 22.34
1,040 18.87 11.11 16.99 22.33 14.89 12.32 12.77 2,512 19.99 15.44 21.17
780 17.54 9.87 15.53 20.51 13.43 11.03 11.48 1,884 18.16 14.03 19.54
520 15.47 9.07 13.86 18.12 11.79 9.37 10.13 1,256 15.87 12.25 17.22
260 13.47 7.59 11.51 15.14 9.4 7.57 8.2 628 13.37 9.93 14.02
0 10.73 5.51 8.69 11.56 7.08 5.85 6.07 0 10.05 6.98 10.07
R P−S 46.62 54.39 52.25 50.98 54.70 55.68 55.79 R P−S 52.02864 57.95181 54.9239
115
116
Table 4.3 Load–settlement results of solid rectangular concrete pile

No 1 2 3 4 5 6 7 8 9 10 11 12 No 13 14 15
Label P1 P91 P238 P124 P137 P173 P147 P93 P85 P48 P16 P4 Label P14 P41 P106
Length 25 26 27 27 27 27 28 28 28 28 28 28 Length 22 22 25
(m) (m)
Load Settl. Settl. Settl. Settl. Settl. Settl. Settl. Settl. Settl. Settl. Settl. Settl. Load Settl. Settl. Settl.
(kN) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (mm) (kN) (mm) (mm) (mm)
0 0 0 0 0 0 0 0 0 0 0 0 0 0 0 0 0
976 0.77 0.89 0.72 0.56 1.29 0.78 0.75 0.71 0.96 0.67 0.94 0.72 720 0.7 0.81 0.8
1,464 1.59 1.89 1.58 1.2 2.56 2.33 1.97 1.35 2.09 2.15 1.7 2.27 1,080 1.59 1.63 1.89
1,952 3.09 3.34 2.76 2.4 4.35 4.05 3.82 2.6 3.95 3.41 3.39 4.5 1,440 2.46 2.59 3.3
2,440 5.34 5.22 4.27 4.36 7.28 6.96 6.39 4.3 5.72 6.09 5.09 6.75 1,800 3.74 3.73 5.02
2,928 7.25 7.4 6.06 6.67 10.05 9.59 9.77 6.61 8.09 8.46 7.11 9.9 2,160 5.08 5.24 6.93
3,416 10.03 9.66 8.42 8.92 13.54 12.74 13.98 9.1 10.42 11.22 9.35 12.8 2,520 6.57 7.31 9.19
3,904 13.59 13.06 11.19 12.23 17.78 16.19 17.85 12.22 13.92 14.11 12.4 16.36 2,880 9.18 10.14 11.83
4,392 17.11 16.57 14.61 15.93 22.35 19.79 22.17 15.91 18.17 17.13 15.59 20.13 3,240 11.85 13.63 14.97
4,880 21.53 21.72 18.77 21.27 27.3 24.5 26.35 20.27 23.82 21.82 19.35 24.5 3,600 15.24 17.77 18.58
3,904 20.35 20.45 17.63 20.46 25.99 23.33 25.16 19.08 22.2 20.91 18.19 23.23 2,880 14.48 16.8 17.54
2,928 18.57 18.88 16.26 18.9 24.07 21.46 23.27 17.57 20.39 19.72 16.82 21.34 2,160 13.59 15.46 16.11
1,952 16.32 16.78 14.28 16.72 21.25 19.07 20.49 15.61 18.2 18.03 14.64 18.86 1,440 12.06 13.83 14.33
976 13.72 13.35 11.98 14.31 17.62 16.36 17.26 13.12 15.51 15.92 12.45 16.08 720 10.41 11.72 12.12
0 10.54 9.91 9.27 11.03 13.45 12.45 13.43 10.35 13.43 12.75 9.51 12.52 0 7.77 8.89 9.46
R P−S 51.05 54.37 50.61 48.14 50.73 49.18 49.03 48.94 43.62 41.57 50.85 48.90 R P−S 49.02 49.97 49.09
4 Field Tests of Precast Concrete Piles
4.3 Non-displacement Square Concrete Piles 117
Fig. 4.4 Location of general engineering construction site (not to scale)
tests were conducted to determine the soil characteristics of specific gravity, relative
density, porosity, void ratio, saturation ratio, water content, plastic limit, liquid limit,
plasticity index, liquidity index, cohesion and friction angle (direct shear test and
triaxle tests [CU and UU]). Through interpreting the borehole logs and parameters
obtained from laboratory tests, the subsurface investigation found the following:
1. miscellaneous fill: contains gravel, concrete and brick stone, construction waste
and plant roots, average depth of 2.5 m; 2. yellowish brown loess: plastic iron-
manganese concretion, average depth of 2.9 m; 3. tawny silt: slightly wet, low
tenacity, average depth of 2 m; 4. silty clay: yellow and brown, medium to high
plasticity, contains iron and manganese oxides, average depth of 11.6 m; 5. fine
sand: brown, medium dense and wet, contains quartz arkose, average depth of 2.1 m;
6. dark brown silty clay: contains iron and manganese oxides, average depth of
10 m; 7. silty clay: yellow, light brown, iron-manganese and calcareous concretion,
contains little sand and gravel, average depth of 8.3 m; 8. coarse sand: brown, dense,
saturated, average depth of 3.6 m; 9. Clay: brown, high viscous, average depth of
2 m. The soil parameters of these subsurface layers are summarized in Table 4.4.
Table 4.4 Soil parameters of subsurface layers

Soil layer ω Gs ρ Es e IP IL c ϕ N qsik qpk
% – g/cm3 MPa – – – kPa o – – –
Fill N/A N/A 1.8 N/A N/A N/A N/A N/A N/A N/A N/A N/A
Loess 24.5 2.7 1.93 9.3 0.75 11.8 0.55 29 24 10 30 N/A
Silt 23.2 2.69 1.96 15.8 0.69 8.7 0.51 19 20 15 35 N/A
Tawny clay 26.4 2.72 1.95 9.1 0.77 12.8 0.53 47 22 22 45 N/A
Fine sand N/A N/A 1.96 N/A N/A N/A N/A N/A N/A 18 50 N/A
Silty clay 27.3 2.72 1.93 7.2 0.8 13.8 0.47 51 23 29 60 N/A
Silty clay 26.5 2.72 1.93 6.8 0.780 13.2 0.46 44 18 25 60 2,200
Coarse sand N/A N/A 1.98 20 N/A N/A N/A N/A N/A 39 60 6,800
Clay 26.8 2.74 1.91 7 0.766 19.2 0.27 39 17 N/A 62 2,400
Table 4.5 Strain gauge

Depth (m) Label of strain Initial value (Hz) K-value
information
gauge
2.3 285,349 1,442.9 0.00006695
6.9 287,918 1,431.0 0.00006768
11 28,588 1,410.5 0.00006423
16 281,245 1,428.1 0.00007353
21 286,639 1,415.2 0.00006673
27 286,556 1,452.2 0.00006921
4.3.3 Pile Descriptions
All six precast piles contained two parts with bolt connections, as shown in Fig. 4.5.
The cross-sections of these tested piles were all rectangular, with side lengths of
700 mm. The three lateral loaded piles and three uplift loaded piles were labelled
P15, P163 and P152 and P138, P151 and P165, respectively, with the same concrete
strength of C50 (characteristic strength of 50 MPa, cubic test sample dimension
of 150 × 150 × 150 mm3 ), and the same horizontal reinforcement of HRB400
Fig. 4.5 Tested

non-displacement precast
pile
(hot-rolled ribbed bar with yield strength of 400 MPa) and vertical reinforcement of
HPB300 (hot-rolled plane bar with yield strength of 300 MPa).
The lengths of lateral loaded non-displacement piles P15, P163 and P152 were
24.7, 26.6 and 28 m, respectively. For the uplift loaded concrete piles P138, P151
and P165, the lengths were 24.7, 26.6 and 28 m, respectively. Specifically, six
pairs of strain gauges were installed in P165 to determine the uplift load transfer
characteristics of this non-displacement precast pile.
The locations, initial value and K-value of the strain gauges are summarized in
Table 4.5. The stress of the strain gauge can be calculated by Eq. 4.1, using the
parameters from Table 4.5 and the collected frequency data recorded onsite. The
axial loading of the pile cross-section can then be determined through Eq. 4.2:
1 2
n
−
σs = K i f ini − f i 2 /As (4.1)
n i=1

− Ec
N P =σs Ac + As (4.2)
Es
where:
−
σs = average stress of vibration wire strain bar.
NP = force of the concrete pile.
Ec = Young’s modulus of concrete.
Es = Young’s modulus of steel.
Ac = cross-section area of concrete.
As = cross-section area of steel.
Ki = coefficient of vibration wire stress rebar.
f ini = initial recorded frequency.
fi = recorded frequency of vibration wire stress rebar.
4.3.4 Installation Process of Piles
These piles were cast in the factory, which can control the contamination caused
by cement dust and exhaust gas released during welding, as shown in Fig. 4.6. The
construction process of installing the non-displacement precast concrete piles was
as follows:
1. The soil was removed from the required location with an auger drilling machine,
as shown in Fig. 4.7. The drilled hole was 550 mm, which was less than the side length
of the pile cross-section, so there was no gap between the pile and soil (Fig. 4.8);
2. After the first piece of the precast concrete pile being transferred to the location
of the drilled hole (Fig. 4.9), the pile could automatically descend by gravity; 3. A
squeezing machine clamped the first piece of the concrete pile, so it could be bolted
with the second piece of pile, as shown in Fig. 4.10.
Fig. 4.6 Precast piles in factory
4.3.5 Test Setup
As shown in Figs. 4.11 and 4.12, two reaction beams were used for lateral and uplift
SLTs. One hydraulic cylinder and two hydraulic jacks were used to provide loads for
the lateral and uplift tests, respectively. Two dial gauges were located at the pile side
for recording the horizontal movement of the pile head, and four dial gauges were
symmetrically installed on the reference bar to measure the vertical uplift movement
of the pile head.
Based on the Chinese codes JGJ 106-2014 (2014), JTJ 041-2000 (2000) and JGJ
94-2008 (2008b), horizontal cycling load tests were conducted. After the maintained
loads were applied, a duration of four minutes was required before recording the
horizontal movement. And after releasing the loads back to 0 kN, a duration of two
minutes was required before recording the residual displacement of pile. During this
test, five load circulations were required before the next loading was applied. The
tests should be ceased if:
1. The monitored horizontal movement of the pile head exceeded 30 mm, or
2. A concrete crack emerged.
For the uplift SLTs, low-speed maintenance tests were conducted. The applied
load was maintained until the rate of axial movement did not exceed 0.1 mm. After
Fig. 4.7 Auger drilling machine
each maintained load was applied, the vertical movement of the piles was recorded
with intervals of five, 10 and 15 min, and 30 min. If the accumulated time exceeded
one hour, record movement of pile with interval of 30 min. The loading could be
terminated in the case of the following conditions:
1. The applied loading was equal to the value of 0.9 times the ultimate strength of
reinforcement
2. Quintuple movement change of the previous loading was discovered
3. The pile head movement increased up to 100 mm.
4.4 Static Load Test Results of Precast Concrete Piles
4.4.1 Results of Small Displacement Precast Piles
As shown in Fig. 4.13, the load–settlement curves of the 21 m pipe pile showed
similar behavior. The total settlement of the 21 m piles was found to be an average of
19 mm, and, after the loading released, the average permanent settlement was found
to be 10 mm. It was also found that the curves after loading released were parallel
with each other.
Fig. 4.8 Drilled hole for precast pile
The load–settlement curves of 22 m piles are presented in Fig. 4.14. It was found
that the total settlement of the 22 m piles was an average of 15 mm, and, after the
loading released, the average settlement was found to be 8 mm. Similar to the 21 m
piles, the curves during unloading between the tested piles were parallel.
The load–settlement curves of the 24 m piles are provided in Fig. 4.15. The
behavior of P1287 was relatively different to the other piles, which requires further
discussion. The other 24 m piles showed similar behavior, and the total settlement
was found to be 15 mm. After the loading was released, the average settlement was
found to be 7 mm. The load–settlement curves of piles with different lengths are
provided in Fig. 4.16. It was found that the total average settlement was 21 mm, and
the permanent average displacement was 10 mm. The curves of the tested piles during
the unloading stages were parallel with each other. The load–settlement curve of the
tested pile located at Area B is provided in Fig. 4.17, the total average settlement
was found to be 18 mm, and the average permanent settlement was found to be
8 mm. Similar to the pile behaviors tested at Area A, the curves of these tested piles
during unloading were also found to be parallel, but the gradient (this is defined as
the settlement change per load from the load release period) was different to the piles
tested in Area A.
4.4 Static Load Test Results of Precast Concrete Piles 123
Fig. 4.9 Transferring precast piles
4.4.2 Results of Large Displacement Precast Piles
The load–settlement curves of the 27 m rectangular piles are presented in Fig. 4.18.
The total average settlement was found to be 23 mm, and the average permanent
settlement was found to be 12 mm. For the 28 m piles shown in Fig. 4.19, the load–
settlement curves illustrated that the total average settlement was 22 mm, and the
average permanent settlement was 12 mm.
As shown in Fig. 4.20, two piles with the same dimensions and soil condition
(distance between piles of 5 m) were tested with a different load increment, and the
same behavior was identified. Figure 4.21 presents the rectangular piles’ behavior
with various pile lengths. The results indicated that the total average settlement was
24 mm, and the average permanent settlement was 12 mm. Similar to the results
obtained from the pipe piles, it was found that, during unloading stages, these curves
were relatively parallel and the gradients were close to each other.
Fig. 4.10 After descending the first piece of pile
4.4.3 Results of Non-displacement Precast Piles
The horizontal loads, H, and movement, X, of three tested precast piles were recorded
as shown in Figs. 4.22, 4.23 and 4.24. It can be seen that the critical load (H cr ) of P15
was 100 kN (to be conservative). The ultimate load appeared to be 400 kN; however,
there were some cracks observed after applying 400 kN. Thus, the ultimate load was
determined to be 300 kN. For P163, the critical load was 200 kN, while the ultimate
load (H u ) was difficult to read and interpretation was required. The H cr and H u of
P152 were determined to be 240 and 480 kN, respectively.
The interpretation of the horizontal SLTs is provided in Figs. 4.25, 4.26 and 4.27.
Through establishing a coordinate system of horizontal loading versus nominated
gradient settlement, three lines were discovered in each diagram. Each two intersec-
tions between two lines demonstrate the corresponding H cr and H u . The equations
of each line are also illustrated in Figs. 4.25, 4.26 and 4.27. The H cr and H u of P15,
P163 and P152 were 107.5, 154.4 and 175.4 kN (H cr ) and 280.9, 447.7 and 470.8
kN (H u ), respectively.
Fig. 4.11 Lateral SLT setup
The results of the H–X curves and interpretation of these SLTs are summarized in
Table 4.6. The results were close to each other and all demonstrated that the critical
and ultimate loads of these non-placement precast piles increased with pile length. It
is worth noting that it is sometimes difficult to determine the critical and ultimate load
through H–X curves; thus, comprehensive analysis between SLTs and interpretation
is required for geotechnical engineers.
The load–movement curve results of three uplift loaded piles are provided in
Fig. 4.28. The maximum uplift movements of these piles’ heads were discovered
to be 1.73 mm (P138), 1.6 mm (P151) and 1.31 mm (P165) when the maximum
loading of 1,900 kN was applied. The corresponding residual uplift displacements
of these three piles were 0.47, 0.5 and 0.35 mm, respectively. As shown in Fig. 4.28,
all three lines from the loading stages were approximately linear; thus, the maximum
loading of 1,900 kN was not the ultimate bearing capacity of these three piles. As
such, interpretation of this diagram was required.
Fig. 4.12 Uplift SLT setup

0 200 400 600 800 1000 1200 1400
0
5
Settlement (mm)
10
P1470
P976
15
P991
P995
20 P1163
P1165
P1175
25
Fig. 4.13 Load–settlement results of 21 m pipe pile


0 200 400 600 800 1000 1200 1400
0
5
Settlement (mm)
10
P630
P1473
15
P643
P1509
20 P1485
P1475
P1580
25
Fig. 4.14 Load–settlement results of 22 m pipe pile
0 200 400 600 800 1000 1200 1400

0
5
Settlement (mm)
10
15
P1275
P1287
20 P1299
P1453
P1465
25
Fig. 4.15 Load–settlement results of 24 m pipe piles
The ultimate uplift bearing capacity of non-failure uplift SLTs can be determined
with the modified Mazurkiewicz method. The assumption of this method is that the
nominate movement (load/displacement) value is equal to 0 when the loading is
small and the displacement is very high, the dash line would be go through by y-axis
which illustrates the uplift ultimate bearing capacity of pile.

0 500 1000 1500
0
5
Settlement (mm)
10
15
P1287-24
20 P1172-20
P995-21
P1509-22
25
Fig. 4.16 Pipe piles at Area A with different lengths

0 1000 2000 3000 4000
0
5
Settlement (mm)
10
15
20 S73-26
S103-26
S126-23
25
Fig. 4.17 Pipe piles at Area B with different lengths
Based on Fig. 4.29, the equations representing these three lines are provided in
Eqs. 4.3, 4.4 and 4.5. When nominate movement is equal to 0, the ultimate uplift
bearing capacities of P138, P151 and P165 were determined as 2,903.6, 3,467.0 and
4,724.2 kN, respectively. As summarized in Table 4.7, the uplift bearing capacity of
these non-displacement piles increased with the pile length:

0 1000 2000 3000 4000 5000 6000
0
10
Settlement (mm)
15
20
P238
P124
25
P137
P173
30
Fig. 4.18 Load–settlement results of 27 m rectangular piles

0 1000 2000 3000 4000 5000 6000
0
5
Settlement (mm)
10
15
P147
20 P93
P85
P48
25
P16
P4
30
Fig. 4.19 Load–settlement results of 28 m rectangular piles
P138: y = −0.9253x + 2903.6 (4.3)
P151 : y = −1.318x + 3467.0 (4.4)
P165: y = −1.9371x + 4724.2 (4.5)
The vibration frequency obtained from strain gauges in P165 was recorded at each
loading stage. Using Eqs. 4.1 and 4.2, the average axial uplift loads were determined

0 1000 2000 3000 4000 5000 6000
0
5
Settlement (mm)
10
15
20
P106
P1
25
Fig. 4.20 Same pipe with different load increments

0 1000 2000 3000 4000 5000 6000
0
10
Settlement (mm)
15
P14-22
20
P1-25
P91-26
25
P137-27
P16-28
30
Fig. 4.21 Rectangular piles at Area C with different lengths
along the pile. As shown in Fig. 4.30, the axial force decreased with depth along the
pile and increased with applied loads provided by hydraulic jacks. This was caused
by friction resistance provided by the soil layers. The accumulated friction resistance
along the pile is provided in Fig. 4.31.
The distribution of unit shaft resistance was obtained by calculating the differences
between the two adjacent axial loads acquired from the strain gauges. As shown in
Number of Cycle Tests

0 5 10 15 20 25 30
0
Hcr
2
4
Concrete Cracking
6 Hu
100kN
200kN
8 300kN
400kN
10 500kN
Loading
Release Loading
12
Fig. 4.22 H–X curves of P15

0 5 10 15 20 25 30 35 40
0
Hcr
5
Horizonntal Movement (mm)
10
100kN
15 200kN
300kN
400kN
20 500kN
600kN
25 700kN
Loading
Release Loading
30
Fig. 4.32, the shaft resistance along the pile shaft is provided. As shown in this figure,
when the uplift loads increases, the friction loads of each soil layer increases. It also
shows that the soil layer from 7 to 16 m plays the most significant role in resisting the
uplift force. Furthermore, it indicates that before the loading stage of 1,330 kN was
applied, the friction resistance of soil layer from 7 to 11 m showed a rapid increasing
trend, and after this stage, the friction resistance of the soil layer from 21 to 27 m
showed a rapid growth.

0 5 10 15 20 25 30 35 40
Horizontal Movement (mm) 0
5
Hcr
10
15 160kN
240kN
Hu
320kN
20
400kN
480kN
25
560kN
620kN
30
Loading
Release Loading
35
0.1
Displacement Gradient (mm/kN)
0.09 y = 0.0003x - 0.0661

0.08 R² = 0.9572
0.07
0.06
0.05
0.04
y = 8E-05x - 0.0043
0.03
R² = 1
0.02 y = 4E-05x
R² = 1 H Hu
0.01 cr
0
0 100 200 300 400 500 600
Horizontal Loading (kN)
Fig. 4.25 Horizontal SLT interpretation of P15
This chapter has presented the investigation of precast concrete piles. Three types of
piles—small, large and non-displacement piles—were examined through two case
studies. For each case study, the background, geotechnical conditions, pile informa-
tion and test setups were illustrated, and the load settlement results were presented
in the final section. For the small and large displacement piles, this chapter focused
on the pile behavior under compressive loading, considering various lengths. For
0.25

y = 0.0006x - 0.2069
0.2
R² = 0.9926
0.15
0.1
y = 0.0002x - 0.0278
R² = 0.9807
0.05 y = 2E-05x
Hcr Hu
R² = 1
0
0 200 400 600 800
0.35
0.3 y = 0.0011x - 0.3632

R² = 0.9999
0.25
0.2
0.15 y = 0.0005x - 0.0807

R² = 0.9817
0.1 Hu
y = 4E-05x - 1E-18
0.05
R² =H1cr
0
0 100 200 300 400 500 600 700
Table 4.6 Summary of capacity and interpretation of horizontal loaded piles

Pile label Pile length (m) Lateral SLT results Interpretation of SLTs
Critical load Ultimate load Critical load Ultimate load
(kN) (kN) (kN) (kN)
P15 24.7 100 300 107.5 280.9
P163 26.6 200 N/A 154.4 447.75
P152 28.0 240 480 175.43 470.8
1.8
1.6
Uplift Movement (mm)
1.4
1.2
0.8
0.6
0.4 P138
0.2 P151
P165
0
0 500 1000 1500 2000
Uplift Static Loading (kN)
Fig. 4.28 Load–displacement results of uplift load piles
5000
4500
4000
3500
Load (kN)
3000
2500
2000 P138
P151
1500
P165
1000 P138
500 p151
P165
0
0 1000 2000 3000 4000
Fig. 4.29 Modified Mazurkiewicz method of uplift loaded piles
Table 4.7 Summary of

Pile label Pile length Uplift SLT results Interpretation of
capacity and interpretation of
(m) SLTs
uplift loaded piles
Ultimate load (kN) Ultimate load
(kN)
P138 24.7 N/A 2,903.6
P151 26.6 N/A 3,467.0
P165 28.0 N/A 4,724.2
Uplift Static Load Test

0 1000 2000 3000
0
10
Depth (m)
15
380kN
570kN
760kN
20 950kN
1140kN
1330kN
25 1520kN
1710kN
1900kN
30
Fig. 4.30 Load transfer mechanism of P156
Accumulated Friction Resistance (kN)

0 1000 2000 3000
0
380kN
570kN
760kN
5
950kN
1140kN
1330kN
10
1520kN
1710kN
Depth (m)
1900kN
15
20
25
30
Fig. 4.31 Accumulated shaft resistance

Unit Shaft Resistance (kN)

0 100 200 300 400 500 600 700 800 900 1000
0
10
Depth (m)
15
380kN
570kN
760kN
20
950kN
1140kN
1330kN
25
1520kN
1710kN
1900kN
30
Fig. 4.32 Unit shaft resistance of P156
the non-displacement pile, uplift and lateral loading conditions were considered.
Further, the interpretation of uplift and lateral SLTs was provided. Finally, this chapter
researched and demonstrated the load transfer mechanism of the non-displacement
pile under uplift loading.
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Ma, S. C., & Li, G. Z. (2006). The relationship between economic increase of China and
environmental contamination based on Kuznets curves analysis. Statistical Research, 8, 37–40.
Marcos, M. C. M., Chen, Y.-J., & Kulhawy, F. H. (2013). Evaluation of compression load test inter-
pretation criteria for driven precast concrete pile capacity. KSCE Journal of Civil Engineering,
17(5), 1008–1022.
Ministry of Construction of the People’s Republic of China. (1999a). Code for construction and
acceptance of metro engineering (GB 50299-1999). Beijing, China: Ministry of Construction of
People’s Republic of China.
Ministry of Construction of the People’s Republic of China. (1999b). Standard for soil test method
Ministry of Construction of the People’s Republic of China. (2000). Technical specifications for
construction of highway bridges and culverts (JTJ 041-2000). Beijing, China: National Standard
of the People’s Republic of China.
Republic of China.
Ministry of Construction of the People’s Republic of China. (2006). Code for seismic design of
railway engineering (GB 50111-2006). Beijing, China: Ministry of Construction of People’s
Republic of China.
Ministry of Construction of the People’s Republic of China. (2007). Code for deformation measure-
ment of building and structure (JGJ 8-2007). Beijing, China: National Standard of the People’s
Republic of China.
Ministry of Construction of the People’s Republic of China. (2008a). Code for urban rail transit
engineering survey (GB 50308-2008). Beijing, China: Ministry of Construction of People’s
Republic of China.
Ministry of Construction of the People’s Republic of China. (2008b). Technical code for building
pile foundations (JGJ 94-2008). Beijing, China: National Standard of the People’s Republic of
China.
Ministry of Construction of the People’s Republic of China. (2009a). Specifications for global
positioning system (GPS) surveys (GB/T 18314-2009). Beijing, China: Ministry of Construction
of People’s Republic of China.
Ministry of Construction of the People’s Republic of China. (2009b). Technical code for monitoring
of building excavation engineering (GB 50497-2009). Beijing, China: Ministry of Construction
Ministry of Construction of the People’s Republic of China. (2012). Specification for Dynamic
Observation of Groundwater in Urban Area (CJJ 76-2012). Beijing, China: National Standard
of the People’s Republic of China.
Ministry of Construction of the People’s Republic of China. (2012). Technical specification for
retaining and protection of building foundation excavations (JGJ 120-2012). Beijing, China:
National Standard of the People’s Republic of China.
Ministry of Construction of the People’s Republic of China. (2013a). Code for design of metro (GB
50157-2013). Beijing, China: Ministry of Construction of People’s Republic of China.
Ministry of Construction of the People’s Republic of China. (2013b). Code for monitoring measure-
ment of urban rail transit engineering (GB 50911-2013). Beijing, China: Ministry of Construction
Ministry of Construction of the People’s Republic of China. (2013c). Standard for test methods of
engineering rock mass (GB/T 50266-2013). Beijing, China: Ministry of Construction of People’s
Republic of China.
Republic of China.
Ministry of construction of the People’s Republic of China. (2015). Code for Tunneling Monitoring
Technology Procedures (QCR 9218–200-15). Beijing, China: National Standard of the People’s
Republic of China.
Yu, F., & Yang, J. (2012). Base capacity of open-ended steel pipe piles in sand. Journal of
Zhusupbekov, A., Alibekova, N., & Morev, I. (2014). Dynamic-penetration method for bearing
capacity determination of precast piles. Soil Mechanics & Foundation Engineering, 51(2).
Chapter 5
Field Performance of Composite Piles
After late nineteenth century when the driven piles design was mainly based on
the experience, the research of pile behavior which suffered static load started
(Paikowsky and Tolosko 1999). Federal Highway Administration provided the static
analysis methods for designing of single piles, the capacity of the pile can be primarily
calculated through Meyerhof Method which is based on the result obtained from Stan-
dard Penetration Tests (SPT), Brown Method, Nordlund Method, α Method (total
stress method) and β method (effective stress method) for piles suffered from vertical
compression loading (Hannigan et al. 2016).
The ultimate bearing capacity of pile could be primarily estimated based on the
static analysis calculation and the empirical method. For example, based on the
Chinese code of JGJ 94-2008 (2008), the ultimate capacity of a single pile is related
to the end bearing (Qpk ) and shaft resistances (Qsik ), which are dependent on the
types and properties of the soil layers as well as piling technology. Similarly, in
Singapore, the ultimate axial load capacity can be determined based on the skin
friction coefficient (K s ), end bearing coefficient (K b ) and unit end bearing (qb ) (Xiao
and Yang 2011). However, due to the uncertainty such as the properties of soil layers,
where the obtained SPT N-value or corrected N value cannot appropriately represent
the properties of the soil, these methods can only be used for primarily design.
The SLTs are the mostly accepted method to determine the ultimate bearing
capacity of piles. These tests can provide accurate data and are commonly used
for investigating the pile behavior. The SLT mostly implements loads from 0 kN to
the value equal to 2 times the required anticipated design loads, and later consecu-
tively releases loads back to 0 kN (JGJ 94-2008 2008; ASTM D-1143/D-07 1994).
In some local standards, the maximum loading value may be equal to 1.5–4 times
the designed loads.
The Load-settlement (Q-s) curve is the typical results obtained after SLTs.
However, mostly, even though the maximum loading of SLT applied, there is no

https://doi.org/10.1007/978-981-33-6183-6_5
140 5 Field Performance of Composite Piles
extraordinary displacement increase observed from in-situ or from the Q-s curve.
This is because the SLTs are usually used for checking if the designed pile achieves
the project requirement instead of determining the ultimate bearing capacity. In
other word, these tests have not been performed to the failure of pile-soil system,
which leads to the requirement of interpretation of SLT results to ultimate capacity
determination.
In current practice, there are some similarities of pile capacity or pile failure
definition among standards. Borel et al. (2004) summarized that there are three types
of failures which are mostly used for determining the pile capacity. The first one
is Qu,F , which is known as plunging failure observed from the SLTs with rapid
increase of pile movement without an increase in loading (or increment loading is
small). The other two are asymptotic ultimate load (Qu,asym ) derived mathematically
or graphically from the Q-s curve and conventional ultimate load (Qu,conv ) which
is defined as the load causing a gross settlement equal to 0.1 times the diameter of
the pile or the load at which the penetration reaches a given value. Some standards
may believe that the Qu,conv occurred when different values of gross settlement was
discovered, for example, different from the value of 0.1 times the diameter, ASTM
D-1143/D-07 (1994) pointed out when the axial settlement achieves to 0.15 times
the diameter or width, failure occurs and the increment loading of SLT should stop.
The interpretations of data obtained from SLTs are Davission’s offset method,
double tangent method, DeBeer’s Log–log method, Brinch-Hansen’s method and
Chin-Kondner method. Methods of interpretation based on maximum allowable gross
movements overestimate the allowable capacities of short piles and underestimate
the abilities of the long piles. AASHTO (2002) recommended the commonly method
for interpreting the pile compression testes results was Davisson’s method (1972)
for driven piles and double tangent method for drilled shafts (Samtani and Nowatzki
2006). DeBeer’s method is the approach that plotting these data using logarithmic
scales (DeBeer 1970). There are two methods suggested by Brinch-Hansen (1963)
to define the failure conditions. They are 90 and 80% criteria, where the first criteria
defines the failure load as the load that is associated with twice the pile head movement
as obtained from 90% of the applied load, the 80% criterion defines the failure load
as the loading that is corresponding to four times the pile head movement obtained
from 80% of the load. Similarly, the Chin’s method can also be used to extrapolate
the SLTs based on the assumption that the Q-s relationship is hyperbolic and hence
the inverse slope of interpreted line is the ultimate bearing capacity (Chin 1970).
Through back analysis of the static pile load test in Singapore, Xiao and Yang (2011)
pointed out that, theoretically, the Chin’s method created two lines which inverse
slope of the first line representing the shaft resistance and the inverse slope of the
second line representing the ultimate resistance.
Nowadays, the utilization of piles is widespread, however, there are issues when
these piles are located in harsh environments, especially in marine or coastal condi-
tions. Piles made with traditional materials can be destroyed due to the corrosion of
steel, the deterioration of timber, and the degradation of concrete. The deterioration
of the timber, concrete, and steel piling systems costs the United States nearly 2
billion dollars per year for repair and replacement (Hassan and Iskander 1998). The
development of concrete piles has continued for more than 200 years, but engineers
are now facing pile-related problems because, even though piles are made of concrete
and steel, which has good rigidity and high strength, damage can still occur. Under
such condition, engineers started to find a new material that has better resistance
ability to face with the harsh environments.
Fiber-reinforced Polymer (FRP) is a composite material that contains resin and
fiber that can provide high tensile strength. This material is also known as fiber-
reinforced plastic or an advanced composite material (Bank and Wiley 2006).
FRP is well-known for its high ratio of strength to weight, high ratio of longitu-
dinal/transversal Young’s and high ratio of longitudinal/transversal shear modulus
(Pecce 2001). This material can be either Carbon-FRP (CFRP), Aramid-FRP (AFRP)
Glass-FRP (GFRP), or Basalt-FRP (BFRP), depending on the fiber used. The compo-
sition of the FRP product is, therefore, flexible depending on the material properties
and the volume ratio of fibers to resin, and the selection of types and orientation
of the fiber. Because of this flexibility, FRP composite material products have been
widely applied for reinforcement and rehabilitation.
FRP material can be manufactured into FRP laminar (slice/sheet) or FRP bars. FRP
bars are usually used for making the GFRP anchors (ribs) and GFRP grilles (Zhang
et al. 2001). The FRP bars can not only be used to replace steel bars inside concrete
structure, but also can be used as anchors for slope reinforcement and support in deep
excavation. The technology of using GFRP reinforcement as an anchor was adopted
in the 2008 Chinese Changji Expressway construction built for red sandstone slope
reinforcement. The results demonstrated the slope was overall stabilized (Luo 2014).
Moreover, when rehabilitation is required, on either upper structure such as beams,
or on columns, and sometimes on walls, the FRP laminar is utilized. The applica-
tion of the FRP slice using resin onto the surface of the cracked beam or wall on
the tensile side recovers the original strength of the components because the FRP
material provides good tensile resistance. A large amount of research has focused
on analyzing the structural behaviors of beams and walls with consideration of FRP
sheet application (Mohammed et al. 2013; Mosallam et al. 2015; Mostofinejad and
Tabatabaei Kashani 2013; Mostofinejad and Mohammadi Anaei 2012). A notable
method using confinement to concrete columns by FRP has been studied, and the
results indicated that this methodology can enhance concrete strength of the column
when the structure suffers axial loading (Kwan et al. 2015; Lo et al. 2015; Youssf et al.
2015). The application history of FRP composite materials has surpassed 80 years;
however, these previous projects mostly focused on rehabilitating or strengthening
the structure members, such as beams and columns.
The history of applying FRP piles is approximately three decades old. Dating
back to April 1987, the first prototype recycled pile was driven at The Port of Los
Angeles to replace the creosote-treated timber piles which successfully avoided any
threat to marine borers (Juran and Komornik 2006). As early as 1998, the Empire
State Development Corporation (ESDC) undertook a waterfront rehabilitation project
known as the Hudson River Park (Robinson and Iskander 2008). This project involved
replacing up to 100,000 bearing piles for lightweight structures. The concrete-filled
FRP composite piles were then employed by the Virginia Department of Transporta-
tion (VDOT) in 2000 for the new Route 40 Bridge over the Nottoway River in Sussex
County, Virginia (Pando et al. 2004).
Extensive investigation had conducted on the FRP composite piles. 4 types of FRP
piles were tested in Elizabeth, New Jersey. These tested FRP piles were concrete filled
fiberglass shell piles (Lancaster Pile), polyethylene piles reinforced with steel bars
(PPI Pile), polyethylene piles reinforced by fiberglass bars (SEAPILES) and solid
polyethylene piles (American Ecoboard Pile). The tests have shown possible applica-
bility of plastic piles to traditional axial loading applications and the further work of
long term creep performance and durability of these piles was highlighted (Robinson
and Iskander 2008). These piles data are analyzed based on Davisson’s method,
DeBeer’s method, Chin’s method and CAPWAP method. In addition, Concrete-
filled FRP piles came into notice and were considered the best FRP piles to resist
upper loading. From a different perspective, the other types of FRP piles showed low
capacity. This is acceptable when using these FRP piles because the steel pipe pile is
mostly used under the conditions of being exposed to water, which means the FRP
pile is mainly used for protection away from corrosion after replacing the steel pile.
Further example like SRP, is mainly used in fendering applications and regarded as
potential load-bearing piles (Guades et al. 2012). FHWA proposed the conclusion
that the FRP composite piles can be used effectively as vertical load-bearing piles
and represent an alternative for deep foundation construction, especially in waterfront
environments and aggressive soils (Juran and Komornik 2006).
For composite piles suffering from lateral loading, a test pile program using a stat-
namic device was conducted and comparisons made between pre-stressed concrete
piles and composite piles. It was concluded that the composite pile exhibited a
much lower stiffness than the pre-stressed concrete pile. This also illustrated that the
composite piles had the ability to sustain lateral load (Pando et al. 2004). To deter-
mine the lateral behavior of large diameter composite piles, tests were conducted by
Thomann and Zoli using two different tests, the inclinometer measurement and survey
measurement instead of a statnamic method. It was concluded that the measured
deflections were much higher than the pre-load test calculation results. It was also
concluded that when using post-load analysis, 85% reduction in soil strength was
required (Thomann et al. 2004). Apart from large diameter composite piles, two
other types of piles, namely the concrete-filled GFRP pipe pile and the standard pre-
stressed concrete pile, were analyzed to determine the behavior lateral loading and
the results compared. These indicated that the concrete-filled GFRP pile was more
flexible than the standard pre-stressed concrete pile. Also, the ultimate lateral load
capacity of the GFRP pipe pile was greater; however, it exhibited brittle behavior at
failure. Additionally, the results concluded that GFRP piles can be modelled using
P-y curves and classical beam theory (Weaver et al. 2008). The limitation of all these
previous investigations into passive composite piles is that there is no consideration
of soil remove during excavation.
As reviewed above, there has been numerous geotechnical applications of FRP
materials being performed, however, the research which aims to study the deflection
behavior of the FRP bar reinforced concrete piles is limited. In addition, the limitation
of the recent research associated with FRP composite material is that there are no tests
on precast concrete piles that are partially confined by FRP laminar. This Chapter
provides two case studies on the purpose of FRP composite pile’s research. The first
case study illustrates the capacity improvement of the composite piles under FRP
sheet confinement, and the second case study investigates the FRP bar reinforced
passive piles used in a deep excavation pit (tunneling project).
5.2 CFRP Laminar Concrete Composite Piles
5.2.1 Background Introduction
To investigate the composite piles’ capacity with FRP confinement effect, compres-
sive SLTs were conducted based on the research method illustrated in Chap. 2. The
tested piles were rectangular and pipe concrete piles, and each type of tested piles
was partially confined by CFRP. For the rectangular piles (P116 and P115), P116 was
applied with CFRP laminar with a space of 120 mm (width of CFRP of 100 mm).
The tensile strength of this fiber material was 760 kPa. The dimensions of these two
piles were the same, with a 400 × 400 mm2 cross-section area and total length of
25 m.
For the concrete pipe piles (P2108 and P2054), P2108 was applied with a CFRP
with material strength of 760 kPa, and the diameters and length of these two piles
were 400 mm and 25 m, respectively. The concrete grade of the pipe pile was C30
(150 × 150 × 150 mm3 cubic compression strength of 30 N/mm2 ) and the concrete
grade of the rectangular pile was C40 (150 × 150 × 150 mm3 cubic compression
strength of 40 N/mm2 ).
The subsurface exploration was discovered through laboratory and in-situ tests. The
In-situ tests such as standard penetration tests (SPT) and cone penetration tests (CPT)
and laboratory tests of consolidation tests, direct shear tests and triaxial tests were
conducted. The local standards of Chinese Code of Standard for Soil Tests Method
(GB/T 50123-1999 1999), Standard for Test Methods of Engineering Rock Mass
(GB 50021-2001 2001) were used in this project for the subsurface exploration.
Based on the borehole logs, the soil layers were discovered as follows.
1. Clay: yellowish brown, plastic, average thickness of 4.55 m; 2. fine sand:
loose, grey-dark, contains ferric oxide, average thickness of 1.98 m; 3. silty clay:
yellowish brown, plastic, partially contains fine sand, average thickness of 4.57 m;
4. sand: yellowish brown, fine to medium density, average thickness of 3.08 m; 5.
sand: brown, saturated, dense, 8 to 12% loess doll, diameter of 1 to 1.5 cm, average
thickness of 6.15 m; 6. strongly weathered glimmerite: average thickness of 2.52 m; 7.
medium weathered glimmerite: very dense, average strength of 14.12 MPa, average
thickness of 12.6 m.
As shown in Fig. 5.1, the concrete rectangular and pipe piles were transferred to
the required location. The CFRP (Fig. 5.2a) was cut into pieces, and then the glue
was prepared by mixing Type A and B resin together. Later, the CFRP pieces were
attached to the concrete surface using the admixed resin, ensuring the resin was
totally permeated through the FRP laminar. The prepared concrete rectangular and
pipe piles are provided in Fig. 5.3. After seven curing days for the composite material,
(a) Concrete pipe pile (b) Concrete rectangular pile
Fig. 5.1 Preparation of the concrete piles
(a) CFRP preparation (b) Resin preparation
Fig. 5.2 Preparation of the composite material

5.2 CFRP Laminar Concrete Composite Piles 145
(a) CFRP-confined pipe piles (b) CFRP-confined rectangular piles
Fig. 5.3 CFRP-confined concrete piles
these piles were ready to install into the ground, as shown in Figs. 5.4 and 5.5. To
enable comparisons between the CFRP-confined piles and the piles without FRP
application, the driven piles were installed close to each other so that the geotechnical
conditions were the same. Mind the overlapping length should be over than the
Fig. 5.4 Installation of

composite rectangular pile
Fig. 5.5 Installation of

composite pipe pile
recommended value from the product specification. For this case, the overlapping
length was 200 mm. Also, do not mixing Type A and B resin in a large amount
because chemical reaction is very fast. There will be inadequate time for operation
(the glue has already hardened).
5.2.4 Test Setup
Similar to the setup depicted in Chap. 4, these SLTs were conducted based on the
Chinese Technical Code for Testing of Building Foundation Piles (JGJ 106-2014
2014). The setup of the SLT is provided in Fig. 5.6. As shown in Fig. 5.7, two
hydraulic jacks were used to provide loads to the pile head. The low-speed main-
tenance methods were used in this case study, and the applied load was maintained
until the rate of axial movement did not exceed 0.1 mm. After each load was applied,
four dial gauges recorded the vertical movement of piles with intervals of five, 10,
15, 15, 15, 30 and 30 min, and one hour duration was required if accumulated time
exceeded two hours.
Fig. 5.6 Compressive SLT
Fig. 5.7 Equipment for SLT
For rectangular piles P116 and P115, the loading started at 976 kN and then
progressed in 488 kN increments until the maximum loading of 4,880 kN, and then
the load was consecutively released back to 0 kN. For the pipe piles, the loading
started at 260 kN, with increments of 130 kN until the maximum loading of 1,300
kN, and then the loading was consecutively released back to 0 kN.
5.2.5 Results of CFRP Laminar–Confined Concrete Piles
The pile head settlements corresponding with each load were recorded, and the load–
settlement curves (Q-s) of these four concrete piles are provided in Figs. 5.8 and 5.9.
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14
Fig. 5.9 Q-s curves of tested pipe piles
As illustrated in Fig. 5.8, after the maximum loads of 4,880 kN were applied, the
maximum settlements of the traditional concrete rectangular pile and the pile with
CFRP application were 26.435 and 24.69 mm, respectively. As shown in Fig. 5.9,
after the maximum loads of 1,040 kN were applied, the maximum settlements of the
traditional concrete pipe pile and the pipe pile with CFRP application were 12.46
and 10.77 mm, respectively.
5.2.5.2 Settlement-lg (Load) Curves
The settlement-lg load (s-lgQ) curves of these four piles are provided in Fig. 5.10.
Similar to DeBeer’s method, which uses lg settlement and lg load data to plot the
lg Load (kN)
1 10 100 1000 10000
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

  P2108-CFRP Confined Pipe
Pile
25


 



 P2054-Concrete Pipe Pile
30
Fig. 5.10 s-lgQ curves of four tested piles
diagram, it was difficult to determine the point that represented the extreme settlement
decreasing trend. Thus, the ultimate bearing capacity was difficult to determine and
some interpretations were required.
5.2.5.3 Settlement-lg (Time) Curves
The settlement-lg time (s-lgt) curves of these four tested piles are provided in
Figs. 5.11, 5.12, 5.13 and 5.14, respectively. These four figures indicate that the
CFRP-confined piles presented less settlement during each loading stage than did
the piles without CFRP application. If there was an extraordinary settlement distance
between two adjacent loading stages, the failure condition occurred, and the ultimate
capacity of the pile could be determined. However, as shown in these figures, it
was difficult to determine such loading stages. Another method is to determine a
load line that illustrates a decreasing trend, which indicates that the soil foundation
system is unstable under one loading, and the previous load is then believed to be
the failure load; however, there was no such line acquired. This also illustrated that
these four piles did not suffer from plunging failure, and interpretation was required
to determine the ultimate bearing capacity.
5.2.5.4 Interpretations of the Static Load Tests
Double Tangent Method
The rectangular pile results based on the double tangent method are provided in
Fig. 5.15. By using two tangent lines and determining the intersection, it was found
lg Time (mins)
5 50 500
0
 
                 
                
                 
                
                 
                
              
5
976kN
Settllement (mm)
10 

 
 

 
  
 
 
 
  
 
  
  
 
  
 
 
 
 
 
 
1464kN
                
                 
                
               
       
1952kN
15 2440kN
2928kN
20
.. ... ... .......
.... . .. ..
........ .....
...............
................
.. .. .. ..
.. ..
... ... ....
................
. ... ..
................
.. .............
.. ..........
.. .. .. ..
..............
......... . ... ..
...... ....
................
........... .
. .
.. .. .. .
......
.. .. ..
3416kN
..
 
 

 
3904kN
25
4392kN
..... . ... ...

. ........
. .
4880kN
30
Fig. 5.11 P116 CFRP-confined concrete rectangular pile
lg Time (mins)
5 50 500
0 

   

 


 

   
      
  
      
5
      
      
      

           


 
976kN
Settllement (mm)
10 






  
  



  



 



 



  



1464kN


 1952kN
15 2440kN
      
      
   
2928kN
      
         


20 3416kN



3904kN
25 4392kN
4880kN
30
Fig. 5.12 P115 concrete rectangular pile
that the ultimate bearing capacities of the CFRP-confined pile (P116) and normal
pile were 2,600 and 2,400 kN, respectively. As illustrated in Fig. 5.16, based on
the double tangent method, the ultimate bearing capacities of the CFRP-confined
concrete pipe pile (P2108) and normal concrete pipe pile were 850 and 760 kN,
respectively. It should be noted that this method uses empirical understanding to find
the tangent line; thus, to some extent, the results may be inaccurate.
lg Time (mins)
5 50 500
0
  
    
      

      
     
2 

  
 
 
   
    
        
       




  260kN
Settllement (mm)
4 390kN



  











 





 



  



 





  520kN
6 650kN
780kN
8 



 



 


  



 

 





 910kN
      
1040kN
10 


1170kN
1300kN
12
Fig. 5.13 P2108 CFRP-confined concrete pipe pile
lg Time (mins)
5 50 500
0 





 


 


 


  





 
      
2 




  















4 


260kN
Settllement (mm)









 

  390kN
6
  
        
     
520kN




650kN
8

  
780kN
      
    
  
 
910kN
  
10



1040kN
12 



1170kN
1300kN
14
Fig. 5.14 P2054 concrete pipe pile
DeBeer’s Method
Based on DeBeer’s method (as illustrated in the review in Chap. 2) of plotting the
settlement and loading into the lg–lg values, as shown in Figs. 5.17 and 5.18, it was
found that the ultimate bearing capacities of these four piles via interpretation were
very difficult to determine. This was because DeBeer’s method is mostly used in the
plunging failure tests.

0 1000 2000 3000 4000 5000 6000
0
 


 

 

 

 



  
  


5

 














10
Settlement (mm)

 
 




 


 
 
15
   
 
 
 
 
 
 

20

 




    
  
 
  
 


 

25

  
 
P116


  


P115
 


30
Fig. 5.15 Double tangent method of rectangular piles

0 200 400 600 800 1000 1200 1400
0 

  









 


2
 
 






 

4



Settlement (mm)

 

  




 
6
 
 
  

  
 


 


8


  


 





  
 
10
 



 
 
  
12




  


 


P2108
 


P2054
14
Fig. 5.16 Double tangent method of pipe piles
Davisson’s Offset Method
Based on Eq. (2.37), the elastic lines could be determined. After offsetting the elastic
lines, the results are provided in Figs. 5.19 and 5.20. It was found that there was no
intersection between the offset line and Q-s curve (loading stages), which illustrated
that the ultimate bearing capacity was over the maximum applied loading, and other
interpretations are required.
lg Load (kN)
100 1000 10000
0.1
1
lg Settlement (mm)
















10











 







 

 
  P116


P115
100
Fig. 5.17 DeBeer’s method of rectangular piles
Chin’s Method
Chin’s (1970) method operates under the assumption that the load–settlement rela-
tionship is hyperbolic. By plotting the displacement/load versus the displacement, the
trend line provides a slope, and the reverse of this slope gives the ultimate bearing
capacity. Theoretically, the curve of settlement versus the ratio (settlement/load)
should comprise two straight lines (Chin 1978). In the plotted diagrams for P116
(Fig. 5.21), P115 (Fig. 5.22), P2054 (Fig. 5.23) and P2108 (Fig. 5.24), the reverse of
the first line represents the shaft resistance, and the reverse of the second line gives the
ultimate resistance. The end bearing could then be determined through calculations
with the obtained values of shaft and ultimate bearing resistance.
Eight functions were found and are provided in these figures. Based on these
functions, for the traditional precast piles, the slopes of P115 and P2054 are deter-
mined as 0.0001 (Qu) and 0.0005 (Qu), respectively, as depicted in Figs. 5.22 and
5.23. For the CFRP confined piles, the slopes of P116 and P2108 are determined as
0.0001 (Qu) and 0.0004 (Qu), respectively, as illustrated in Figs. 5.21 and 5.24. It
can be found that, for the concrete pipe pile, the ultimate bearing capacity of the pile
with CFRP application is determined as 2500 kN with shaft capacity of 909.1 kN.
For the concrete rectangular pile, the ultimate bearing capacity of pile with CFRP
application is determined as 10,000 kN with shaft capacity of 1428.6 kN. Table 5.1
summarises the results of these four piles based on different methods.
lg Load (kN)
100 1000 10000
0.1

 



1
lg Settllement (mm)


 


 

















10

 


 




P2108


P2054
100
Fig. 5.18 DeBeer’s method of pipe piles

0 1000 2000 3000 4000 5000 6000
0 
 

 
 

 

 

 


5
 

 

 



 

 

 

10
 

Settlement (mm)
 

 



  
  

 
 
  
15
  
  
  
  
  
  
 
  
  
  
20

 
 
  

   

 
  
  
  
 
  
  
  
25
  

  
P116
  
  
 
  

 
30 P115


 
Elastic Line
35
Offset Line
40
Fig. 5.19 Davisson’s offset method of rectangular piles

5.3 GFRP Bar–Reinforced Concrete Composite Piles 155

0 500 1000 1500
0 
  



 

 











5
 
 
 
 
 
  

  
  
Settlement (mm)
 




 
 
 
 
 
  
10
 
  

    
 
 
 




15


  P2108

P2054
20

Elastic Line
Offset Line
25
Fig. 5.20 Davisson’s offset method of pipe piles
0.006
0.005






0.004 


y = 0.0001x + 0.0017


  
R² = 0.99
0.003











Qs 

 Qu
0.002



  


 
y = 0.0007x + 0.0001
R² = 0.9421
0.001



0





0 5 10 15 20 25 30
Settlement (mm)
Fig. 5.21 Chin’s method of P116
5.3 GFRP Bar–Reinforced Concrete Composite Piles
As stated in Chap. 2, intensive research has focused on the field test on pile foundation
yet has seldom considered the passive pile with FRP bar reinforcement, and has
very limited analysis of the deflection behaviour of this pile under changed earth
pressure. Further, lateral SLTs are short-term loading, and the research associated
with long-term loads and deflection behaviour is limited. As discussed in Chap. 4, the
research on non-displacement piles commenced in a tunnelling project, that project
0.006
Settlement/Load (mm/kN) 0.005

 


 


0.004  


 
y = 0.0001x + 0.0017
 


  
R² = 0.99
0.003



 


 
Qs Qu

  
0.002
  



 

y = 0.0007x + 0.0002


0.001 R² = 0.9304
0
 


0 5 10 15 20 25 30
Settlement (mm)
0.012
0.01 


0.008
 


 
y = 0.0005x + 0.0032
 


 

0.006




 


R² = 0.9759





Qs Qu
    
 
0.004


 


 
0.002 

y = 0.0014x + 0.0013

R² = 0.9876
0
0 5 10 15
Settlement (mm)
contained two parts—the first part used non-displacement piles, and the second part
used bored piles. The study in this section was based on the project illustrated in
Chap. 4, Sect. 4.3.
As shown in Fig. 5.25, the excavation construction required passive piles to resist
the lateral loading. As a result of the location of P1 obstructing the Tunnelling Boring
Machine (TBM) for tunnelling, where the TBM would have to break through, use
of the traditional steel-reinforced piles was not appropriate and even dangerous to
do so. The reason is that the TMB can easily be wound with steel reinforcement.
The TBM will stop boring underground, and the labour required to untangle the
wound steel. In such a case, the potential for loss of life is high because the broken
0.009


 
0.008


 
0.007





y = 0.0004x + 0.0039
0.006 







 
 

R² = 0.9956

 
0.005 



Qs Qu

  

  
 

0.004  


 
y = 0.0011x + 0.0022
0.003

 

R² = 1
0.002
0 2 4 6 8 10 12
Settlement (mm)
Table 5.1 Results of piles under different interpretation methods

Methods Double tangent Davisson’s criteria Chin’s method
Pile Label Qu (kN) Qu (kN) Qu (kN) Qs (kN) Qt (kN)
P116 2600 >4880 10,000 1428.6 8571.4
P115 2400 >4880 10,000 1428.6 8571.4
P2108 850 >1300 2500 909.1 1591.0
P2054 760 >1300 2000 714.3 1285.7
Fig. 5.25 Construction plan map (not to scale)

Table 5.2 Support

Type of Support Excavation Excavation
installation information and
support installation date duration
excavation duration
date
Concrete 18 March 22 April 2016 34 days
support 2016
Steel 22 April 2016 27 May 2016 35 days
support
Steel 27 May 2016 3 July 2016 36 days
support
N/A 3 July 2016 29 July 2016 26 days
concrete and the adjacent soil is prone to collapse. By taking advantage of the GFRP
property, the project decided to use FRP bar reinforced concrete pile instead of the
steel reinforced pile from the soft-eye opening area (Type A). As shown in Fig. 5.25,
P1 was a GFRP-reinforced concrete pile, while the adjacent pile P258 was a steel-
reinforced pile (Type B). These two piles’ deflection were monitored every day by
considering the support strut application and excavation process for the deflection
behaviour analysis.
To investigate the pile deflection, it is vital to consider the construction process.
In this particular construction, three supports were used to help the piles resist the
changed earth pressure caused by excavation. The construction process is illustrated
as follows: (1) The concrete supports were first installed (no earth pressure) and then
the excavation began. These passive piles started to resist the loading caused by soil
movement. Note that near pile Type C (Fig. 5.25), anchorage was installed. (2) After
6.5 m of soil was removed, the second support (steel) was installed. Following this,
a further 4.8 m of soil was removed, and the third support strut (steel) was installed.
(3) After the third steel support was installed, an additional 3 m of soil was removed,
and the total excavation depth of 16 m was achieved.
As summarised in Table 5.2, three supports were installed on 18 March, 22 April
and 27 May 2016. Once the support installation was finished, the excavation durations
were 34, 35 and 36 days, respectively. To monitor the behaviour of the passive pile,
with the aim of checking if these passive piles were stable after the final excavation
finished, a further 26 days of monitoring was required.
After conducting the laboratory and in-situ tests and interpreting the borehole logs
near these tested two piles, the subsurface condition was determined. The simplified
soil profile contained seven layers, and the parameters of each layer are summarised
in Table 5.3. As shown in Fig. 5.26, these piles reached the seventh layer with embed-
ment of 5.69 m. It should be noted that Table 5.3 was determined from boreholes
near these monitored two piles. For other pile investigations, the soil profile may be
different.
Table 5.3 Properties of simplified soil layers

Soil Thickness ω Gs ρ ρd Es e wL wP IP IL c ϕ
layer (m) % – g/cm3 g/cm3 MPa – % % – – kPa o
Fill 2.5 N/A N/A 1.8 N/A N/A N/A N/A N/A N/A N/A N/A N/A
Loess 2.9 24.5 2.7 1.93 1.55 9.3 0.75 30.7 18.9 11.8 0.55 29 24
Silt 1.2 23.2 2.69 1.96 1.59 15.8 0.69 27.5 18.8 8.7 0.51 19 20
Silty 4.5 26.4 2.72 1.95 1.54 9.1 0.77 32.5 19.7 12.8 0.53 36 22
clay
Fine 2.6 N/A N/A 1.96 N/A N/A N/A N/A N/A N/A N/A 0 22
sand
Silty 6 27.3 2.72 1.93 1.52 7.2 0.8 34.3 20.7 13.8 0.47 42 16
clay
Clay 4.0 28.6 2.75 1.93 1.51 6.9 0.80 43.5 23.5 20 0.3 38 15
Fig. 5.26 Soil layers adjacent to pile (not to scale)

These two monitored piles, P1 (GFRP bar reinforcement) and P258 (steel bar rein-
forcement), were designed to be 22 m long, with pile diameters of 700 mm. The
manufacture process of GFRP bar–reinforced concrete piles and steel bar–reinforced
concrete piles involves driving holes, assembling reinforcement, and transferring into
the holes and casting.
The GFRP stirrups were 10 mm in diameter with spacing of 150 mm (Fig. 5.27).
The longitudinal reinforcement was 26 GFRP bars with diameters of 28 mm
(Fig. 5.28). The ultimate tensile strength of the GFRP bar was over 500 MPa and
the designed grade of the concrete was C30. Based on the local standard of Code
for Design of Concrete Structures GB 50010-2002 (2002), the cubic compression
strength of C30 was 30 MPa (the dimensions of the cubic concrete sample were
150 × 150 × 150 mm3 ). The GFRP cage was assembled as shown in Fig. 5.29. The
Fig. 5.27 GFRP stirrups of

piles
Fig. 5.28 GFRP

reinforcements of piles
Fig. 5.29 Assembled GFRP

cage of pile
traditional steel-reinforced concrete pile was also made of C30 concrete. The stirrups
were 16 mm diameter HPB300 (hot-rolled plan) bars with spacing of 150 mm. The
horizontal reinforcements were 26 HRB400 (hot-rolled ribbed) bars with diameters
of 28 mm. The ultimate compression strength of the concrete was 30 MPa, and the
yield strengths of HPB300 and HRB400 were 300 and 400 MPa, respectively. The
steel cages were welded. In Fig. 5.30, the left side of the figure presents the completed
steel cages. Table 6.4 summarises the parameters and prices of the GFRP and steel
reinforcements.
As illustrated in Table 5.4, the cost of GFRP material was approximately four
times that of the steel reinforcement. If substituting all steel-reinforced concrete
piles with GFRP-reinforced concrete piles, this project (516 piles, all with a length
of over 20 m) would cost extra ¥3,096,000 (AU$595,384.70) or more. Nevertheless,
this GFRP design is underestimated its capacity through applying high safety factor
due to its rare usage.
Fig. 5.30 Assembled steel

cage of pile
Table 5.4 Parameters and prices of reinforcements

Type of pile GFRP pile Steel pile
Reinforcement Horizontal reinforcement Stirrup Horizontal reinforcement Stirrup
Brand GFRP500 GFRP bar HRB400 HPB300
Diameter 28 mm 26 mm 26 mm 26 mm
Yield strength 500 MPa 300 MPa 400 MPa 300 MPa
Price ¥18.5/m ¥18.5/m ¥4.8/m ¥4.8/m
5.3.4 Test Setup and Procedures
The measurement system to determine the lateral deflection was a plastic PVC tube
and a readout probe. The internal diameter of the PVC tube was 75 mm with four
grooves cut at 90° intervals. Figures 5.31 and 5.32 illustrate the installation of PVC
Fig. 5.31 PVC tube on

GFRP cage
Fig. 5.32 PVC tube on steel

cage
Fig. 5.33 Readout probe for deflection measurement of pile
tubes on a GFRP cage and steel cage, respectively.

The readout probe that fit into the grooves is demonstrated in Fig. 5.33. The amount
and location of horizontal movement in a deep foundation could be determined
through lowering the readout probe into the bottom of the PVC tube and pulling
upwards. During system installation, the top of the PVC was covered by a cap to
avoid the concrete being poured inside and was then covered again with a weaving
bag in case the plastic cap being damaged by coarse aggregate.
After the reinforcement cages were assembled and the PVC tubes were installed,
the cages were transferred to the acquired location. The cages of each pile were
lifted by two cranes, and were then filled into the guide slots. To ensure the concrete
cover met the requirements, concrete cushion blocks were used with spacing of 2 m.
Figure 5.34 shown the installation of the steel-reinforced concrete bored passive pile.
After the installation of the GFRP and steel reinforcements into the bored holes, fresh
concrete was poured into the holes.
Because of the installation of concrete support which ultimate compression strain
is very small (0.003), assuming that pile head should not move and the accumulated
deflection value changes from pile head to the pile toe. The lateral displacement
measurement was recorded every day during the support installation and excavation
process.
5.3.5 Results of GFRP Bar–Reinforced Concrete Piles
The concrete supports were cast before excavation. As shown in Fig. 5.35, the exca-
vator dug out soil after the concrete support was cured (cured date: 18 March 2016).
It was believed that there was no lateral displacement of the pile at this stage, since
Fig. 5.34 Installation of steel-reinforced concrete pile
Fig. 5.35 Excavation after first concrete support being installed

the earth pressure was relatively small during the first excavation. However, there
were three excavators working together, which created extra pressure to push the pile
head outside the excavation site.
The obtained lateral displacement of the upper part of the GFRP pile was discov-
ered within +0.5 mm (negative value represents load direction pointing to foundation
pit), possibly caused by the excavators. This small value could also have been caused
by the accuracy of the inclinometer.
Under the assumption of zero movement of the pile head, the accumulated
displacement values were overestimated by calculation, especially from the pile
bottom. As shown in Fig. 5.36, the data showed that, during the five days of exca-

-5.0 -3.0 -1.0 1.0 3.0 5.0
0 
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Concrete support
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Changed earth pressure due

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10 
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Depth (m)
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20-03-2016
20
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23-03-2016
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01-04-2016
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Fig. 5.36 Deflection of CFRP pile at the beginning of excavation

vation from 18 to 23 March 2016, lateral deflection was detected within +0.4 mm
from the pile head (error because of equipment or effect of the excavators), and
the accumulated lateral deflection was +0.73 mm at the bottom of pile, which was
calculated by adding the error values from the pile top.
As the excavation continued, the earth pressure started to increase, which meant
that the data could be accurately captured by the equipment. As shown in Fig. 5.36,
after 1 April 2016, the recorded data could be trusted because a negative value
occurred, which represented that the earth pressure had pushed the pile to the
excavation pit.
The first and second steel supports were installed on 22 April and 27 May
2016, respectively, as shown in Figs. 5.37 and 5.38. After these three supports
Fig. 5.37 First steel support applications
Fig. 5.38 Second steel support applications

being installed, the deflection behaviours of the GFRP-reinforced concrete pile were
recorded, as shown in Fig. 5.39a, with labels a, b and c representing the collected
data after installation of the three supports. As demonstrated in Fig. 5.39b, during
26 days of continuous data collection (no soil excavation), the horizontal movement
remained stable.
Through analysis of the data obtained from the inclinometer (Fig. 5.39a), it was
detected that the lateral deflection of the FRP-reinforced concrete pile was increasing
as the excavation depth was increasing. It was detected that the maximum horizontal
deflection was −1.56 mm at the depth of 12 m when the second support was installed
(the first steel support). After the first and second steel supports were installed, the
maximum deflections were discovered at the depths of 12.5 m, with values of −4.73
and −6.02 mm, respectively.
Additionally, after the last steel support was installed, the data collection
continued. The deflection became stable after 3 July 2016 (Fig. 5.39b). It was also
detected that the horizontal movement of the pile toe was stable at the value of −
1.1 mm, which led to the heave in the excavation pit (also discovered by total station).
This phenomenon proved the correct assumption of zero movement from the pile
head.
The steel-reinforced concrete pile was also simultaneously monitored. After
approximately 15 weeks of monitoring, the lateral deflection data of the traditional
steel-reinforced concrete pile obtained from the inclinometer system demonstrated
that, after the concrete support and two steel supports being installed, the maximum
lateral deflections were −4.70 mm (at the depth of 12.5 m), −7.95 mm (at the depth
of 12.5 m) and −10.56 mm (at the depth of 12.5 m), as shown in Fig. 5.40. For
comparison, the deflection behaviour of these two piles is summarised in Table 5.5.
This chapter has investigated composite piles with FRP application. Two types of
concrete piles (small and large displacements piles) were examined, with considera-
tion of the CFRP confinement effect. The conventional results such as load-settlement
curves and interpretations of the compressive SLTs were presented to determine the
ultimate bearing capacity. Furthermore, in this Chapter, another two bored piles
were monitored to research the deflection behaviour of the traditional steel rein-
forced and GFRP-bar reinforced pile foundations. The long-term load deflection
behaviour with consideration of support strut installation and excavation process was
provided. In addition, the behaviours between the GFRP-bar reinforced concrete pile
and traditional steel bar-reinforced concrete pile were compared in the last section.
168
Horizontal Movement (mm) Horizontal Movement (mm)

-8.0 -3.0 2.0 7.0 -8.0 -6.0 -4.0 -2.0 0.0 2.0 4.0 6.0
0 Concrete 0
Concrete support Changed earth pressure
Concrete support Changed earth pressure
5 5
Steel support Steel support

After 1st support installed
After 2nd support installed
After 3rd support installed
10 10
Steel support Steel support
Depth (m)
Depth (m)
15 15
Embedded depth 22-04-2016 03-07-2016
20 27-05-2016 20 22-07-2016
03-07-2016 29-07-2016
Fig. 5.39 a Lateral deflection of GFRP pile during three stages; b Total horizontal deflection of FRP pile
5 Field Performance of Composite Piles
Horizontal Movement (mm) Horizontal Movement (mm)
-12.0 -7.0 -2.0 3.0 -12 -7 -2 3
0 0
Concrete support Changed earth pressure Changed earth pressure

5 5
Steel support
20-03-2016 20-03-2016
22-04-2016 22-04-2016
10 20-05-2016 10 20-05-2016
Steel support 21-06-2016
21-06-2016
03-07-2016 03-07-2016
Depth (m)
Depth (m)
15 15
Embedded depth
20 20
Fig. 5.40 Deflection of a GFRP bar–reinforced and b steel bar–reinforced concrete bored piles
169
Table 5.5 Deflection behaviours of CFRP- and steel-reinforced concrete piles

Construction condition GFRP-reinforced Concrete Pile Steel-reinforced concrete pile
(after) Maximum Depth (m) Maximum Depth (m)
deflection (mm) deflection (mm)
First concrete support −1.56 12.0 −4.70 12.5
installed
First steel support −4.73 12.5 −7.95 12.5
installed
Second steel support −6.02 12.5 −10.56 12.5
installed
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Hassan, M., & Iskander, M. G. (1998). State of the practice review in FRP composite piling. Journal
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Chapter 6
Field Tests of Super-Long and Large
Diameter Piles
A compressive loaded pile is a slender element that transfers the loading from the
upper structure to the soil layers. Nowadays, because pile foundations with large
capacities are required, increasing numbers of super-long and large-diameter piles
are being designed for high-rise buildings and bridge projects. Usually, in China, a
pile longer than 50 m with a diameter greater than 0.8 m is recognised as a super-long
and large-diameter pile foundation (Yang and Zhong 1998). Lin et al. (1999) pointed
out that the pile with length to diameter ratio of 100 or more, though the length may
be less than 50 m, can also be deemed to super long pile in practice.
Numerous studies have been conducted to examine the compressive loaded piles
including encompassing H-steel piles, pretensioned spun high-strength concrete
(PHC) piles and cast-in-situ piles with post grouting (Zhou et al. 2012a, b; Wang et al.
2015; Feng et al. 2016, Li et al. 2016; Michael and Jae 2017). Most of the research has
focused on the behaviour of these piles with respect to different geotechnical condi-
tions, driving process, analytical method, theoretical load transfer method and finite
element method (Zhou et al. 2014; Zhang et al. 2014; Zou and Zhao 2016, Sun et al.
2016; Oliaei and Ghotbi Siabil 2017; Lu et al. 2019). Previous investigations also
include self-balanced testing method, determination of the effective length, effect of
soil stiffness and negative skin friction of the large diameter and supper long pile
foundation (Zhang et al. 2012, 2014; Zhou et al. 2015; Xing et al. 2017).
Static field tests are one of the most accurate methods to provide valuable informa-
tion, such as pile capacity and the load transfer mechanism. However, it is preferable
to perform these tests on piles less than 40 m in length. For a super-long pile, the
capacity always achieves up to 10,000 kN, which means that more than 1000 t of
reaction system are required. Thus, when using a weighted platform, the platform is
unsafe and prone to collapse. Moreover, if using reaction beams, the beams are prone
to fail because the bending movement is too great. Under such conditions, other test
methods are preferred, such as dynamic load tests and O-cell tests. However, these

https://doi.org/10.1007/978-981-33-6183-6_6
174 6 Field Tests of Super-Long and Large Diameter Piles
tests have some limitations. Some types of pipe piles or H-steel piles, which do not
contact the rock, may behave different under dynamic and static loading conditions
(Hannigan et al. 2016). For O-cell testing, many engineers believe that the method
is too costly because the O-cell is cast inside the pile (and is not recyclable) and the
pile will be damaged after load is applied.
This chapter presents a case study referring to super-long and large-diameter
piles, and introduces an improved static load test (SLT) method. In this research,
three drilled shaft pile foundations with various lengths and diameters were tested in
a bridge project. Different to the previous chapters which focused on the compressive,
uplift or horizontal ultimate bearing capacity, this chapter however, aims to study
the behaviour relating to load transfer mechanism and shaft resistant development
and distribution of the super-long and large diameter pile foundation. It is expected
that the results of this field tests’ research will provide practical information for
geotechnical engineers.
6.2 Compressive Static Load Tests
6.2.1 Geotechnical Conditions
A bridge project was constructed in Henan Province, China. The subsurface explo-
ration was conducted based on the Specifications for Highway Engineering Geolog-
ical Remote Sensing (JTG/T C21-01-2005 2005) and Standard for Soil Test Method
(GB/T 50123-1999 1999). Based on the interpretation of borehole logs, the soil layers
near these three tested piles were determined as illustrated in Fig. 6.1. It was found
that the main layers were dense to very dense sand and silty sand. Further, it was
found that the thickness of the soil layers varied, and the soil layers were loose in
the very upper region and dense to very dense in the rest of the region. As shown
in Fig. 6.1, all three piles were embedded in the bearing stratum with depth of 1 m.
The bearing stratum are dense sand, very dense grait and dense grait for piles with
labels of P12, P66 and P105, respectively. Further, in order to determine the shaft
resistance, strain gauges were installed in each soil layer. An example in relation to
the strain gauge location is also demonstrated in this figure.
6.2.2 Description of Tested Drilled Shaft Piles
Figure 6.2 presents information of the tested piles. As shown in the figure, the diam-
eter of P12 and P66 are the same of 1500 mm, and the diameter of P105 is 1800 mm.
The pile lengths varied, the shortest pile is 52 m, the medium length pile is 73 m,
and the longest pile is 83 m. All pile foundations were made of concrete with the
same concrete strength of 30 MPa. The designed capacity for the P12, P66 and P105
Tickness Elevation Layers Soil Condition P12 Tickness
Elevation Layers Soil Condition P66 Tickness Elevation Layers Soil Condition P105
Layer 1a 3 95.758-92.758 Silty Clay Very soft Layer 1b 3 96.466-93.686 Silty Sand Loose to medium dense Layer 1c 2.8 97.110-94.310 Silty Sand Loose
Layer 2a 3.808 92.758-88.950 Silty Clay Soft Layer 2b 6 93.686-87.396 Silty Sand Medium dense Layer 2c 3 94.310-91.310 Silty Sand Loose to medium dense
Layer 3c 4 91.310-87.310 Silty Sand Medium dense
Layer 3a 2.6 88.950-86.350 Silty Clay Soft to medium stiff
Layer 4a 3.1 86.350-83.250 Silt Loose to medium dense Layer 3b 3 87.396-84.496 Fine Sand Medium dense Layer 4c 5.3 87.310-82.010 Silty Sand Dense
Layer 5a 2.6 83.250-80.650 Silty Sand Medium dense Layer 4b 10 84.496-74.096 Fine Sand Dense
Layer 5c 5.3 87.310-82.010 Silty Sand Very dense
Layer 6a 10.6 80.650-70.050 Fine Sand Dense
Layer 6c 5.3 87.310-82.010 Silty Sand Loose to medium dense
Layer 5b 14 74.096-60.046 Fine Sand Very dense

6.2 Compressive Static Load Tests
Layer 7c 5.3 87.310-82.010 Sand with gravel Medium dense
Layer 6b 12 60.046-48.216 Fine Sand Very dense

Layer 8c 5.3 87.310-82.010 Silty Clay Stiff
Layer 7b 6 48.216-42.089 Fine Sand Very dense Layer 9c 5.3 87.310-82.010 Cementation Weak
Layer 9a 20.778 42.128-21.350 Silty Sand Dense to Very Dense Layer 8b 6 42.089-36.269 Silty Sand Very dense
Layer 9b 9 36.269-26.969 Silty Sand Very dense
Layer 10b 3 26.969-24.419 Silty Sand Very dense

3 24.419-21.010 Grait Very dense
Pile Label Dia. Length Concrete Strength 12 83.01.-71.010 Grait Dense

(mm) (m) (MPa)
P12 1,500 52 30
P66 1,500 73 30
P105 1,800 83 30
Fig. 6.1 Subsurface conditions and tested piles (not to scale)

175
Fig. 6.2 Simplified diagram of the tested piles (not to scale)
are 28,000, 40,000 and 64,000 kN, respectively, and the reaction frame capacity is
designed to be 1.2 times the designed capacity which are 33,600, 48,000 and 76,800
kN, respectively.
The locations of the strain gauges are also illustrated in Fig. 6.2. These locations
are selected according to the soil layers’ locations. All the strain gauges are installed
symmetrically in case one of the strain gauges being failed (if a strain gauge failed,
the other symmetrically installed strain gauge with the same depth can be used for
data collection; if there is no failure of strain gauges, the average values can be
used). The telltale is also installed in each pile so that the pile end displacement can
be determined.
In each pile, the reinforcement ratio from the upper part is twice the lower part.
The diameter of the auxiliary steel bar and the main longitudinal steel are all 28 mm in
6.2 Compressive Static Load Tests 177
diameter, and the diameter of the stirrup is 25 mm (the upper part spacing is 100 mm,
other part is 150 mm). Though total reinforcement ratio of these three tested piles
are different, the reinforcement effect is ignored in this research. This is because
this case study is researching the capacity under compressive loads, all these applied
loads are resisted by concrete material, the compressive contribution by the steels
are very small.
6.2.3 Test Setup
6.2.3.1 Reaction System
The traditional reaction system (SLT) is mostly used to test piles with a capacity lower
than 10,000 kN (or 1000 t) to avoid the reaction beams suffering too much bending
moment. In this case study, the maximum reaction frame capacity required 76,800
kN (1.2 times the maximum test load); thus, the traditional reaction beam could not
be used. Under this condition, a reaction system was designed. As shown in Fig. 6.3,
this reaction system contained eight anchoring piles that were connected with the
concrete beams and steel strand. The load on the reaction device (provided by the
hydraulic jacks) could be resisted by the steel strand, and the strand force could be
resisted by the vertical friction force (by an anchoring pile) and horizontal force (by
a concrete beam). There was a small bending moment acting on the reaction device;
hence, the reaction frame capacity was improved. As shown in Fig. 6.4, hydraulic
jacks were used to provide loads, and a reaction device was employed. Two reference
steel beams were also used, so that dial gauges could be installed.
Steel strand Reaction device

6
Anchoring 7pile 8
Hydraulic jacks
Concrete beam
Reaction column
Concrete beams 11
10
4 9 5
Anchoring pile
Tested pile Tested pile
1 2 3
(a) Plain view of test set-up (not to scale) (b) Mechanical mechanism
Fig. 6.3 Reaction system design

Reaction device
Hydraulic jacks
Reaction column
Anchoring
Reaction beam
Reaction pile
Test pile head

Reference steel beam
Fig. 6.4 Anchoring reaction system
6.2.3.2 Experimental Principle
The strain of the rebar can be determined through the ratio between the stress and
modulus of the rebar, as illustrated in Eq. (6.1). The stress of the rebar at a time of i can
be calculated by using the force values transferred through the rebar over the rebar
cross-sectional area, as depicted in Eq. (6.2). Before the pile foundation is loaded,
the initial coefficients ( f 0 and T0 ) should be determined. As shown in Eq. (6.3), the
force transferred from the pile can be determined after recording the coefficients at
a time of i. It should be noted that these two initial factors should be recorded after
the concrete is cured and the test set-up is finished.
εri = σri /Er (6.1)
σri = Pri /Ari (6.2)

Pri = K f i2 − f 02 + K T (Ti − T0 ) (6.3)
where:
εri = the strain at a time of i developed inside the rebar;
σri = the rebar stress at a time of i from the cross area of Ari (kN/m2 );
Er = individual rebar modulus provided from the manufacturer (kN/m2 );
Pri = vertical force transferred to the rebar (kN);
6.2 Compressive Static Load Tests 179
Ari = area of the rebar at a time of I (m2 );

K = calibration coefficient of rebar (kN/Hz2 );
K T = temperature compensation factor (kN/°C);
f 0 = initial frequency recorded from rebar (Hz);
f i = frequency recorded at a time of i (Hz);
T0 = initial temperature (°C);
Ti = temperature at time of i (°C).
Theoretically, as shown in Eq. (6.4), the strain of the rebar is equal to the strain of
the steel reinforcement and the concrete strain after the loading is applied because
these three materials act as one element. The concrete cross-sectional axial force
can be determined based on Eqs. (6.1)–(6.4), as illustrated in Eq. (6.5). The force
difference (Pi ) between different concrete cross-sections represents the soil friction
forces, and the friction stress between the concrete pile and soil can be determined
using force difference divided by friction area ( Ash ), as depicted in Eq. (6.6).
εci = εri = εsi (6.4)
E c Aci + E s Asi
Pi = Aci E c εci + Asi E s εsi = Pri
E s Ari
E c Aci + E s Asi 2
= K f i − f 02 + K T (Ti − T0 ) (6.5)
E s Ari
qs = Pi / Ash (6.6)
where:
Pi = concrete cross-sectional axial force.
Aci , E c , εci = concrete section area, modulus and strain at a time of i, respectively.
Asi , E s , εsi = steel section area, modulus and strain at a time of i, respectively.
qs = friction stress between concrete pile and soil (kN/m2 ).
Pi = force difference between two cross-section axial forces (kN).
Ash = column friction area (m2 ).
6.2.3.3 Construction Process
Figure 6.5 displays the preparation of the pile test, eight anchor piles were initially
cast (Fig. 6.3a), and later the test pile was cast. After 28 days of concrete curing,
Fig. 6.5 Construction process
the reaction column was cast. Later, the reaction column, concrete beam, anchoring
strand and test pile were connected, as demonstrated in Fig. 6.3b. After the installation
of hydraulic jacks and dial gauges, the compressive SLT could be performed and,
finally, the results could be obtained.
6.2.4 Test Process
The maintained compressive load tests were selected to perform. As shown in Table
6.1, there were 14, 10 and 16 loading stages for P12, P66 and P105, with the first
loadings of 4000 kN, respectively. The number of unloading stages was half of the
loading stages, and the decrement load was twice the increment load. During each
loading stage, the vertical settlement was recorded at time intervals of 15 min, and
when the accumulated time exceeded to one hour, the time interval changed to 30 min.
For the unloading stages, the time interval was 30 min. Among these recorded values
during loading stages, if the difference settlement value was <0.1 mm, it could be
primarily recognised as a stable condition. If the difference value of 0.1 mm occurred
again, it could be viewed as stable and the next load could be applied.
Table 6.1 Information of test process

Pile label Loading stage Unloading First load (kN) Increment (kN) Decrement
stage (kN)
P12 14 7 4000 2000 4000
P66 10 5 4000 4000 8000
P105 16 8 4000 4000 8000
6.3 Test Results and Discussion 181
6.3 Test Results and Discussion
6.3.1 Settlement–lg Time and Load–Settlement Curves
Figures 6.6, 6.7, and 6.8 display the settlement–lg time of the three tested piles.
As indicated in these figures, there were no points that illustrated a dramatic
turning trend, and all settlement values during each applied load were close, which
demonstrated that these pile foundations were stable during each applied load.
As shown in Fig. 6.6, after the maximum loading of 28,000 kN was applied,
the maximum vertical pile head settlement of P12 was determined as 87.42 mm.
Comparing to P66, as demonstrated in Fig. 6.7, the stable settlement of the P66
was found as 22.92 mm when loads of 28,000 kN was applied and the total head
settlement was small (55.49 mm) when maximum loads of 40,000 kN was applied.
This represents that the capacity of P66 is greater than P12. Also, this phenomenon
also illustrated that increasing the pile length, especially when reaching a harder
bearing stratum, can effectively increase the ultimate bearing capacity of the single
pile.
As shown in Fig. 6.8, when loading of 28,000 kN was applied, the pile head
displacement of P105 was determined as 17.10 mm, which was less than P66
(22.92 mm) and P12 (87.42 mm). Furthermore, when loads of 40,000 kN was applied,
the vertical settlement of P105 was determined as 27.33 mm, which was also less
lg Time (min)
15 150 1500
0
10
20
4000kN
30 6000kN
8000kN
Settlement (mm)
40
10000kN
50 12000kN
14000kN
60 16000kN
18000kN
70
20000kN
80 22000kN
24000kN
90 26000kN
28000kN
100
Fig. 6.6 s–lgt results of P12

lg Time (min)
15 150 1500
0
10
20
Settlement (mm)
30 4000kN
8000kN
12000kN
40
16000kN
20000kN
50 24000kN
28000kN
60 32000kN
36000kN
40000kN
70
Fig. 6.7 s-lgt results of P66
than the settlement of P66 (55.49 mm). This is primarily due to the increment of
the pile diameter and pile length that improves the total capacity of a single pile.
In addition, when the pile was loaded 64,000 kN, the total settlement of P105 was
determined as 45.07 mm as shown in Fig. 6.8, this represents that this pile possesses
greatest load capacity. From another point of view, the design of this pile foundation
is too conservative (45.07 mm < 0.1 * D = 0.1 * 1800 mm = 180 mm).
The load–settlement curves of these tested piles in relation to the pile toe, pile
head and pile deformation are presented in Figs. 6.9, 6.10 and 6.11. As shown in
Fig. 6.9, after the maximum loads were applied, the pile toe displacement of P12
was determined as 74.2 mm, which represents that the compression of concrete is
very small. This is because the bearing stratum being the silty sand (compressible)
as depicted in Fig. 6.1. This also illustrates that the loads have transferred to the pile
toe and the transferred loads are resisted by the bearing stratum.
As shown in Fig. 6.10, the pile toe movement of P66 was determined as 21.73 mm
(ratio between toe and head settlement: 21.73/55.49 = 39%). Two reasons may have
led to this phenomenon, the first one is that the bearing stratum is the very dense
grait; the second reason is that the shaft friction decreases the transferred load (P66
is longer than P12).
lg Time (min)
15 150 1500
0
4000kN
10 8000kN
12000kN
16000kN
20 20000kN
Settlement (mm)
24000kN
28000kN
30 32000kN
36000kN
40000kN
40 44000kN
48000kN
52000kN
50
56000kN
60000kN
64000kN
60
Fig. 6.8 s-lgt results of P105
Fig. 6.9 Load–settlement Load (kN)

curves of P12 0 20000
0
20
40
Settlement (mm)
60
80
100 Pile Head Settlement
Pile Teo Settlement
Pile Deformation
120
Load (kN)
0 10000 20000 30000 40000
0
10
20
Settlement (mm)
30
40
50

Pile Teo Settlement
Pile Deformation
70
Fig. 6.10 Load–settlement curves of P66
Load (kN)
0 10000 20000 30000 40000 50000 60000 70000
0
10
20
Settlement (mm)
30
40

Pile Teo Settlement
Pile Deformation
60
Fig. 6.11 Load–settlement curves of P105

Load (kN)
0 10000 20000 30000 40000 50000 60000 70000
0
10
20
30
Settlement (mm)
40
50
60
70
80 P12
90 P66
P105
100
Fig. 6.12 Load–settlement curves of tested piles
As demonstrated in Fig. 6.11, the pile toe displacement of P105 was very small
(only 1.59 mm). This illustrates that the load did not transfer to the pile toe because
this pile is too long, in other word, this pile is designed very conservative.
The load–settlement curves of these tested piles in relation to the pile head are
shown in Fig. 6.12. It can be found that, after the maximum loadings of 28,000, 40,000
and 64,000 kN were applied, the maximum settlements of P12, P66 and P105 were
determined to be 87.42, 55.49 and 45.07 mm, respectively. This demonstrates that
the pile capacity of P105 is greater than that of the other two piles, and the smallest
capacity is for P12. Further, it was found that, for P105, a linear line was observed
(close to elastic deformation of PL/EA), which represented that the settlement of
P105 was mainly due to compression deformation. Given that the deformation of the
concrete element is the main factor indicating the ultimate capacity of a pile (P105),
the ultimate bearing capacity would not effectively increase if applying base grouting
technology.
6.3.2 Load Transfer Mechanism
The load transferred to the pile cross-section along the pile length can be determined
using Eq. (6.5). Figures 6.13, 6.14, and 6.15 present the load transfer mechanism
of these three tested piles. It can be seen that the transferred load decreased along
the pile length during each applied load, but the gradients were different. This was
because the pile shaft area and soil layers were different.
As shown in Fig. 6.13, the change of the axial force of P12 from 0 to 15 m was
smaller than rest of the part. This was because the layers from 0 to 15 m were mostly
silty clay with soil condition of ‘very soft and soft’. Also, it can be found that below
Load (kN)
0 5000 10000 15000 20000 25000 30000
0 95
90
10 4000kN
85 6000kN
80 8000kN
20
10000kN
75
Elevation (m)
12000kN
Depth (m)
30 70 14000kN
16000kN
65
18000kN
40
60 20000kN
22000kN
55
50 24000kN
50 26000kN
28000kN
60 45
Load (kN)
0 5000 10000 15000 20000 25000 30000 35000 40000 45000
0
5 93
10
15 83
20
25 73
4000kN
30
Elevation (m)
8000kN
Depth (m)
35 63
12000kN
40
45 16000kN
53
50 20000kN
55 24000kN
43
60 28000kN
65 32000kN
70 33
36000kN
75
40000kN
80 23
15 m, the gradients from each layer were similar, this was due to similar soil layers
and conditions being existed (Fig. 6.1, dense fine sand). As shown in Fig. 6.14, the
gradients of P66 from 0 to 12 m was greater than the rest part, this was because the
existence of loose sand and soft clay. As shown in Fig. 6.15, the gradient from 0 to
10 m was large, which illustrates small change of axial force. This was because the
loose to medium dense sand existed in this range. Also, the smallest gradient was
found from 30 to 38 m, this was due to the stiff soil layer existing in this range and
more shaft resistance was provided by this layer.
Load (kN)
0 10000 20000 30000 40000 50000 60000 4000kN
0 94 8000kN
12000kN
10
84 16000kN
20 20000kN
74
24000kN
30
Elevation (m)
28000kN
64
Depth (m)
40 32000kN
54 36000kN
50 40000kN
44 44000kN
60
48000kN
70 34 52000kN
56000kN
80 24
60000kN
64000kN
90 14
For P12 and P66, the bearing capacity included end bearing as demonstrated
in Figs. 6.13 and 6.14, respectively. Further, it can be found that the end bearing
percentage of P12 (9523/28,000 = 34%) was greater than P66 (8117/40,000 =
20.2%), which represented that the increase of pile length can make contribution
to the increase of the shaft resistance and hence reduce the transferred loads to the
bearing stratum. However, for P105, the end bearing was very small. This indicated
that P105 was a shaft-resistant pile (Fig. 6.15). Under this condition, seeking to
improve the ultimate bearing capacity by base grouting would be pointless for P105.
As illustrated in Fig. 6.1, the soil layers around P105 were mostly sandy material;
thus, shaft grouting technology was preferred to increase the total ultimate bearing
capacity. In addition, since no loads were transferred to the pile end, increasing the
pile length would not increase the pile capacity, and the shaft resistance near the pile
toe would not be fully developed.
Table 6.2 provides the proportion of shaft and toe resistance under each working
load. It can be seen that, for P105, a very small end bearing was observed. For P12
and P66, before RW was <80%, the shaft and toe resistance proportions were similar.
Further, when RW was 100%, the shaft resistance proportion of P12 was less than
P66, which was mainly because the pile length of P66 was greater than that of P12.
6.3.3 Shaft Resistance Development and Distribution
Figures 6.16, 6.17, and 6.18 present the load–shaft resistance curves of the three tested
piles in relation to each soil layer. As shown in Fig. 6.16, for P12, it was observed
that, for most layers, when increasing the load, the shaft resistance developed from
zero to a maximum value and was then maintained. Moreover, the shaft resistance
Table 6.2 Proportion of shaft and toe resistance under working load
P105 P66 P12
RW (100%) RE (100%) RS (100%) RW (100%) RE (100%) RS (100%) RW (100%) RE (100%) RS (100%)
6.25 0.00 100.00 10.00 0.00 100.00 14.29 0.00 100.00
12.50 0.00 100.00 20.00 0.19 99.81 21.43 0.00 100.00
18.75 0.00 100.00 30.00 0.39 99.61 28.57 0.63 99.38
25.00 0.13 99.87 40.00 0.66 99.34 35.71 0.83 99.17
31.25 0.21 99.79 50.00 1.80 98.20 42.86 1.15 98.85
37.50 0.27 99.73 60.00 5.18 94.83 50.00 1.46 98.54
43.75 0.30 99.70 70.00 13.32 86.68 57.14 5.84 94.16
50.00 0.33 99.67 80.00 18.70 81.30 64.29 9.25 90.75
56.25 0.53 99.47 90.00 19.56 80.44 71.43 11.84 88.17
62.50 0.75 99.25 100.00 20.29 79.71 78.57 15.76 84.24
68.75 0.92 99.08 85.71 22.18 77.83
75.00 1.02 98.98 92.86 29.07 70.93
81.25 1.19 98.81 100.00 34.01 65.99
87.50 1.26 98.74
93.75 1.47 98.53
100.00 1.80 98.20
Note RW = working load proportion; RE = end bearing percentage; RS = shaft resistance percentage
100
90
80
Shaft resistance (kPa)
70
60
Layer 1
50
Layer 2
40 Layer 3
30 Layer 4
Layer 5
20
Layer 6
10 Layer 7
0 Layer 8
0 5000 10000 15000 20000 25000 30000
Load (kN)
Fig. 6.16 Shaft resistance development of P12

140
120
Layer 1
100
Layer 2
80 Layer 3
Layer 4
60 Layer 5
Layer 6
40 Layer 7
Layer 8
20
Layer 9
Layer 10
0
0 10000 20000 30000 40000
Load (kN)
180
160
Layer 1
140 Layer 2
Layer 3
120
Layer 4
100
Layer 5
80 Layer 6
Layer 7
60
Layer 8
40
Layer 9
20 Layer 10
Layer 11
0
0 10000 20000 30000 40000 50000 60000 70000
Load (kN)
from the upper layers developed first (the shaft resistance from layer 7 and layer 8
did not well developed before loading of 8000 kN was applied). Also, it can be found
that, after loads of 8000 kN were applied, the shaft resistance from layer 6, 7 and
8 increased dramatically, and the shaft resistance of layer 6 and 7 maintained after
loads 22,000 kN were applied, but the resistance was keep going for layer 8, which
represented that the total shaft resistances were almost fully developed.
As illustrated in Fig. 6.17, similar to P12, the shaft resistance of P66 from the upper
layers developed first. The shaft resistance from layer 1, 2 and 3 was fully developed
after 8000 kN, and the resistance from layers 8, 9 and 10 started to increase after
16,000 kN was applied. In addition, it was found that some curves fluctuated. As
for layer 3, the shaft resisting softening was found when loading of 16,000 kN was
applied, and after 24,000 kN, the stress hardening occurred. The reason was that the
pile was compressed vertically and expanded horizontally during load was applied,
and the sandy soil of layer 3 (medium dense) near the pile rearranged the location. The
slippage of the sand particles decreased the development of the shaft resistance, but
later, the soil particles became stable, the shaft resistance developed. Interestingly,
the soil softening and soil hardening occurred again as shown in the figure, the
reason may be related to the slurry wall thickness. Unfortunately, that data is limited,
further research referring to the amount of slurry admixture is required. It is worth
to notice that, from the second softening of layer 3 (loading of 32,000 kN), the shaft
resistance of layer 10 increased dramatically, and from the second hardening of layer
3 (loading of 36,000 kN), the shaft resistance of layer 10 decreased dramatically. This
phenomenon can be explained as ‘Mutual Compensation’.
As demonstrated in Fig. 6.18, it is evidently to see that the shaft resistance of
layer 1, 2 and 3 were fully developed after 4000 kN, and the shaft resistance of
4, 5, 6, 8 were fully developed after 20,000 kN. Later, with the increasing of pile
head loading, the layers of 8, 9, 10 and 11 played a significant role in resisting the
transferred load. Also, it can be seen that the shaft resistance of layer 7 was keep
going to increase (though tress softening occurred which was due to slippage of sand
and gravel), this illustrated that the shaft resistance from the upper layer did not have
to fully developed when the shaft resistance from the lower part started to develop.
Figures 6.19, 6.20, and 6.21 present the shaft resistance distribution along the
pile depth of the tested piles. As shown in Fig. 6.19, before loading of 10,000 kN
was applied, the shaft resistance of P12 was mainly provided by layers 3, 5 and
6. Moreover, with the loads increasing, the shaft resistance generally developed
along the pile length. Similar to the finding obtained from Fig. 6.16, after loading of
22,000 kN was applied, except for the bottom layer (the shaft resistance have kept
increasing), the shaft resistance from the other layers almost fully developed (in each
layer, the distances among vertical lines are very close to each other).
As shown in Fig. 6.20, for the P66, before loading of 16,000 kN was applied, the
shaft resistance was mainly provided from upper layers, the bottom layer provided
small shaft resistance, which was less than 5 kPa. Moreover, if connecting the shaft
resistances with a smooth line, the resistance along the pile illustrated an ‘R’ shape.
Figure 6.20 also presented that the fluctuation curves of layer 3, the shaft resistance
softening and hardening observed.
The shaft resistance distribution of P105 is represented in Fig. 6.21. It can be
found that there was almost no shaft resistance from the bottom layer before 28,000
kN was applied. Also, it can be seen that the layers 1, 2 and 3 were fully developed
firstly and 4, 5, 6 and 8 were developed later because the vertical lines’ distances are
very close. Similar to the findings obtained from Fig. 6.18, the shaft resistance of
layer 7 kept going to develop during the whole loading tests.
Shaft Resistance (kPa)
0 10 20 30 40 50 60 70 80 90 100
0
10
15
4000kN
20
6000kN
8000kN
25
10000kN
12000kN
Pile Depth (m)

30
14000kN
16000kN
35
18000kN
20000kN
40
22000kN
24000kN
45
26000kN
28000kN
50
Fig. 6.19 Shaft resistance distribution along the pile for P12
191
192

0 20 40 60 80 100 120 140
0
10
20
30
40
4000kN
8000kN
Pile Depth (m)

50 12000kN
16000kN
20000kN
60
24000kN
28000kN
70 32000kN
36000kN
40000kN
80
6 Field Tests of Super-Long and Large Diameter Piles
0 20 40 60 80 100 120 140 160 180 200
0
10
4000kN
20
8000kN
12000kN
30 16000kN
20000kN
24000kN
40
28000kN
Pile Depth (m)

32000kN
50 36000kN
40000kN
44000kN
60
48000kN
52000kN
70 56000kN
60000kN
64000kN
80
193
This chapter has investigated the axial behaviour of super-long and large-diameter
piles through performing SLTs. A newly developed reaction system was introduced,
and three piles with various lengths and diameters were tested. Detailed experimental
principle used to determine the transferred load by using sister rebar was provided
with formula derivation. Also, the test process of the super-long and large-diameter
was demonstrated in this chapter.
For investigation of the pile capacity, the traditional presentations of Q-s curves
including the load versus pile head and pile toe settlement as well as settlement-lg time
curves were illustrated. By analysing the load versus depth diagrams, the load transfer
mechanism of these tested piles was researched. Also, the load versus shaft resistance
was provided, which was used for the study of shaft resistance development. In
detail, with load increment, the development of shaft resistance in each soil layer
was researched. Lastly, the shaft resistance distribution of these three super-long and
large-diameter piles were presented.
References
Feng, S. J., Lu, S. F., & Shi, Z. M. (2016). Field investigations of two super-long steel pipe piles in
offshore areas. Marine Georesources & Geotechnology, 34(6), 559–570.
tion of driven pile foundations (FHWA-NHI-16-009) (FHWA-NHI-16-009). Federal Highway
Administration: USA.
Li, S. C., Zhang, Q., Zhang, Q. Q., & Li, L. P. (2016). Field and theoretical study of the response
of super-long bored pile subjected to compressive load. Marine Georesources & Geotechnology,
43(1), 71–78.
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capacity of large diameter bored piles. MATEC Web of Conferences, 22, 1–7.
Xing, H. F., Wu, J., & Luo, Y. (2017). Field tests of large-diameter rock socketed bored piles based
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Chapter 7
Piles Under Ultimate Loads
During the construction of a bored pile, the remove of soils by machines may cause
the soil residue to remain in the drilled hole. This leads to a decrease in the end
resistance capacity of the pile and an increase of pile settlement. Because of these
remaining soil deposits from the base of hole, it has been found that the capacity of
some piles from a mansion was much lower than the designed requirement. It has
also been found that the settlement of these piles was much greater than the treated
piles through performing the SLTs in Taiyuan City, China, (Shi 2005).
For this particular problem, it can be solved by applying the slurry support tech-
nology. However, it is reported that, sometimes, this slurry support admixture layer
(bentonite or polymeric material, soils and water) deceased the friction resistance
between soil layers and concrete pile, which consequently decreased the ultimate
bearing capacity of the piles. It is reported that this admixture layer or composite
layer decreases 30–40% of the pile bearing capacity (Li et al. 2000).
Under condition of the existed admixture layer, the problem can be solved by
applying the post grouting technology. During this process, cement admixture is
pressured down, and the slurry support layer is then forced out (Zhou et al. 2017).
There are intensive researches that focused on the ultimate bearing capacity of these
piles which suffered from compressive, uplift and horizontal forces during SLTs
(Castelli and Wilkins 2004; Nguyen et al. 2012; Sinnreich and Simpson 2013; Patel
et al. 2015). The ultimate bearing capacity of post pile is mostly determined by
interpretation of non-plunging curve result. Other investigations paid attention to
the methods of grouting to find out the best way to increase the bearing capacity
under various subsurface condition (Ho 2003). Some of the researchers take the load
transfer mechanism into account. However, the SLTs of these piles under failure load
are rarely conducted, and the research of these piles’ behaviours is limited.
This chapter provides an understanding of pile failure criteria under SLTs to
enable continuous improvement in construction practices and primary foundation

https://doi.org/10.1007/978-981-33-6183-6_7
198 7 Piles Under Ultimate Loads
design. There is very limited information referring to the behaviour of a pile under
ultimate loading in AS 2159-2009 (Standards Australia 2009). Thus, the objectives
of this chapter are to illustrate the behaviours of piles suffering from ultimate failure
condition.
Three dominant factors can result in pile failure. Commonly in practice, pile failure
occurs because of inadequate rigidity of the concrete. Inadequate concrete vibration
may lead to decrease of the concrete strength. If the vibration is insufficient, the
deficient concrete will crack. If the vibration is too great, the lower part of concrete
will achieve the required strength, but the pile head (which is mainly made of cement,
sand and water without coarse aggregate) will fail to achieve the required strength.
The second common reason for pile failure is that the rigidity of the concrete pile
is overdesigned, or the engineer overestimates the soil rigidity. The pile foundation
element will not break, but the soil fails when the ultimate load is applied. The final
reason for pile failure is that eccentric loading is applied, and the pile is broken by
bending moment.
To obtain the pile behaviours under ultimate loading, three piles were cast and
SLTs were performed. Piles P60, P40 and P12 were loaded until failure occurred
for the aforementioned conditions. Specifically, P60 failed because of inadequate
concrete rigidity, P40 failed because of bending movement and P12 failed because
of the insufficient rigidity of the soil. The other failure conditions depicted in Figs. 7.1
and 7.2—failure of the tension bar system in the reaction pile and platform bearing
failure—are not included because there was no pile soil destruction.
Fig. 7.1 Failure of tension bar system in reaction pile (Handley et al. 2006)
7.2 Subsurface Explorations 199
Fig. 7.2 Platform bearing failures under Kentledge test (Handley et al. 2006)
7.2 Subsurface Explorations
The construction site was located in western Jinan city, Shandong Province, China.
The subsurface exploration was discovered through laboratory and in-situ tests. The
in-situ tests of SPTs and CPTs and laboratory tests of consolidation, direct shear and
triaxial tests were conducted based on the Chinese Standard for Soil Test Method
(GB/T 50123-1999 1999), Standard for Test Methods of Engineering Rock Mass
(GB 50021-2001 2001), respectively. Based on the borehole logs and soil samples,
ten layers were discovered as follows:
1. fill: yellowish-brown: plastic clay with ash concrete and brick crumbs, thickness
ranges from 0.6 to 3.4 m; 2. silty clay: yellow, loess-shaped, saturated, thickness
ranges from 1.2 to 3.2 m; 3. silty clay: yellowish-brown, plastic, thickness ranges
from 0.4 to 4.4 m; 4. clay: grey-brown to beige, plastic, thickness ranges from 1.2 to
2.5 m; 5. clayey silt: it contains ferric oxide, thickness ranges from 0.6 to 1.6 m; 6.
clay, brown: partially contains loess-doll, particle size between 1 and 2 cm, thickness
ranges from 1.2 to 4.1 m; 7. silty clay: yellowish-brown, plastic, contains iron and
manganese oxides, thickness ranges from 3.7 to 6.8 m; 8. clay: yellow to pale brown,
plastic, thickness ranges from 3.4 to 6.2 m; 9. silty clay to clay: pale brownish red,
plastic stiff, thickness ranges from 12.3 to 34.2 m; 10. clay: pale brownish red, very
stiff, contains gravel with particle size between 2 and 3 cm.
Fig. 7.3 Plan view of the tested piles (not to scale)
As shown in Fig. 7.3, three piles with labels of P80, P60 and P40 are next to
building 6# . These piles are in the same diameter and length. Pile 12 which is near
building 3# has a larger diameter and length. Based on the borehole logs from the
study area that shown in Fig. 7.3, it is discovered that the subsurface condition near
building 6# is similar. The soil parameters of the test piles are summarized in Figs. 7.4
and 7.5.
7.3 Designed Ultimate Bearing Capacity
JGJ 94-2008 (Technical Code for Building Pile Foundations) has depicts some of
the methods for determining the ultimate bearing capacity of pile foundation when
designing a pile foundation. The equations used in this project are illustrated in
Eqs. 7.1 and 7.2 which are calculations for the traditional pile foundation and post
grouted pile foundation, respectively. Equation 7.1 represent the ultimate bearing
capacity of single pile without any grouting application hence the total capacity
(Q uk ) is the sum up of shaft capacity (Q sk ) and end capacity (Q pk ). The values of
shaft stress (qsik ) and end stress (q pk ) are relating to types of soils, state of soils (such
as void ratio, liquid index, SPT N-value), types of pile foundation and construction
method. Different to Eq. 7.1, the total capacity of pile as demonstrated in Eq. 7.2
includes the parts where resistances with grouting strengthening effect from shaft
(Q gsk ) and base (Q gpk ). The strengthening coefficients of shaft (βsi ) and base (β p )
are relating to the types of soils.
7.3 Designed Ultimate Bearing Capacity
Fig. 7.4 Subsurface conditions of P80/P60/P40

201
202
Fig. 7.5 Subsurface condition of P12

7 Piles Under Ultimate Loads
7.3 Designed Ultimate Bearing Capacity 203

Q uk = Q sk + Q pk = u qsik li + q pk A p (7.1)

Q uk = Q sk + Q gsk + Q gpk = u qs jk l j + u βsi qsik l gi + β p q pk A p (7.2)
where:
Q uk —Ultimate bearing capacity of single pile.

Q sk —Shaft capacity without grouting strengthening.
Q pk —Base capacity without grouting strengthening.
qsik —Shaft resistance of each soil layers.
q pk —End resistance from the pile toe.
A p —End area from the pile toe.
Q gsk —Shaft capacity with grouting strengthening.
Q gpk —Base capacity with grouting strengthening.
qs jk —Shaft resistance without grouting strengthening.
u—Perimeter of pile.
l j —Thickness of layers within soil strengthening part.
βsi —The strengthening coefficients of shaft.
β p —The strengthening coefficients of base.
The capacity of single piles based on the In-situ method and API method are
also considered in this case study. Different to the JGJ 94-2008, In-situ method
considers the SPT N-value and the API method considers α coefficient which is
relating to the overburden pressure. The ultimate bearing capacity of these four piles
are summarized in Table 7.1. It can be seen that, the total capacity obtained from
Table 7.1 Ultimate bearing capacity of piles

Standards Shaft End Total Designed Reaction Pile label
resistance resistance capacity load weight
(kN)
JGJ 94 4973 804 5777 8400 10,080 P12
(Non-grouting)a
In-situ method 3610 1055 4665
API method 4285 904 5190
JGJ 94 5696 1768 7465
(Grouting) b
JGJ 94 3156 382 3538 5200 6240 P80/P60/P40
(Non-grouting) a
In-situ method 2669 466 3135
API method 3335 203 3538
JGJ 94 3699 877 4577
(Grouting)b
Note a JGJ 94 (non-grouting)
b JGJ 94 (Grouting) represent the ultimate bearing capacity based on Eqs. 7.1 and 7.2, respectively
Table 7.2 Information of tested piles

Pile label Pile Before cutting Pile length Pile diameter Concrete
condition (m) (m) (mm) strength (MPa)
P40 Eccentricity 32 32 − 2 = 30 600 35
P60 Inadequate 30 30 − 0 = 30 600 Upper 2 m: 25
concrete − 30
strength Lower part: 35
P80 Standard pile 32 32 − 2 = 30 600 35
P12 Inadequate 37 37 − 2 = 35 800 50
soil strength
In-situ method is conservative, and the capacity gained from JGJ 94-2008 and API
method are similar. Finally, the designed maximum loads for SLTs are determined as
8400 and 5200 kN for P12 and P80, P60, P40 respectively. For security consideration,
the reaction weight (1.2 times the designed testing loads) are defined to be 10,080
kN (P12) and 6240 kN (P80/P60/P40) for the testing piles.
7.4 Pile Construction
There are 4 piles designed to be tested. As depicted in Table 7.2, after the concrete
were cured, the pile length of P40, P60, P80 and P12 were 32, 30, 32 and 37 m,
respectively (before pile cut). After cutting of 2 m from the pile head as shown in
Fig. 7.6, all three piles (P40, P60, P80) were in the same dimensions of length of
30 m. Except for P12, the diameter of the other three piles was 600 mm. Because
there was no cutting for P60, it was believed that the top 2 m’ concrete did not
achieve the required strength of C35. Note that, after the concrete cured, the strength
of the concrete from the upper pile was discovered from 25 to 35 MPa via conducting
concrete rebound tests by resiliometer.
7.5 Test Setup
The SLTs were conducted based on the Chinese Technical Code for Testing of
Building Foundation Piles (JGJ 106-2014 2014). The setup of the SLTs is provided
in Fig. 7.7. As shown in Fig. 7.8, four hydraulic jacks (QF630T-20) were used to
provide loads to the pile head. The low-speed maintenance methods were used in this
case study, and the applied load was maintained until the rate of axial movement did
not exceed to 0.1 mm. After each load was applied, the vertical movement of the piles
was recorded with intervals of five, 10 and 15 min, and 30 min if the accumulated
time exceeded 1 h.
7.5 Test Setup 205
Fig. 7.6 Percussion of pile head
Fig. 7.7 Compressive SLT
For piles P40, P60 and P80, the loading started at 1,040 kN and progressed with
520 kN increments. The vertical settlements corresponding to each loading were
recorded with four automatic dial gauges. For the piles that achieved the maximum
loading of 5,200 kN, the applied loads decreased with decrements of 1,040 kN, and
the unloading terminated at 0 kN. For pile P12, the loading started at 1,680 kN and
progressed with 840 kN increments.
Fig. 7.8 Equipment of SLT
7.6 Observation of Pile Under Failure Condition
7.6.1 Piles with Inadequate Rigidity
As shown in Fig. 7.9, a concrete crack in P60 occurred when the loading was at
3,640 kN. The crack was discovered at the pile head, and the maximum width of the
crack was up to 30 mm. Through a low-strain integrity test, this pile was found to be
broken at the depth of 1.2 m underground. After digging out the pile and breaking the
concrete from the pile head, it was found that the reinforcement was bent outside at a
depth of 0.9–1.2 m, as shown in Fig. 7.10. This occurred because, after the concrete
was damaged, the reinforcement tried to resist the vertical loading, and, for a second,
the reinforcement was bent. It could not be bent inside the pile because the concrete
core resisted this deflection.
7.6.2 Piles Suffering from Eccentric Loads
As shown in Fig. 7.11b, pile P40 was tested with eccentricity of 200 mm. After a load
of 4,680 kN was applied, the pile seemed to successfully resist the loading; however,
the concrete actually crushed at the compression zone, and the reinforcement at this
zone (Fig. 7.11a) tried to resist this compressive load, yet failed, and the orientation
was outside. At the same time, the concrete ruptured in the tensile side, but the
reinforcement in this side provided tensile resistance for a couple of minutes and
failed later. There was no reinforcement hump discovered in this zone (Fig. 7.10c).
7.6 Observation of Pile Under Failure Condition 207
20–30 mm
Fig. 7.9 Concrete crack from pile head
Fig. 7.10 Deformed

reinforcement
Eccentricity
200 mm
(a) (b) (c)
Fig. 7.11 Pile suffering from eccentric loads. a Bent reinforcement of left part b Pile suffering
from eccentricity c Fractured concrete of right part
7.6.3 Failure from Inadequate Soil Rigidity
For pile P12 that suffered from plunging failure, in which concrete was not damaged.
After a load of 8,400 kN applied, nothing occurred until 115 min, when the pile was
suddenly driven into the ground. This phenomenon can be primarily explained by
the fact that the shaft resistance was fully developed, and, after 115 min, the loading
transferred to the pile tip. As a result of the inadequate stiffness of the clay-bearing
stratum, the pile was plunged into the ground. As shown in Fig. 7.12, cracks in the
ground were discovered up to 100 mm.
100 mm
Fig. 7.12 Plunging failure of pile

7.7 Test Results 209
7.7 Test Results
7.7.1 Pile with Achieved Design Requirements
The results of pile P80 are provided in Figs. 7.13, 7.14 and 7.15. The interpretation
results based on the double tangent method, Chin’s method and offset method are
provided in Fig. 7.16a–c, respectively.
From the Q-s curve (Fig. 7.13), the ultimate loading was conservatively deter-
mined as 4,680 kN, with a corresponding maximum settlement of 4.9 mm. For
some local standards, the ultimate bearing capacity could be determined as 5,200 kN
because the corresponding settlement of 7.88 mm was relatively small. According
to the Chinese Standard JGJ 106-2014 (2014), Clause 4.4.2.1, through analysing
thes-lgt results at each loading, the capacity of the pile could be determined where
there was a deep downwards curve occurring at the end of s-lgt curve. As shown in
Fig. 7.14, there were two evident downward trends from the loading stages of 4,680
Fig. 7.13 Q-s curve of P80 Compressive Load (kN)

0 2000 4000 6000
0
3
Settlement (mm)
9
lg Time (min)
1 10 100 1000
0
4
Settlement (mm)
8
1040kN
1560kN
10 2080kN
2600kN
3120kN
12
3640kN
4160kN
14 4680kN
5200kN
Fig. 7.14 s-lgt curve P80
and 5,200 kN, but the settlement remained stable in the end, without an extreme
settlement increasing. Hence, the capacity of the pile was determined as 5,200 kN,
or 4,680 kN for the conservative consideration. Based on the s-lgQ curve, the ultimate
loading was determined as 4,680 kN, with difficulty (Fig. 7.15).
As stated in AASHTO (2002) and FHWA (1992), the double tangent method is
more commonly used for drilled shafts. Based on plotting two tangent lines, the
intersecting point that presents the ultimate bearing capacity of P80 was determined
to be 3,900 kN, as shown in Fig. 7.16a. Through Q-s curve, plotting elastic line, and
offset this line based on Eq. 2.38. As shown in Fig. 7.16b, there was no intersect
between the offset line and curve from the loading stages, which indicated that the
ultimate bearing capacity of P80 was over the maximum applied load of 5,200 kN.
When plotting the load/settlement versus settlement, the result indicated a line with
a gradient of 0.0002. Based on Chin’s method, the ultimate bearing capacity of P80
was determined to be 5,000 kN, as shown in Fig. 7.16c.
Compressive Load-lgQ (kN)

1 10 100 1000 10000
0
3
Settlement (mm)
Fig. 7.15 s-lgQ curve P80
7.7.2 Failure from Inadequate Concrete Strength
The Q-s curve, s-lgt curve and s-lgQ curve of P60 are provided in Figs. 7.7, 7.17,
7.18 and 7.19. As shown in Fig. 7.17, a small settlement 14.13 mm was discovered
from a loading of 3,640 kN, and there was an extremely increasing (from 14.13 to
83.2 mm) discovered when the next loading applied, the ultimate bearing capacity
was determined to be 3,640 kN for P60 based on the Q-s curve.
As shown in Fig. 7.18, there was a dramatic decreasing trend when the loading
of 4,160 kN was applied, which showed increments in settlement with increasing
time, this represented that the failure criteria of the soil occurred. So, the ultimate
bearing capacity of P60 was then determined to be 3,640 kN. This figure also provides
valuable information illustrating the pile failure behaviour. As shown in Fig. 7.18,
when loads of 3,640 kN were applied, the concrete from the upper part of pile
foundation broken and resulted in the settlement increment. Later, under such vertical
loads, the pile foundation could still resist the applied vertical load and thus the
settlement data remained stable. In practice, when finding a curve similar to the
3,640 kN-curve, the pile is expected to be broken, checking the pile integrity is
indispensable, thus low strain detection test is strongly recommended to perform.
As shown in Fig. 7.19, a point with corresponding load of 3,640 kN was found
to be the ultimate load because, after this point, the settlement of the pile head
decreased dramatically (the settlement changed from 14.13 to 83.2 mm with small
load increment of 520 kN).
Compressive Load (kN) Compressive Load (kN)

0 2000 4000 6000 0 2000 4000 6000
0 0
1
2
2
4
3
Settlement (mm)
Settlement (mm)
4 6
5 8
6
10
7
12
8 P80
Elastic Line
Offset Line
9 14
(a) (b)
0.0016
0.0014
0.0012
0.001
0.0008
0.0006
0.0004
0.0002
0
0 2 4 6 8 10
Settlement (mm)
(c)
Fig. 7.16 Interpretations a Double tangent method b Davisson’s offset method c Chin’s method

0 2000 4000 6000
0
10
20
30
Settlement (mm)
40
50
60
70
80
90
Fig. 7.17 Q-s curve of P60
Interpretations of the Q-s curve based on the double tangent method, Davisson’s
offset method and Chin’s method are provided in Fig. 7.20a–c. As shown in Fig. 7.20a,
the intersection of the two tangent lines was determined and the corresponding load
was found to be 3,600 kN. For Davisson’s offset method, using the concrete modulus
of 31,500 N/mm2 to determine the elastic line (PL/EA). As shown in Fig. 7.20b, the
ultimate capacity was determined to be 3,600 kN. However, because the concrete
modulus from the upper part was 30,000 N/mm2 , the actual capacity should be
slightly over 3,600 kN. By using the data obtained during the SLT before failure
occurred, and then plotting the settlement/load versus settlement, a gradient of 0.0003
was discovered, as shown in Fig. 7.20c. Finally, the bearing capacity of P60 was
determined to be 3,333 kN.
lg (Time)
1 10 100 1000
0
10
20
30
Settlement (mm)
40
50
1040kN
60 1560kN
2080kN
70 2600kN
3120kN
80 3640kN
4160kN
90
Fig. 7.18 s-lgt curve of P60
7.7.3 Failure from Eccentricity
For the pile that suffered from eccentric loading, the typical results presentation is
provided in Figs. 7.21, 7.22 and 7.23. As shown in Fig. 7.21, after the loading of
4,680 kN was applied, a huge increase in settlement was discovered. The ultimate
capacity was then determined to be 4,160 kN. As shown in Fig. 7.22, after loading
of 4,160 kN was applied, the s-lgt curve demonstrated that the soil settlement was
relatively stable with time increasing; however, after the next loading was applied
(4,680 kN), there was an extreme increase in settlement, which represented the failure
condition occurring. The ultimate bearing capacity was then determined as 4,160
kN. It is worth noting that there are two ‘drops’ discovered under loads of 4680 kN.
The first settlement drop is referring to the crush of concrete from the compression
zone (around 14 mm: 6 mm to 20 mm settlement change); the second drop relates
to the failure of the pile-soil system. As shown in Fig. 7.23, beyond the point of
4,160 kN loading being applied, a dramatic downwards trend was discovered, which

1 10 100 1000 10000
0
10
20
30
Settlement (mm)
40
50
60
70
80
90
Fig. 7.19 s-lgQ curve of P60
represented the ultimate condition. The ultimate load was then determined as 4,160
kN.
The interpretations based on the double tangent method, Davisson’s method and
Chin’s method are illustrated in Fig. 7.24a–c, respectively. The intersection of two
tangent lines was discovered from Fig. 7.24a, and the ultimate bearing capacity of
P40 was then discovered to be 4,190 kN. As illustrated in Fig. 7.24b, the intersection
between the offset line and Q-s curve was determined, and the ultimate capacity
of P40 was obtained as 4,250 kN. As shown in Fig. 7.24c, Chin’s method was not
applicable to the piles suffering from failure caused by eccentric loading.
7.7.4 Plunging Failure
The typical presentations are provided in Figs. 7.25, 7.26 and 7.27. For pile P12,
the capacity was determined to be 7,560 kN, as shown in Fig. 7.25, because, after
this 7,560 kN point, the Q-s curve indicated a dramatic downwards trend, which
illustrated the failure condition. As shown in Fig. 7.26, under loading of 7,560 kN,
the pile was relatively unstable, but failed when 8,400 kN was applied. The ultimate
capacity of this pile was determined to be 7,560 kN, based on the s-lgt curve. In
addition, from the s-lgQ curve, the ultimate bearing capacity was determined to be
7,560 kN (Fig. 7.27).
The results based on the double tangent method, Davisson’s method and Chin’s
method are presented in Fig. 7.28a–c, respectively. As shown in Fig. 7.28a, the
intersection between two tangent lines was found, and the ultimate bearing capacity

0 2000 4000 6000 0 2000 4000 6000
0 0
10 10
20 20
30 30
Settlement (mm)
Settlement (mm)
40 40
50 50
60 60
70 70
P60
80 80 Elastic Line
Offset Line
90 90
(a) (b)
0.0045
0.004 y = 0.0003x + 0.0002
R² = 0.9987
0.0035
0.003
0.0025
0.002
0.0015
0.001
0.0005
0
0 5 10 15
Settlement (mm)
(c)
was determined to be 7,560 kN. For Davisson’s method, Eq. 2.39 was required instead
of Eq. 2.38 to determine the offset line because this was a large pile with an 800 mm
diameter. The intersection between the offset line and Q-s curve was discovered
(Fig. 7.28b), and the capacity was determined to be 7,900 kN. By plotting the data
obtained from the tested pile before failure occurred, based on Chin’s method, a
function with a slope of 0.001 was determined (Fig. 7.28c) and the ultimate bearing

0 1000 2000 3000 4000 5000
0
10
15
Settlement (mm)
20
25
30
35
40
45
50
Fig. 7.22 s-lgt curve of P40 lg Time (min)

1 10 100 1000
0
10
15
Settlement (mm)
20
25
30
1040kN
1560kN
35
2080kN
2600kN
40 3120kN
3640kN
45 4160kN
4680kN
50

1 10 100 1000 10000
0
5
10
15
Settlement (mm)
20
25
30
35
40
45
50
Fig. 7.23 s-lgQ curve of P40
capacity was obtained, which was equal to 1/0.0001 = 10,000 kN. A summary of
these pile capacities is provided in Table 7.3.
As shown in Table 7.3. It can be seen that the ultimate bearing capacity obtained
from traditional methods (Q-s, s-lgt and s-lgQ curves) and interpretations (double
tangent, Davission’s and Chin’s) are generally close. For the standard pile, comparing
with the ultimate bearing capacity of grouted pile as depicted in Table 7.1 (4,577 kN),
the tested pile’s capacity is overall 4,777 kN, which illustrates the design method
provided by JGJ 94-2014 (2014) (Grouting) is ideal. It should be noted that, in this
chapter, the diagram based on the Chin’s method is determined using the test data
before failure occurred (the maximum applied loads with corresponding settlement
are not used in the plotted Chin’s diagram). The reason to do this is to check if the
Chin’s method can provide appropriate results under non-failure condition or if this
method can be used for capacity determination under Proof Tests (PTs). As shown
in Table 7.3, the capacity obtained by Chin’s method is close to the results acquired
from the other methods, which confirms that the Chin’s method can be appropriately
used for pile capacity prediction when the PTs being performed. Also, it found that
the Chin’s method cannot be used when pile suffered from eccentricity.

0 2000 4000 6000 0 2000 4000 6000
0 0
5 5
10 10
15 15
Settlement (mm)
Settlement (mm)
20 20
25 25
30 30
35 35
40 40
P40
45 45 Elastic Line
Offset Line
50 50
(a) (b)
0.002
0.0018
0.0016
0.0014
0.0012 y = -4E-05x + 0.0015

R² = 0.1502
0.001
0 2 4 6 8
Settlement (mm)
(c)

0 2000 4000 6000 8000 10000
0
10
20
30
Settlement (mm)
40
50
60
70
80
90
100
Fig. 7.26 s-lgt curve of P12 lg Time (min)

1 10 100 1000
0
10
20
30
Settlement (mm)
40
50
60 1680kN
2520kN
70 3360kN
4200kN
80 5040kN
5880kN
6720kN
90
7560kN
8400kN
100

1 10 100 1000 10000
0
10
20
30
Settlement (mm)
40
50
60
70
80
90
100
Fig. 7.27 s-lgQ curves of P12
This chapter has presented a pile investigation for performing compressive SLTs with
the ultimate load applied. To determine the pile behaviour under ultimate loading,
four piles were cast. Three types of tests that lead to failure were considered. The
first test involved loading the pile to failure caused by inadequate concrete strength,
the second test involved loading the pile to failure because of eccentricity and the
final test involved loading the pile until failure because of the soil being soft (or
the pile being overdesigned for its capacity). The behaviours of these three piles
were compared with a pile that achieved the design. In this chapter, after the conven-
tional result presentation as well as different data interpretations being provided,
the pile capacity of these tested piles was summarised. To check whether the Chin’s
method can provide appropriate results under non-failure condition or to verify if this
method can be used for capacity determination under PTs, the interpreted diagrams
by Chin’s method was based on the obtained test data before failure occurred (the
maximum applied loads with corresponding settlement are not used in the plotted
Chin’s diagram).

0 5000 10000 0 5000 10000
0 0
10 10
20 20
30 30
Settlement (mm)
Settlement (mm)
40 40
50 50
60 60
70 70
80 80
P12
90 90 Elastic Line
Offset Line
100 100
(a) (b)
0.012
0.01
0.008
y = 0.0001x + 0.0005
0.006
R² = 0.9991
0.004
0.002
0
0 20 40 60 80 100
Settlement (mm)
(c)
References 223
Table 7.3 Summarisation of the pile capacity

Pile Pile condition Q-s (kN) s-lgt s-lgQ Double Davisson’s Chin’s
label (kN) (kN) tangent (kN) (kN)
(kN)
P80 Standard pile 4,680 5,200 4,680 3,900 >5,200 5,000
P60 Inadequate 3,640 3,640 3,640 3,600 >3,600 3,333.3
concrete
strength
P40 Eccentricity 4,160 4,160 4,160 4,190 4,250 N/A
P12 Inadequate 7,560 7,560 7,560 7,560 7,900 10,000
soil strength
References
American Association of State Highway and Transportation Officials (AASHTO). (2002). Standard
Specifications for Highway Bridges, Washington, D. C.
Castelli, R. J., & Wilkins, E. (2004). Osterberg load cell test results on base grouted bored piles
in Bangladesh. In GeoSupport 2004: drilled shafts, micropiling, deep mixing, remedial methods,
and specialty foundation systems (pp. 587–602).
Handley, B., Ball, J., Bell, A., & Suckling, T. (2006). Federation of piling specialists handbook on
pile load testing. Forum Court: FPS.
Republic of China.
Ministry of Construction of the People’s Republic of China. (2013). Standard for test methods of
Republic of China.
Republic of China.
Nguyen, V. L., Nie, L., & Zhang, M. (2012). Method cement post-grouting to increase the load
capacity for bored pile. Research Journal of Applied Sciences, Engineering and Technology,
5(19), 4727–4732.
Patel, D., Glover, S., Chew, J., & Austin, J. (2015). The pinnacle-design and construction of large
Sinnreich, J., & Simpson, R. C. (2013). Base grouting case studies including full scale comparative
load testing. In Seventh International Conference on Case Histories in Geotechnical Engineering.
No. 2.16, pp. 1–8.
Standards Australia. (2009). Piling—design and installation (AS 2159) (pp. 1–97).
Zhou, J. L., Zhang, X., Jiang, H. S., Lyu, C., & Oh, E. (2017). Static and dynamic load tests of shaft
and base grouted concrete piles. Advances in Civil Engineering, 2017, 1–11.
Chapter 8
Capacity and Settlement Analysis
This chapter presents the results discussion for the post grouted piles, displacement
and non-displacement piles, composite piles and piles tested until ultimate load, with
consideration of the state-of-art construction technology.
For the grouted piles, the behaviours of piles under various grouting technologies
are compared, and the ultimate compressive and uplift bearing capacity is determined
though SLTs and dynamic load tests. These traditional results are presented and
compared with the extrapolated results. Further, the static and dynamic results are
compared and discussed in reference to the ultimate bearing capacity. In addition,
the method used for displacement prediction under compressive and uplift loads is
provided.
After the investigation of post grouted bored piles, the investigations of precast
piles are presented in Sect. 8.3, which include small and large displacement piles and
non-displacement piles. For the displacement piles, the pile behaviours driven from
different locations are discussed. The behaviours of small and large displacement
piles with various lengths are researched with consideration of soil layer analysis.
Further, the behaviours of non-displacement piles suffering from uplift and horizontal
loading are discussed, and the empirical formula for pile displacement prediction are
provided.
This chapter also presents a discussion of the composite piles. For the piles with
FRP confinement, the capacity obtained from various interpretation methods is deter-
mined and compared, and the improved capacity caused by FRP confinement is
examined. For the FRP inside reinforced pile, the horizontal deflection of this pile is
compared with traditional pile (steel bar reinforcement). The discussion of composite
pile considering the construction process (excavation depth and strut application) is
also illustrated.

https://doi.org/10.1007/978-981-33-6183-6_8
226 8 Capacity and Settlement Analysis
Finally, the behaviours of piles applied with ultimate loading are discussed. The
result presentations are compared and the interpretation methods are discussed.
Moreover, the behaviour of piles with three types of failure conditions are detailed.
After working through the material in this chapter, the reader can analyse the
settlement and capacity of the super-long and large diameter pile by themselves
using the same procedures. In detail, for the capacity analysis of these piles, the
interpretation methods in Sect. 8.2.1.1 should be used, the readers can also base on
the methods that are illustrated in Chaps. 2 or 3. For the settlement prediction and
empirical formula determination, find the method in Sect. 8.2.2.1.
8.2 Post Grouted Concrete Piles
8.2.1 Capacity Discussion
8.2.1.1 Compressive Static Load Tests
The load settlement results of three post grouted piles (P51, P121 and P126) obtained
from compressive SLTs illustrated the settlement behaviours under increment and
decrement loading. It was found that the final settlements of P51 (no grouting pile)
and P121 (base grouting pile) were 8.11 and 7.36 mm, respectively. This represented
a small decrement of total settlement after using base grouting technology, which is
highly because the end bearing stratum being the diorite. In other words, the base
grouting may effectively decrease the total settlement for a pile that reaches the
stratum such as loose sands. P126 showed a small total settlement of 2.8 mm. This
demonstrated that the shaft grouting would decrease the total settlement of the pile
foundation.
Presenting the load settlement data into logarithmic scale (s-lgQ) as illustrated
in Fig. 3.8. Through s-lgQ curvatures, the ultimate compressive bearing capacities
of the piles could be determined, albeit with difficulty, to be 6,800, 7,100 and 8,000
kN, respectively. This is because the tests were PTs, which aimed to ensure the pile
settlement was acceptable under the maximum loading (equal to twice the design
load). The non-failure settlements led to difficulty in determining the capacity from
the s-lg Q curves. Thus, determination of the ultimate pile capacity had to be based
on interpretation.
Similar to the analysis of s-lgQ curves, as shown in Fig. 3.15 (DeBeer’s method
interpretation), the ultimate bearing capacities of P51, P121 and P126 obtained from
the plotted lgs–lgQ curves were determined, with difficulty, to be 7,600, 7,900 and
8,300 kN, respectively. For Davisson’s offset limit method, the intersection between
the load–settlement curve and offset line could not be found as shown in Fig. 3.16,
which illustrated that the ultimate bearing capacity was over the maximum applied
loads. Thus, more comprehensive analysis was still needed. Another common method
to plot the failure criteria of a drilled pile is the double tangent method, emphasised
8.2 Post Grouted Concrete Piles 227
by AASHTO (2002) and FHWA (1992) (Samtani and Nowatzki 2006b). As shown
in Figs. 3.12, 3.13 and 3.14, through finding two tangent lines from the loading
curve, the intersection represents the ultimate bearing capacity and corresponding
settlements. The capacities of P51, P121 and P126 were determined to be 7,560,
7,700 and 8,250 kN, respectively. The double tangent method and DeBeer’s method
both aim to find a point to represent a ‘turning trend’ through two tangent lines and
logarithmic scales; thus, the settlements determined from these two methods were
close to each other, as shown in Table 3.4.
As shown in Fig. 3.17, the slopes of these three piles were determined to be 1E-4,
9E-5 and 8E-5, and the inverse of these slopes gave the ultimate capacity. Thus, the
capacity of piles P51, P121 and P126 were determined to be 10,000, 11,111 and
12,500 kN. It was found that Chin’s method provided the greatest values, compared
with the other methods. All results obtained from the different interpretation methods
demonstrated that the base grouting and base-and-shaft grouting improved the ulti-
mate bearing capacity of the piles. For instance, based on double tangent method,
the capacity of the base-and-shaft grouting pile (P126) was 8,250 kN, and the base
grouting pile (P121) was 7,700 kN. These two values were greater than the non-
grouted pile (P51), which had a capacity of 7,560 kN. This was because the grouting
technique increased the shaft and base area to resist the compressive working loads.
Further, the results obtained from Chin’s method and the dynamic methods were
relatively close to each other.
8.2.1.2 Uplift Static Load Tests
The load displacement behaviours of piles suffering from uplift loading are shown
in Fig. 3.18. It can be seen that, during the loading stages from 0 to 3,000 kN, the
displacement between two piles were similar, and these two increment load lines were
approximately linear. By plotting these data into s-lgQ curves, no evident ‘turning
point’ was found, which aligned with the load settlement results (approximately
linear from the loading stages). Through analysing the s-lgt diagrams, as shown in
Figs. 3.20 and 3.21, it was found that the upwards displacement was stable overall
with time before the next loading was applied. Further, there was no evident increase
in upwards displacement between the two loading stages.
As depicted in Fig. 3.22, the interpretation of uplift SLTs through the offset method
gave the ultimate bearing capacity of P15 and P16 to be 1,500 and 1,700 kN, respec-
tively. However, as discussed above, all result analyses of the Q-s, s-lgQ and s-lgt
curves indicated that the ultimate bearing capacity was greater than the maximum
loading of 3,000 kN. Hence, the ultimate load capacity or failure load obtained
through the offset method will underestimate the real uplift ability of the pile; thus,
the recommended uplift design load (stated by FHWA NHI-16-009), which is equal
to half to two-thirds of the failure load, will be conservative. Another interpreta-
tion for the uplift loaded tests was using the modified Mazurkiewicz method. As
illustrated in Figs. 3.23 and 3.24, two functions were determined, and, when x =
0, the ultimate uplift bearing capacities of P15 and P16 were 6,299 and 5,442 kN,
respectively. This indicated that the base-and-shaft grouted technology increased the
ultimate uplift bearing capacity.
8.2.1.3 Dynamic Load Tests
Through dynamic load tests, the shaft resistances along the pile and the end bearing
of the pile could be determined. As provided in Fig. 3.27, the shaft resistance of P121
along the pile length was not uniformly distributed compared with P126 as depicted
in Fig. 3.28. This was because the shaft grouting material which has mixed with
soil layers resulting in the shaft around the pile demonstrated a relatively uniform
property. From the load transfer mechanism diagrams of P121 and P126, it was found
that the transferred loads decreased with depth.
As illustrated in Table 3.6, the ultimate bearing capacities of P121 and P126 were
determined to be 9,071 and 9,851.5 kN, respectively. Further, the shaft resistances
of P121 and P126 were determined to be 4,217.3 and 4,850.0 kN. This represented
that the base-and-shaft grouting increased the total pile capacity (9,071 kN < 9,851.5
kN) and shaft resistance (4,217.3 kN < 4,850.0 kN). The phenomenon was caused
by the post grouting technique, which increased the total friction area between the
concrete surface and soil layers. It can also be seen that the grouting technique could
also improve the end bearing capacity of piles (9,851.5–4,850.0 > 9,071–4,217.3).
Compared with shaft resistance obtained from dynamic load tests and interpretation
by modified Mazurkiewicz’s method, this interpretation overestimated the ultimate
uplift bearing capacity (4,850 kN < 6,299 kN).
From the ultimate compressive SLT results, the results indicated that: (1) the base-
and-shaft grouting pile increased about 9.82% of its capacity without any grouting;
(2) the base grouting pile increased approximately 2.89% of its capacity without
grouting; (3) the capacity of the shaft-and-base grouted pile increased by 6.1% of
the base grouting pile, and this value was close to the results obtained from the
dynamic load tests (8.6%). By comparing the ultimate uplift SLT results, it could be
determined that there was a 15.7% increment in the ultimate pile capacity after using
the base-and-shaft grouting technology. By comparing the shaft resistance capacities
of the base grouted piles and base-and-shaft grouted piles obtained from the dynamic
load test, a 15.0% increment was demonstrated.
8.2.2 Settlement Discussion
8.2.2.1 Compressive Loaded Piles
The maintained SLTs could provide information about which loading stage was
unstable. As shown in Fig. 8.1, during each loading stage, the soil foundation system
was unstable within a duration of 30 min, and, later, the soil foundation system
became stable with increasing time, during which the load was maintained. It can also
Time (min)
0 500 1000 1500 2000
0
Loading Stage 1
Loading Stage 2
2
Loading Stage 3
Loading Stage 4
4
Loading Stage 5
Settlement (mm)
Loading Stage 6
6
Loading Stage 7
Loading Stage 8
8
Loading Stage 9
Unloading Stage 1
10
Unloading Stage 2
Unloading Stage 3
12
Unloading Stage 4
Unloading Stage 5
14
Fig. 8.1 Settlement curves versus maintained time of P51
be seen that, after the loading of Stage 7 was applied, huge settlement was discovered,
and the soil foundation system was unstable after a relatively long duration. Thus, the
loading of Stage 7 was found to be the critical loading. Similarly, from Figs. 8.2 and
8.3, the critical loading stages of P121 and P126 were determined to be loading Stage
9. From the Figs. 8.1, 8.2 and 8.3, it can also be seen that, during each unloading stage,
the soil foundation systems were stable with maintained loading applied. Further,
by comparing the final settlements of these three piles, the results indicated that the
settlement of the base-and-shaft grouted pile was less than the base grouted pile, and
the settlements of these two piles were both less than the traditional pile without
grouting application.
Define the settlement ratio Rs = St /s(Fi ) . Under a certain compressive loading,
the settlement at a certain time (St ) divided by the settlement (s(Fi ) ) which represents
the final settlement at the end of this loading will be less than 1, but over 0. The value
Rs = 1 illustrates that the soil is stable. For example, for the pile P126, from a
loading stage of 7,200 kN, before a time of 120 min, St /s(Fi ) is less than 1. After the
time of 120 min, St /s(Fi ) = 1, the next loading stage will be applied soon because the
settlement is stable at this 7,200 kN load. By plotting the time versus the settlement
ratio, Rs , which is obtained from various loading stages, scatter diagrams of P51,
P121 and P126 can be obtained, as shown in Figs. 8.4, 8.5 and 8.6. By using the
non-linear regression, a settlement equation is proposed, as illustrated in Eq. 8.1:
tn
St = s(Fi ) × (8.1)
mn + t n
Time (min)
0 500 1000 1500 2000
0
2 Loading Stage 1
Loading Stage 2
4 Loading Stage 3
Loading Stage 4
Settlement (mm)
6 Loading Stage 5
Loading Stage 6
8 Loading Stage 7
Loading Stage 8
10
Loading Stage 9
Unloading Stage 1
12
Unloading Stage 2
Unloading Stage 3
14
Unloading Stage 4
Unloading Stage 5
16
Time (min)
0 500 1000 1500 2000
0
4 Loading Stage 1
Settlement (mm)
Loading Stage 2
Loading Stage 3
6 Loading Stage 4
Loading Stage 5
Loading Stage 6
8
Loading Stage 7
Loading Stage 8
10 Loading Stage 9
Unloading Stage 1
Unloading Stage 2
12 Unloading Stage 3
Unloading Stage 4
Unloading Stage 5
14

Fig. 8.4 Non-linear regression of P51 (loading stage)

where:
St = Settlement at a certain loading stage at a certain time in mm
s(Fi ) = Settlement at the end of a certain loading stage in mm
m and n = Settlement coefficients obtained from non-linear regression curve
(Table 8.1).
Comparing the scatter diagrams of P51 with P126, P126 shows better results
because most points assemble on the regression line, which represents that the shaft-
and-base grouting technology reinforced the soil layers around the pile foundation
and made the soil layers stable. Comparing P121 with P126 indicates that the points
obtained from the base grouted pile (P121) were dispersed, which illustrates that
the shaft grouting technique reinforced the shaft soil layers. However, comparing
the results between P51 and P121 indicates a controversial outcome. Although the
locations among these three piles were close with each other, more boreholes were
required to explain the mentioned conflict. This phenomenon could also be caused by
Table 8.1 Settlement coefficients of compressive loaded bored pile

Type of pile Compressive loading Compressive unloading
m n R2 m n R2
Traditional pile 0.1872 0.6795 0.9908 0.8265 4.814 0.9996
Base grouted pile 0.1127 0.5618 0.947 0.3928 3.268 0.9999
Base-and-shaft grouted pile 0.3426 0.9395 0.9921 0.5395 3.942 0.9999
Fig. 8.7 Non-linear regression of P51 (unloading)
the construction process associated with the disturbance of soil layers. For example,
the vibrations during grouting may cause the soil to be disturbed.
Equation 8.1 was only used to predict the settlement from loading stages. In
order to obtain the empirical equation which was used to predict the settlement
during unloading stages, define the settlement ratio Rs = St /s(in) . Under a certain
compressive unloading, the settlement at a certain time (St ) divided by the settlement
(s(in) ) which represents the initial settlement at the beginning of this unloading will
be less than 1, but over 0. The value Rs < 1 illustrates that the soil is relatively stable.
By plotting the time versus the settlement ratio, Rs , which is obtained from various
unloading stages, scatter diagrams of P51, P121 and P126 can be obtained, as shown
in Figs. 8.7, 8.8 and 8.9. By using non-linear regression, a settlement equation is
proposed, as illustrated in Eq. 8.2:
tn
St = s(I n) × (8.2)
mn + tn
where:
St = Settlement at a certain loading stage at a certain time in mm
s(I n) = Settlement at the beginning of a certain loading stage in mm
m and n = Settlement coefficients obtained from non-linear regression curve
(Table 8.1).
These two empirical equations can be used to predict the settlement under a certain
loading at a certain time, while the coefficients of m and n can be determined from
Table 8.1. Equation 8.1 is similar to the equations proposed by Yang et al. (2012),

Fig. 8.10 Non-linear

regression of single pile
(Yang et al. 2012)
with different coefficient values of m and n. As shown in Fig. 8.10, the scatters
assembled very close to each other. This was because the tests loaded to a maximum
of 200 kPa, where loading was relatively small, and the record frequency (number
of times) are lesser under the maintained loads. In contrast, in the case of P51, P121
and P126, where loading was up to 9,000 kN (17,914 kPa), they required much more
time to allow the soil settlement to be stable, especially from the beginning of the
loading (five to 60 min in each loading stage).
Based on the proposed equations of Eqs. 8.1 and 8.2, the predicted settlement and
test settlement of P51, P121 and P126 from the loading stages and unloading stages
are provided in Figs. 8.11, 8.12, 8.13, 8.14, 8.15 and 8.16. It can be seen that these
two equations provided accurate computed results.
Equation 8.1 was proposed based on the settlement data from time five to 120 min
(the settlement difference was less than 0.1 mm when the time was beyond 120 min;
thus, these data were sufficient to determine the fitting line). As shown in Table 8.2,
for P126, it was found that, beyond a time of 120 min, the predicted settlements at a
time of 150 and 180 min were very close to the test settlements. For another example,
for P51, from the loading stage of 5,400 kN, when the time was beyond 120 min,
the predicted settlement was 5.312 mm, which was very close to the collected test
settlement value of 5.395 mm. Similarly, Eq. 8.2 could also predict the settlement
value beyond a time of 95 min.
Time (min)
0 20 40 60 80 100 120 140
0 1800kN Field Test
1800kN Computed
Settlement
2700kN Field Test
2
2700kN Computed
Settlement
3600kN Field Test
4 3600kN Computed
Settlement
Settlement (mm)
4500kN Field Test
6 4500kN Computed
Settlement
5400kN Field Test
5400kN Computed
8 Settlement
6300kN Field Test
6300kN Computed
Settlement
10 7200kN Field Test
7200kN Computed
Settlement
8100kN Field Test
12
8100kN Computed
Settlement
9000kN Field Test
14
Fig. 8.11 Test and computed settlements from loading stages of P51
Time (min)
0 50 100 150 200
7
8
Settlement (mm)
7200kN Field Test

9
7200kN Computed
Settlement
5400kN Field Test
5400kN Computed
10 Settlement
3600kN Field Test
3600kN Computed
Settlement
1800kN Field Test
11
1800kN Computed
Settlement
0kN Field Test
0kN Computed
Settlement
12
Fig. 8.12 Test and computed settlements from unloading stages of P51
Time (min)
0 20 40 60 80 100 120 140
0
1800kN Field Test
1800kN Computed
2 Settlement
2700kN Field Test
2700kN Computed
4 Settlement
3600kN Field Test
3600kN Computed
Settlement (mm)
6 Settlement
4500kN Field Test
4500kN Computed
Settlement
8 5400kN Field Test
5400kN Computed
Settlement
6300kN Computed
Settlement
7200kN Field Test
12
7200kN Computed
Settlement
8100kN Field Test
14
8100kN Computed
Settlement
9000kN Field Test
16
Time (min)
0 50 100 150 200
7
9
Settlement (mm)
7200kN Field Test

10 7200kN Computed
Settlement
5400kN Field Test
11 5400kN Computed
Settlement
3600kN Field Test
3600kN Computed
12 Settlement
1800kN Field Test
1800kN Computed
13 Settlement
0kN Field Test
0kN Computed
Settlement
14
Time (min)
0 20 40 60 80 100 120 140
0 1800kN Field Test
1800kN Computed
Settlement
2700kN Field Test
2
2700kN Computed
Settlement
3600kN Field Test
4 3600kN Computed
Settlement
Settlement (mm)
4500kN Field Test
6 4500kN Computed
Settlement
5400kN Field Test
5400kN Computed
8 Settlement
6300kN Field Test
6300kN Computed
Settlement
7200kN Computed
Settlement
8100kN Field Test
12
8100kN Computed
Settlement
9000kN Field Test
14
Time (min)
0 50 100 150 200
0
4
Settlement (mm)
7200kN Field Test

6 7200kN Computed
Settlement
5400kN Field Test
5400kN Computed
8 Settlement
3600kN Field Test
3600kN Computed
Settlement
1800kN Computed
Settlement
0kN Field Test
12
Table 8.2 Test and computed settlements of P126 from loading 8,100 kN
Time Load LVDT1 LVDT2 LVDT3 LVDT4 Average Computed Information
(min) (kN) (mm) (mm) (mm) (mm) (mm) (mm)
5 8,100 8.12 8.41 6.69 8.77 7.9975 7.998 Data used for
15 8,100 8.17 8.49 6.82 8.91 8.0975 8.098 non-linear
regression
30 8,100 8.24 8.6 6.93 9.01 8.195 8.195
45 8,100 8.29 8.67 6.96 9.06 8.245 8.245
60 8,100 8.34 8.72 7.06 9.13 8.3125 8.313
90 8,100 8.38 8.77 7.09 9.16 8.35 8.350
120 8,100 8.44 8.83 7.17 9.24 8.42 8.420
150 8,100 8.46 8.84 7.2 9.27 8.4425 8.443 Data used for
180 8,100 8.49 8.87 7.23 9.29 8.47 8.470 comparison
8.2.2.2 Uplift Loaded Piles
The vertical displacements of two piles versus time are provided in Figs. 8.17 and
8.18. As shown in these diagrams, it was found that, during each loading stage, the
settlement trends during each maintained loading were almost parallel to the x-axis,
and the displacement difference between two loading stages were small and similar,
which represented no critical loading information. By comparing these gradients of
each loading/unloading stage between two piles, it was discovered that the settlement
16
14 Loading Stage 1
Loading Stage 2
12 Loading Stage 3
Displacement (mm)
Loading Stage 4
10 Loading Stage 5
Loading Stage 6
8 Loading Stage 7
Loading Stage 8
6
Loading Stage 9
Unloading Stage 1
4
Unloading Stage 2
Unloading Stage 3
2
Unloading Stage 4
Unloading Stage 5
0
0 500 1000 1500 2000
Time (min)
Fig. 8.17 Vertical displacement of P15 under maintained load

18
16 Loading Stage 1
Loading Stage 2
14
Loading Stage 3
Displacement (mm)
12 Loading Stage 4
Loading Stage 5
10
Loading Stage 6
Loading Stage 7
8
Loading Stage 8
6 Loading Stage 9
Unloading Stage 1
4
Unloading Stage 2
2 Unloading Stage 3
Unloading Stage 4
0 Unloading Stage 5
0 500 1000 1500 2000
Time (min)
Fig. 8.18 Vertical displacement of P16 under maintained load
of P16 was unstable. In other words, the gradient of P16 was relatively higher than
P15, which indicated that the base-and-shaft grouting technology improved the uplift
displacement behaviour.
Similar to the method illustrated in the previous section which is used to determine
the non-linear regression lines, define the displacement ratio Rd = Dt /D(Fi ) and
displacement ratio Rd = Dt /D(in) to determine the non-linear regression lines (uplift
loaded bored piles) during loading and unloading stages, respectively:
tn
Dt = D(Fi ) × (8.3)
mn + t n
tn
Dt = D(I n) × (8.4)
mn + tn
where:
Dt = Displacement at a certain loading stage at a certain time in mm
D(Fi ) = Displacement at the end of a certain loading stage in mm
D(in) = Displacement at the beginning of a certain loading stage in mm
m and n = Displacement coefficients obtained from non-linear regression curve
(Table 8.3).
From the loading stages, the regression lines were determined for P15 and P16,
as shown in Figs. 8.19 and 8.20, respectively. It was found that the data of P16
scattered, and the points of P15 assembled better and close to the regression line.
This illustrated that the grouting technology made the soil layers stable when the
Table 8.3 Settlement coefficients of uplift loaded bored piles

Type of pile Uplift loading Uplift unloading
m n R2 m n R2
Base-and-shaft grouted pile 0.00702 0.581 0.9997 0.01761 1.359 1.0
Traditional pile 0.01287 0.5059 0.9952 0.4913 3.921 0.999

maintained loading was applied. As shown in Figs. 8.21 and 8.22, the data indicated
that the displacement of uplifted piles during unloading stages was stable, which
was similar to the outcome of the compressive loaded pile during unloading stages.

16
600kN Field Test
600kN Computed
14 Settlement
900kN Field Test
900kN Computed
12 Settlement
1200kN Field Test
1200kN Computed
Settlement (mm)
10 Settlement
1500kN Field Test
1500kN Computed
Settlement
8 1800kN Field Test
1800kN Computed
Settlement
6 2100kN Field Test
2100kN Computed
Settlement
2400kN Field Test
4
2400kN Computed
Settlement
2700kN Field Test
2
2700kN Computed
a
3000kN Field Test
0
0 100 200 300 400 500
Time (min)
Based on these regression lines, the displacement equations were determined, and
the m and n values are provided in Table 8.3.
Based on the proposed Eqs. 8.3 and 8.4, the computed displacements of uplift
loaded piles with test data were compared, as shown in Figs. 8.23, 8.24, 8.25 to 8.26.
It can be seen that the proposed equations provided accurate results for the piles’
displacement.
8.3 Precast Concrete Piles
8.3.1 Small and Large Displacement Piles
8.3.1.1 Capacity Discussion
The capacity design of displacement piles as illustrated in Sect. 4.2, Chap. 4 is

detailed in this section. The capacity of a pile foundation can be determined by
calculating the shaft and end capacity, and based on the local project standard of
Technical Code for Building Pile Foundations, JGJ 94-2008 (2008), the capacity of
the precast rectangular piles and pipe piles can be determined as follows:
For solid rectangular piles:
15
14
Settlement (mm)
13
2400kN Field Test
2400kN Computed
Settlement
1800kN Field Test
12 1800kN Computed
Settlement
1200kN Field Test
1200kN Computed
Settlement
600kN Field Test
11
600kN Computed
Settlement
0kN Field Test
0kN Computed Settlement

10
0 50 100 150 200
Time (min)
18 600kN Field Test

600kN Computed
16 Settlement
900kN Field Test
900kN Computed
14 Settlement
1200kN Field Test
1200kN Computed
12
Settlement (mm)
Settlement
1500kN Field Test
1500kN Computed
10 Settlement
1800kN Field Test
8 1800kN Computed
Settlement
2100kN Field Test
6 2100kN Computed
Settlement
2400kN Field Test
4 2400kN Computed
Settlement
2700kN Field Test
2 2700kN Computed
Settlement
3000kN Field Test
0
0 50 100 150 200 250
Time (min)
8.3 Precast Concrete Piles 245
16
15
14
Settlement (mm)
2400kN Field Test
2400kN Computed
13 Settlement
1800kN Field Test
1800kN Computed
Settlement
1200kN Field Test
12
1200kN Computed
Settlement
600kN Field Test
11 600kN Computed
Settlement
0kN Field Test
0kN Computed
10 Settlement
0 50 100 150 200
Time (min)

Q uk = Q sk + Q pk = u qsik li + q pk A p (8.5)
For pipe piles:

Q uk = Q sk + Q pk = u qsik li + q pk A j + λ p A p1 (8.6)
λ p = 0.16h b /d when h b /d < 5
λ p = 0.8 when h b /d ≥ 5
where:
Q uk = Ultimate bearing capacity of pile
Q sk = Shaft capacity of pile
Q pk = End capacity of pile
u= Cross-section perimeter of pile
qsik = Shaft resistance
li = Thickness of ith soil
q pk = End resistance
Ap = Area from pile toe
π
Aj = Effective area of pile toe and A j = 4
(d 2 − d12 )
λp = Plug effect coefficient

A p1 = Hollow area of pile toe
hb = Embedment depth of pile in bearing stratum
d= External diameter of pipe pile
d1 = Internal diameter of pipe pile.
The shaft resistance, qsik , and end resistance, q pk , can be determined through
Tables 8.4 and 8.5, respectively. The pile capacity should be based on the nearest
borehole information, instead of the average soil thickness. Further, it is important
to determine the elevation of the first soil layer based on the construction elevation.
For example, for the concrete rectangular pile near the borehole with a label of
BH162, the construction elevation was about +4.4 m; hence, there would be around
1 m excavation, as shown in Fig. 8.27. After the soil thickness near the pile was
determined, based on the parameters obtained from laboratory and in situ tests, the
calculation of shaft and end capacity could be determined, as shown in Table 8.6.
The shaft, end and total capacities of all the tested pipe and rectangular piles
(Chap. 4) are summarised in Table 8.7. Based on the double tangent method, the
capacities of the tested concrete rectangular piles were determined. By plotting the
interpretation versus the calculated ultimate bearing capacity as shown in Fig. 8.28, it
can be seen that all the points were located in the lower part, which illustrated that the
double tangent method was conservative in determining the ultimate bearing capacity.
The ultimate bearing capacity obtained from the double tangent interpretation can be
modified by multiplying a factor. In this case, the factor η D was determined to be 1.83.
By plotting the designed capacity versus the modified double tangent interpretation,
as shown in Fig. 8.29, it can be seen that the modified double tangent interpretation
could provide a good prediction of the ultimate bearing capacity
Traditionally, the allowable loading, Qa , is equal to the calculated ultimate bearing
capacity divided by the safety factor of 2. By plotting the allowable load versus
the double tangent method, as shown in Fig. 8.30, it can be seen that the double
tangent method could appropriately predict the allowable loads. Compared with the
double tangent method, Chin’s method overestimated the ultimate bearing capacity.
A reduction factor (ηCh ) of 1.35 for a pile with a length less than 25 m and of 1.2 for
a pile with a length over 25 m is recommended. As shown in Fig. 8.31, based on the
proposed reduction factor, by plotting the modified Chin’s interpretation versus the
calculated capacity, all points assembled in the angle bisector of the first quartile.
For the precast pipe piles, the calculated ultimate bearing capacity was around
2,500 kN and the allowable load of a single pile was 1,250 kN. Based on the Technical
Code for Testing of Building Foundation Piles, JGJ 106-2014 (2014) the maximum
load applied during tests should be twice the allowable load—which was 2,500 kN.
However, the required capacity of this project was 520 kN, and these SLTs only
applied maximum loads of 1,040 kN (This is a very uneconomical design). Further
research on precast pipe piles with more loading application is required.
Table 8.4 Shaft resistance, qsik

Soil type Soil state Concrete Post grouting Traditional
precast pile bored pile bored pile
Fill 22–30 20–28 20–28
Sludge 14–20 12–18 12–19
Mucky soil 22–30 20–28 20–28
Cohesive IL > 1 24–40 21–38 21–38
soil 0.75 < I L ≤ 1 40–55 38–53 38–53
0.5 < I L ≤ 0.75 55–70 53–68 53–66
0.25 < I L ≤ 0.50 70–86 68–84 66–82
0 < I L ≤ 0.25 86–98 84–96 82–94
IL ≤ 0 98–105 96–102 94–104
Red clay 0.7 < aw ≤ 1 13–32 12–30 12–30
0.5 < aw ≤ 0.7 32–74 30–70 30–70
Silt Soft e > 0.9 26–46 24–42 24–42
Medium soft 0.75 ≤ e ≤ 0.9 46–66 42–62 42–62
Stiff e ≤ 0.75 66–88 62–82 62–82
Silty sand Loose 10 < N ≤ 15 24–48 22–46 22–46
Medium 15 < N ≤ 30 48–66 46–64 46–64
loose
Dense N > 30 66–88 64–86 64–86
Medium Medium 15 < N ≤ 30 54–74 53–72 53–72
sand dense
Dense N > 30 74–95 72–94 72–94
Coarse sand Medium 15 < N ≤ 30 74–95 74–95 76–98
dense
Dense N > 30 95–116 95–116 98–120
Gravel sand Loose 5 < N 63.5 ≤ 15 70–110 50–90 60–100
Medium to N 63.5 > 30 116–138 116–130 112–130
dense
Gravel Medium to N 63.5 > 10 160–200 135–150 135–150
dense
Cobble Medium to N 63.5 > 10 200–300 140–170 150–170
dense
Weathered 30 < N ≤ 50 100–120 80–100 80–100
soft rock
Weathered 30 < N ≤ 50 140–160 120–140 120–150
rag-stone
Highly N 63.5 > 10 160–240 140–200 140–220
weathered
soft rock
(continued)
Table 8.4 (continued)

Soil type Soil state Concrete Post grouting Traditional
precast pile bored pile bored pile
Highly N 63.5 > 10 220–300 160–240 160–260
weathered
rag-stone
Source JGJ 94-2008 (Ministry of Construction of the People’s Republic of China 2008)
8.3.1.2 Settlement Discussion
As illustrated in Chap. 4, Fig. 4.1 shows the tested piles at Areas A, B and C. The soil
profiles of Areas A, B and C are shown in Figs. 8.32, 8.33 and 8.34. It was found that
a pile length ranging from 20 to 30 m may be required to reach the bearing stratum
(angular gravel) in Area A, a pile length ranging from 24 to 32 mm may be required
in Area B, and a pile length ranging from 24 to 39 m may be needed in Area C.
The load settlement results of compressively loaded piles were presented in
Chap. 4. As shown in Fig. 4.13, these Q-s curves of all 21 m piles were different.
It was found that the maximum settlements of piles ranged from 16 to 24 mm. This
was because the bearing stratum was brecciated gravel, and the shaft resistance was
different (thickness of soil layers varied). This total head displacement is the sum up
of pile deformation and soil deformation. It was also found that, after consecutive
unloading back to 0 kN, all piles curves were parallel (unloading curves). This was
because all these piles were made of the same concrete with the same moduli. Once
the loading was released, the shortened pile would expand back at the same rate
(settlement/load, unit of mm/kN), and the plastic deformation of soil dominates
elastic deformation.
When defining the permanent settlement ratio, R P−S , governed by Eq. 8.7, the
average R P−S of small displacement pipe piles was determined to be 52.2% in Area
A (strength of concrete of 30 MPa) and 54.9% in Area B (strength of concrete of
50 MPa):
smax − s p
R P−S = × 100 (8.7)
smax
where:
R P−S = Permanent settlement ratio;
smax = Settlement under maximum applied load;
sp = Permanent settlement after releasing load.
As provided in Tables 4.1 and 4.2, the ratios of all these tested piles were close
to the average permanent ratio in Areas A and B, respectively. Similar to these pipe
piles, as depicted in Table 4.3, the average R P−S of large displacement rectangular
piles was determined to be 49.0% (strength of concrete was 40 MPa) and the ratios
of rectangular piles were close to the average value of 49.0%.
Table 8.5 End resistance, qpk
Soil type Soil state Concrete precast pile Post grouting bored pile Traditional bored pile
Pile Type l l l l l l l l l l l
Cohesive 0.75 < I L ≤ 1 210–850 650–1,400 1,200–1,800 1,300–1,900 150–250 250–300 300–450 300–450 200–400 400–700 700–950
soil 0.5 < I L ≤ 0.75 850–1,700 1,400–2,200 1,900–2,800 2,300–3,600 350–450 450–600 600–750 750–800 500–700 800–1,100 1,000–1,600
0.25 < I L ≤ 0.50 1,500–2,300 2,300–3,300 2,700–3,600 3,600–4,400 800–900 900–1,000 1,000–1,200 1,200–1,400 850–1,100 1,500–1,700 1,700–1,900
0 < I L ≤ 0.25 2,500–3,800 3,800–5,500 5,500–6,000 6,000–6,800 1,100–1,200 1,200–1,400 1,400–1,600 1,600–1,800 1,600–1,800 2,200–2,400 2,600–2,800
Silt Medium 0.75 ≤ e ≤ 0.9 950–1,700 1,400–2,100 1,900–2,700 2,500–3,400 300–500 500–650 650–750 750–850 800–1,200 1,200–1,400 1,400–1,600
dense
Dense E ≤ 0.75 1,500–2,600 2,100–3,000 2,700–3,600 3,600–4,400 650–900 750–950 900–1,100 1,100–1,200 1,200–1,700 1,400–1,900 1,600–2,100
Silty sand Medium 10 < N ≤ 15 1,000–1,600 1,500–2,300 1,900–2,700 2,100–3,000 350–500 450–600 600–700 650–750 500–950 1,300–1,600 1,500–1,700
dense
Dense N > 15 1,400–2,200 2,100–3,000 3,000–4,500 3,800–5,500 600–750 750–900 900–1,100 1,100–1,200 900–1,000 1,700–1,900 1,700–1,900
Fine sand Medium N > 15 2,500–4,000 3,600–5,000 4,400–6,000 5,300–7,000 650–850 900–1,200 1,200–1,500 1,500–1,800 1,200–1,600 2,000–2,400 2,400–2,700
Medium dense to 4,000–6,000 5,500–7,000 6,500–8,000 7,500–9,000 850–1,050 1,100–1,500 1,500–1,900 1,900–2,100 1,800–2,400 2,800–3,800 3,600–4,400
sand dense
Coarse 5,700–7,500 7,500–8,500 8,500–10,000 9,500–11,000 1,500–1,800 2,100–2,400 2,400–2,600 2,600–2,800 2,900–3,600 4,000–4,600 4,600–5,200
sand
Gravel Medium N > 15 6,000–9,500 9,000–10,500 1,400–2,000 2,000–3,200 3,500–5,000
sand dense to
Gravel dense N63.5 > 10 7,000–10,000 9,500–11,500 1,800–2,200 2,200–3,600 4,000–5,500
sand
Cobble N63.5 > 10 8,000–11,000 10,500–13,000 2,000–3,000 3,000–4,000 4,500–6,500
Weathered 30 < N ≤ 50 4,000–6,000 1,000–1,600 1,200–2,000
soft rock
Weathered 30 < N ≤ 50 5,000–8,000 1,200–2,000 1,400–2,400
rag-stone
Highly N63.5 > 10 6,000–9,000 1,400–2,200 1,600–2,600
weathered
soft rock
(continued)
249
250
Soil type Soil state Concrete precast pile Post grouting bored pile Traditional bored pile
Pile Type l l l l l l l l l l l
Highly N63.5 > 10 7,000–11,000 1,800–2,800 2,000–3,000
weathered
rag-stone
Source JGJ 94-2008 (Ministry of Construction of the People’s Republic of China 2008)
8 Capacity and Settlement Analysis
Fig. 8.27 Borehole log 162
Table 8.6 Calculation of pile

1. Shaft capacity
adjacent to borehole 162
Label Soil thickness (m) qsik (kPa) Shaft resistance
(kN)
1 1 20 20
1-1 0 20 0
2 3 50 150
3 5 38 190
3-1 3 18 54
4 1.2 65 78
5 3 78 234
5-1 1.2 50 60
5 8 78 624
6 0 150
Total shaft resistance 2,256
2. End resistance
End layer Ap Toe area End bearing
(kN)
6 9,500 0.16 1,520
3. Ultimate bearing capacity 3,776
(kN)
Table 8.7 Summarization of tested precast piles

Concrete rectangular pile Concrete pipe pile
Pile Pile kN Pile Pile kN Pile Pile kN
label capacity label capacity Label Capacity
P1 Shaft 2,256 P1172 Shaft 1,333.37 P643 Shaft 1,496.65
capacity capacity capacity
Total 3,776 Total 2,329.24 Total 2,492.52
P106 Shaft 2,447.36 P1470 Shaft 1,358.87 P1509 Shaft 1,531.44
Total 3,967.36 Total 2,354.74 Total 2,527.32
P124 Shaft 2,634.88 P1175 Shaft 1,343.92 S73 Shaft 1,653.40
(continued)

Concrete rectangular pile Concrete pipe pile
Pile Pile kN Pile Pile kN Pile Pile kN
label capacity label capacity Label Capacity
P137 Shaft 2,611.36 P1287 Shaft 1,714.44
capacity capacity
Total 4,131.36 Total 2,710.31
capacity capacity
P4 Shaft 2,638.4 P1299 Shaft 1,729.39
capacity capacity
Total 4,158.4 Total 2,725.26
capacity capacity
P16 Shaft 2,651.36 P1453 Shaft 1,727.13
capacity capacity
Total 4,171.36 Total 2,723.00
capacity capacity
P85 Shaft 2,748.8 P630 Shaft 1,412.25
capacity capacity
Total 4,268.8 Total 2,408.12
capacity capacity
P93 Shaft 2,693.12 P1473 Shaft 1,432.34
capacity capacity
Total 4,213.12 Total 2,428.22
capacity capacity
6000
5500
Double Tangent Interpretation
5000
4500
22m
4000
25m
3500 26m
27m
3000
28m
2500
2000
1500
1500 2500 3500 4500 5500
Designed Capacity
Fig. 8.28 Designed capacity versus double tangent interpretation

6000
Modified Double Tangent Interpretation

5500
5000
4500 22m
25m
4000 26m
27m
3500 28m
3000
2500
2500 3500 4500 5500
Designed Capcity
Fig. 8.29 Designed capacity versus modified double tangent interpretation
4500
Double Tangent Interpretation
4000
3500
22m
3000 25m
26m
27m
2500
28m
2000
1500
1500 2000 2500 3000 3500 4000 4500
Allowable Load
Fig. 8.30 Allowable load versus double tangent interpretation
As illustrated in Tables 4.1, 4.2 and 4.3, the permanent settlement ratio of small
displacement piles was greater than the large displacement piles; however, this was
not a general outcome, since there was a lot of uncertainty. Thus, further research is
required with consideration of the strength of concrete used, pile length, shaft areas
and so forth.
6000
5500
Modified Chin's Interpretation
5000
22m
4500
25m
4000 26m
27m
3500 28m
3000
2500
2500 3000 3500 4000 4500 5000 5500 6000
Designed Capacity
Fig. 8.31 Designed capacity versus modified Chin’s interpretation
Fig. 8.32 Soil profile of Area A based on borehole logs
Similar to the 21 m piles, the load settlement results of the 22 and 24 m piles were
provided in Figs. 4.14 and 4.15. These piles were made of the same concrete material
of 30 MPa. It was found that the maximum settlements were different because the
shaft resistances of each layer were different. It could also be seen that the curves
from the unloading stages were parallel with each other because the same material
was being used, and the elongation deformation would develop at the same rate.
To further confirm this, the load–settlement curves of piles with the same concrete
(50 MPa) were parallel from the unloading stage, as shown in Fig. 4.17. If plotting
Fig. 8.33 Soil profile of Area B based on borehole logs
Fig. 8.34 Soil profile of Area C based on borehole logs
the load–settlement curves of piles with different concrete strengths, these curves
would not be parallel with each other from the unloading stage. The same outcome
was obtained for the 27 and 28 m rectangular piles, as shown in Figs. 4.18 and 4.19.
Under consideration of pipe and rectangular pile lengths, as shown in Figs. 4.16
and 4.21, respectively, the maximum settlement of pipe and rectangular piles were
different, yet the R P−S values of the pipe and rectangular piles were close to the same
type of pile. In detail, the R P−S of P1172 (20 m), P1165 (21 m), P1485 (22 m) and
P1299 (24 m) was 55.14, 53.72, 51.86 and 54.7%, respectively, for the pipe piles,
and the R P−S of P14 (22 m), P106 (25 m), P91 (26 m), P124 (27 m) and P147 (28 m)
was 49.02, 49.09, 54.37, 48.14 and 49.03%, respectively, for the rectangular piles.
This indicates that, for the piles that reached the hard soil stratum, during loading
and unloading, the structure deformation was a dominant factor. Thus, the permanent
settlement ratios of one type of pile will be close to each other.
To further prove this outcome with consideration of loading value, the results of
P1 and P106 were compared. These piles were made of the same concrete and were
25 m. As shown in Fig. 4.20, different loading increments were applied, and this
demonstrated that the two curves were parallel with each other during unloading
stages. The R P−S of P1 and P106 was found to be 51.05% and 49.09%, respectively,
which was close to each other.
8.3.2 Non-displacement Precast Piles
8.3.2.1 Capacity Discussion
As discussed in literature review of Chap. 2, there has been limited previous research
on non-displacement precast piles—the construction process in which soil is first
removed and then the precast pile is lowered. Thus, this study investigated the
behaviours of six piles under uplift and lateral loading, and the results were provided
in Chap. 4.
For the horizontal load movement results obtained from the lateral SLTs, the crit-
ical load and ultimate load could be preliminarily determined from Figs. 4.22, 4.23
and 4.24 for P15 (24.7 m), P163 (26.6 m) and P152 (28 m), respectively. However, this
is an empirical method that is highly dependent on practical experience. For example,
the ultimate horizontal capacity of P163 was difficult to determine. Further, it was
found that, before the ultimate loading being applied, the piles’ behaviours were
similar. As shown in Figs. 4.22 and 4.23, the maximum horizontal movements of
P15 and P163 were 5.56 and 5.125 mm, respectively, when 400 kN was applied. In
addition, the residual movements of P15 and P163 were 0.59 and 0.57 mm, respec-
tively. To further prove this outcome, the maximum horizontal movements of P15
and P163 were 11.28 and 11.91 mm, and the residual movements were 1.51 and
2.66 mm, respectively, when 500 kN was applied.
These three typical presentations preliminarily illustrated that the critical loads
and ultimate loads increased with pile length. However, the ultimate load of P163 was
difficult to determine. The method of horizontal SLT interpretation involved plotting
horizontal loading versus nominated gradient settlement. As shown in Figs. 4.25,
4.26 and 4.27, theoretically, two ‘turning points’ could be determined that represented
critical and ultimate loading conditions. Through determining these points, it was
found that critical load and ultimate load increased with pile length, as illustrated in
Table 4.6.
The load–settlement curves of the three uplift loaded piles were provided in
Fig. 4.28. It can be seen that the curves from the loading stages were relatively
linear, and the maximum uplift displacements were small. Through interpretation of
Mazurkiewicz’s method, the ultimate uplift capacities of the three piles were deter-
mined in Fig. 4.29. It can be seen that the capacity increased with pile length. The load
transfer mechanism of the non-displacement precast pile was presented in Figs. 4.30
and 4.31. This behaviour was similar to the compressive loaded piles, where loading
was transferred into piles resisted by shaft force developed by soil and pile shaft
surface, and the there is no load transferred from the pile end.
8.3.2.2 Displacement Discussion
The proposed empirical Eqs. 8.1 to 8.2 and 8.3 and 8.4 can be used to determine the
compressive and uplift displacement prediction respectively, with various coefficient
m and n values which can be obtained from non-linear regression lines. Similar
to these equations, for the lateral loaded piles, during the loading and unloading
stages, the horizontal displacement ratio was defined as Rh = Ht /H(Fi ) . Under a
certain lateral loading, the horizontal displacement at a certain time (Ht ) divided by
the displacement (H(Fi ) ) which represents the final displacement at the end of this
loading or unloading will be less than 1, but over 0. By plotting the time versus the
horizontal displacement ratio, Rh , from the loading and unloading stages, as shown
in Figs. 8.35, 8.36, 8.37 and 8.38, the non-linear regression lines were determined.
Finally, the horizontal displacement equation was proposed, as illustrated in Eq. 8.8:
tn
Ht = H(Fi ) × (8.8)
mn + tn
where:
Ht = Settlement at a certain loading stage at a certain time in mm;
H(Fi ) = Settlement at the end of a certain loading/unloading stage in mm;
m and n = Settlement coefficients obtained from non-linear regression curve.

Fig. 8.36 Non-linear regression of P15 (unloading stage)
As shown in Fig. 8.36, 8.37 and 8.38, for the piles during loading and unloading
stages, some points should be excluded because the horizontal displacement ratio
value was defined as 0 to 1. This was mostly because of the error caused by dial
gauges. Also, for the pile during unloading stages, when the pile foundation deflects
back to its initial location, the soil layers may collapse and consequently lead to the
inaccurate reading of horizontal displacement. As shown in Fig. 8.36, this regression
line was relatively trustworthy and could provide good displacement predictions,
Fig. 8.38 Non-linear regression of P163 (unloading stage)
12
10
8 100kN Field Test

Settlement (mm)
100kN Computed
Settlement
200kN Field Test
6
200kN Computed
Settlement
300kN Field Test
4 300kN Computed
Settlement
400kN Field Test
400kN Computed
2 Settlement
500kN Field Test
500kN Computed
Settlement
0
0 10 20 30 40
Time (min)
2.5
100kN Field Test

Settlement (mm)
1.5 100kN Computed

Settlement
200kN Field Test
200kN Computed
Settlement
1 300kN Field Test
300kN Computed
Settlement
400kN Field Test
0.5 400kN Computed
Settlement
500kN Field Test
500kN Computed
0 Settlement
0 10 20 30 40 50
Time (min)
as illustrated in Fig. 8.40. The regression line with an R-squared value of 0.7583,
as depicted in Fig. 8.38, is not recommended for use. From the loading stages, as
shown in Figs. 8.35 and 8.37, these regression lines could provide accurate values of
coefficients m and n; hence, the predicted horizontal displacement was trustworthy,
as illustrated in Figs. 8.39 and 8.41. The limitation of this research was that there is
no comparison between this non-displacement pile and displacement driven precast
pile. Hence, the effect caused by removing soil first is unknown. Only the behaviours
of these non-displacement piles have been investigated; thus, further research is
required.
8.4 Concrete Composite Piles
8.4.1 FRP Laminar–Confined Composite Piles
By comparing the Q-s curves of CFRP laminar–confined composite piles with tradi-
tional concrete piles, as depicted in Figs. 5.8 and 5.9, it was found that the settlements
of the CFRP-confined pile were less than the traditional piles during the application
of loading. After the loads were unloaded, the permanent settlements of these two
30
100kN Field Test

25
100kN Computed
Settlement
200kN Field Test
20 200kN Computed
Settlement (mm)
Settlement
300kN Field Test
300kN Computed
15 Settlement
400kN Field Test
400kN Computed
10 Settlement
500kN Field Test
500kN Computed
Settlement
5 600kN Field Test
600kN Computed
Settlement
700kN Field Test
0
0 10 20 30 40
Time (min)
composite piles were less than the traditional concrete piles. Since the geotechnical
condition was the same (the soil layers were the same around the tested piles) and all
piles reached the bearing stratum, the settlement was primarily due to the concrete
shortening. Under the same loading value, the shortening of the CFRP composite
pile was less than the normal pile because of the circumferential stress caused by
CFRP confinement.
Through the s-lgQ curves, as shown in Fig. 5.10, it was difficult to find a turning
point that could represent the failure criteria; hence, the determination of the ultimate
bearing capacity of these tested piles needed to be based on interpretation. The s-
lgt curves provided the same information as the Q-s curves, and indicated that the
settlements of the CFRP-confined piles were less than the traditional piles. Similar
to the s-lgQ curves, DeBeer’s method also provided the same outcome because there
was no turning point, as shown in Figs. 5.17 and 5.18.
As displayed in Figs. 5.11, 5.12, 5.13 and 5.14, there was no point showing
a dramatic increasing of vertical settlement in each loading line, and no extreme
‘drop’ illustrating dramatic settlement increase under one specific loading stage.
The ultimate bearing capacity could not be determined via these four diagrams, yet
provided valuable information that these four piles’ capacity should be more than
the maximum applied loads. Through determining the elastic lines and offsetting
these lines based on Eq. 2.38, Davisson’s method also provided the same outcome.
Given that the offset line does not intersect with the Q-s curves obtained from the
8.4 Concrete Composite Piles 263
loading curves, the ultimate bearing capacity of these piles should be more than the
maximum loads applied from the SLTs, as shown in Figs. 5.19 and 5.20.
Based on the double tangent method, the pile capacities of P116, P115, P2108 and
P2054 were determined to be 2,600, 2,400, 850 and 760 kN, respectively, as shown
in Figs. 5.15 and 5.16. However, as discussed above, the capacity of these four piles
should be greater than 4,800, 4,800, 1,300 and 1,300 kN, respectively. The reason
is that the double tangent method is a conservative method; thus, the determined
ultimate load would be less than the maximum applied loads, especially when used
for obtaining the pile capacity under non-plunging SLTs and proof tests. Following
the method provided by Yang and Xiao (2011), which is based on Chin’s method,
the functions were used to evaluate the capacities of piles, as depicted in Figs. 5.21,
5.22, 5.23 and 5.24.
The values of ultimate bearing capacity, shaft resistance and toe resistance were
summarised in Table 5.1. It was found from Chin’s method that the CFRP-confined
pipe pile had a greater capacity than the traditional pipe pile, and that this confine-
ment increased the shaft and toe resistance simultaneously. From the double tangent
method, it was also found that the CFRP confinement increased the pipe pile capacity.
Further, the double tangent method illustrated that the CFRP confinement could
increase rectangular pile capacity.
This investigation focused on the behaviour of piles under CFRP confinement.
However, the limitation was that the loading applied was not sufficiently large, which
led to difficulty in determining the capacity. Further research is required with consid-
eration of the types of CFRP material used, quantity of laminar used, FRP orientation
arrangement and so forth.
8.4.2 FRP Bar–Reinforced Composite Piles
The deflection behaviour of GFRP-reinforced concrete piles was researched in

comparison with traditional steel-reinforced concrete piles in Chap. 5. These two
piles suffered from lateral loading caused by changed earth pressure during excava-
tion construction, and the deflection behaviours were monitored with consideration
of support strut application, as shown in Figs. 5.39 and 5.40.
Analysis of the lateral displacement diagrams of these piles showed that the deflec-
tion increased from pile head along the pile to a certain depth, and this maximum
horizontal movement would increase along the pile. Also, it could be seen that the
horizontal movements along the pile shaft are small, thus, the project was relatively
overdesigned. The GFRP composite pile demonstrated a maximum lateral deflection
of −6.02 mm at the depth of −12.5 m, while the maximum lateral deflection of the
steel-reinforced concrete bored pile was −10.56 mm at the depth of 12.5 m. Further,
it was found that, after each support strut was applied, the maximum horizontal
deflection of the GFRP composite pile was less than the normal reinforced concrete
pile, and the maximum horizontal movement of piles was approximately 12 to 13 m
underneath, as depicted in Table 5.5.
It cannot be concluded that the GFRP bar–reinforced concrete pile demonstrated

better than the traditionally reinforced concrete pile, since the ultimate tensile strength
of GFRP and steel are different (GFRP > 500 MPa; steel = 400 MPa). However,
it can be concluded that the GFRP bar–reinforced concrete bored pile could be a
suitable alternative to replace the traditionally reinforced concrete bored pile in deep
excavation. Further research is required to determine the stresses along the pile, and
more advanced analysis is needed, such as a bending moment study along the pile
length.
8.5 Piles Under Ultimate Load
Chapter 7 provided a pile behaviour investigation with consideration of the inad-

equate concrete strength of piles, punching failure of piles and piles applied with
eccentric loads. SLTs were conducted and the results of the Q-s, s-lgt and s-lgQ
curves were analysed for capacity determination. Chapter 7 also presented the double
tangent method and Davisson’s offset method interpretations of these failed piles.
By using the data before failure condition occurred and selecting Chin’s method,
prediction of these piles’ capacities was also presented.
Through observation of these piles under ultimate loading, it was discovered that,
for the pile with low concrete strength caused by inappropriate construction, the
concrete will crack at the pile head, which may extend up to 30 mm, and the concrete
will crush at a certain depth. The reinforcement in the crushed concrete zone will be
forced outwards. In the case of the pile which suffered from eccentricity, concrete will
crush in the compressive zone, and, in this zone, the reinforcement will be forced
outwards. The steel in the tensile zone, however, resists loads after the concrete
has ruptured, and the pile consequently fails. In the case of the pile suffering from
punching failure, there were no concrete cracks discovered, but the soil crack could
extend up to 50 mm.
By comparing the Q-s and s-lgQ curves, the ultimate bearing capacities were the
same as those summarised in Table 7.3, this is because these methodologies are to find
a turning point, which illustrated a huge settlement increase. It can also be seen that,
for these piles under ultimate loading, the double tangent method can provide good
results, although they may be slightly conservative (3,900 kN < 4,680 kN of P80).
Further, it can be seen that Davisson’s method can provide a good interpretation,
although it may sometimes overestimate the ultimate capacity (5,200 kN > 4,680
kN). In addition, it was found that Chin’s method can provide acceptable prediction
of ultimate bearing capacity, although it may sometimes overestimate the capacity
(10,000 kN > 7,560 kN). Further, it can be seen that Chin’s method was not fit to
predict the pile capacity under eccentric loading.
For the results of the s-lgt curves, as shown in Fig. 7.18, the behaviour of the
pile with inadequate concrete strength could be determined. It can be seen that,
during the applied loading of 3,640 kN, there was a decreasing curve. This increase
in settlement was caused by the concrete being crushed. Given that there was only
8.5 Piles Under Ultimate Load 265
compressive loading, this failure in the concrete structure did not lead to failure of
the pile foundation. It was not until the loading of 4,160 kN was applied that the soil-
pile system failed. The behaviour of the pile with eccentric loading can be seen in
Fig. 7.22. It was found that the concrete structure failed when a loading of 4,680 kN
was applied. Two downwards trends were determined. The first downwards trend
was caused by the total failure of the concrete structure, as proven by analysing
Fig. 7.11 (After the field test was performed, the steel in the tensile zone yielded).
The second downwards trend was caused by the failure of the pile soil system. As
shown in Fig. 7.26, for plunging failure pile, the pile foundation was unstable under
the loading of 7,560 kN, and resisted the loading of 8,400 kN, but finally failed.
This Chapter has detailed analysis and discussion referring to the capacity and the
settlement of those different types of pile foundations. In terms of the capacity,
traditional result presentations were discussed and the extrapolations of different
types of piles were compared. Most importantly, this Chapter provides a method
for the reader to derive the empirical formula, which can be used for settlement
prediction. This method considered three loads orientations, namely the compressive,
uplift and horizontal directions. Also, the empirical formula considered the loading
and unloading stages.
References
American Association of State Highway and Transportation Officials (AASHTO), Standard

Specifications for Highway Bridges, Washington, D. C., 2002.
Ministry of Construction of the People’s Republic of China. (2008). Technical code for building
China.
Republic of China.
Samtani, N. C., & Nowatzki, E. A. (2006b). Soils and foundations—Volume II (No. FHWA-NHI-
16-009). US Department of Transportation.
Yang, P., Hu, H.-S., & Xu, J.-F. (2012). Settlement characteristics of pile composite foundation
under staged loading. Procedia Environmental Sciences, 12, 1055–1062.
Yang, H., & Xiao, D. (2011). Back analysis of static pile load test for SPT-based pile design: A
Singapore experience. In Advances in Pile Foundations, Geosynthetics, Geoinvestigations, and
Foundation Failure Analysis and Repairs (pp. 144–152).
Chapter 9
Conclusions and Recommendations
This book provides the theoretical background, technical concepts and extrapola-
tion of pile foundations related to state-of-the-art field tests. Based on the literature,
Chap. 2 reviewed the up-to-date research and determined its limitations, the general
principles and practices are detailed. Following this, based on the observed limita-
tions, further research was provided in Chaps. 3 to 7 though testing a total number of
64 piles. After discussion of these results which was provided in Chap. 8, the current
chapter presents the study conclusions and recommendations with reference to the
tested pile foundations.
9.2 Post Grouted Piles
To determine the capacity of piles with different grouting technology applications,

a total number of 5 piles were selected to test. Compressive and uplift loaded SLTs
were conducted and two dynamic load tests were performed on two of these piles.
Though these tests, the behaviours of these piles under different load directions were
identified, and the ultimate bearing capacity was determined based on interpretation.
The results led to the following conclusions:
For these bored piles under compressive loads:
1. From the SLT results, the base-and-shaft grouted pile increased about 9.82% of
its capacity (compared with the pile without any grouting).
2. From the SLT results, the base grouting pile increased approximately 2.89% of
its capacity (compared with the pile without any grouting).
3. From the SLT results, the base-and-shaft grouted pile increased around 6.1% of
its capacity (compared with the base grouting pile).

https://doi.org/10.1007/978-981-33-6183-6_9
268 9 Conclusions and Recommendations
4. From the dynamic load test results, the base-and shaft grouted pile increased
around 8.6% of its capacity (compared with the base grouting pile).
5. An empirical formula for compressive settlement prediction was proposed
(including loading and unloading stages, Chap. 8, Sect. 8.2.2.1).
For piles under uplift loads:

1. The base-and-shaft grouted pile increased about 15.7% of its capacity (compared
with the pile without any grouting).
2. The shaft resistance of the base-and-shaft grouted pile increased about 15.0% of
its capacity (compared with the pile with base grouting only).
3. An empirical formula and uplift displacement prediction were proposed
(including loading and unloading stages, Chap. 8, Sect. 8.2.2.2).
The interpretation methods indicated that the double tangent method was conser-
vative and Chin’s method slightly overestimated the pile capacity for compressively
loaded piles. Mazurkiewicz’s method overestimated the capacity for uplift loaded
piles. Further, the offset method for uplift loaded piles was too conservative. After
the uplift SLTs were performed, the recommended uplift design load should be equal
to two-thirds to four-fifths of this failure load.
For the precast displacement piles, a total number of 40 piles were selected. Twenty-
two piles with various lengths were tested in Area A, three piles with various lengths
were tested in Area B, and fifteen piles with various lengths were tested in Area C.
By defining the permanent settlement ratio, RP-S , it was found that:
1. The RP-S from all piles with concrete strength of 30 MPa was close to 52.2%,
and the Q-s curves from the unloading stages were parallel with each other in
Area A.
Area B.
Area C.
The other conclusions were as follows:
1. For the precast piles, if the tested piles are the same type, the Q-s curves from
the loading stages will be different (shaft resistance varied) and the Q-s curves
from the unloading stages will be parallel with each other.
2. For the proof test result, if using double tangent method, the ultimate bearing
capacity is the interpreted value multiplying by factor ηD . If using Chin’s method,
the ultimate bearing capacity is the interpreted value divided by reduction factor
ηCh . In the case of the project in China, ηD = 1.83, ηCh = 1.35 (L < 25) and ηCh
= 1.2 (L > 25) are recommended.
If in one area, the Q-s curve of a pile from the unloading stages are not parallel to
the curves of other piles which are made of same materials (gradient of the unloading
trend line is evidently different), it represents that the soil layers from this location
may be very different to the others, or the pile are damaged. Thus, more borehole
observations are recommended, and a low stain integrity test is suggested. It seems
that the RP-S value of large displacement piles is smaller than that of small displace-
ment piles; however, there is much variables such as concrete strength and geometric
configuration being different among these tested piles, further research is required.
For the behavior determination of non-displacement precast piles, 6 piles were
selected with different design lengths. Three uplift and three horizontal SLTs were
performed, with conclusions as follows:
1. The ultimate uplift capacity of the pile increased with pile length.
2. The ultimate horizontal capacity of the pile increased with pile length.
3. The transferred loads from the pile decreased along the pile length, and shaft
resistance increased along the pile.
4. Shaft resistance will develop under increased loads, and particular soil layers
provide dominant shaft resistance.
5. An empirical formula for the horizontal displacement prediction were proposed
(including loading and unloading stages). The proposed equation from the
loading stages is recommended; however, the equation from the unloading stages
may provide inaccurate results. Hence, it is only used for optional analysis
(Chap. 8, Sect. 8.3.2.2).
Similar to the interpretation of grouted piles illustrated in the previous section,
the offset method of uplift loaded piles is not recommended. For the horizontal
SLTs, the nominated gradient settlement method is recommended. Given that there
was no driven precast pile being tested, the capacity comparison between the non-
displacement and displacement pile is unknown; hence, the effect caused by removing
the soil first is uncertain. Thus, further research is required.
9.4 Concrete Composite Piles
To determine the effect of CFRP confinement, two types of precast piles were
selected. For this research, a total number of 4 piles were tested, including one
pipe pile and one pipe pile with CFRP application, one rectangular pile and one
rectangular pile with CFRP application. Through results analysis and interpretation,
it was concluded that:
1. The CFRP confinement increased the total capacity of rectangular pile by about
8.3%.
2. The CFRP confinement increased the total capacity of pipe pile by about 11.8%
and 25% based on the double tangent method and Chin’s method, respectively.
3. For the pipe pile, the CFRP confinement increased the shaft resistance by 27.2%.
4. For the pipe pile, the CFRP confinement increased the end resistance by 23.7%.
The limitation of this research was that the applied loads were insufficient, which
led to the small effect of CFRP confinement. Further research is also required with
consideration of the types of FRP material selected, quantity of laminar applied
and FRP orientation arrangement. This new type of pile with FRP application is
recommended for use in coastal conditions if the pile is fully covered and confined by
FRP (compressive loaded pile). From the structural point of view, this FRP confined
pile should perform very well when the pile toe reaching strong bearing strum such
as rocks.
For the study of FRP inside reinforcement application in geotechnical conditions,
2 piles were monitored by inclinometer. The deflection behaviors of these two piles
were determined and the results were compared. It could not be concluded that
the GFRP bar–reinforced concrete pile demonstrated better than the traditionally
reinforced concrete pile, since the ultimate tensile strength of GFRP and steel are
different (GFRP > 500 MPa; steel = 400 MPa). However, it could be concluded
that the GFRP bar–reinforced concrete bored pile could be a suitable alternative to
replace the traditionally reinforced concrete bored pile in deep excavation. Further
research is required with determination of stresses along the pile, and more advanced
analysis is needed, such as a bending moment study along the pile length.
9.5 Super-Long and Large Diameter Piles
To research the load transfer, shaft resistance development and distribution of the
super-long and large-diameter piles, 3 drilled shaft concrete piles were tested. From
the conventional diagram presentation, it was concluded that:
1. By increasing the pile length, especially when reaching a harder bearing stratum,
it can effectively increase the ultimate bearing capacity of the single pile. Also,
increment of the pile diameter and pile length can improves the total capacity of
a single pile. By comparison of these piles, the 83 m pile with 1800 mm diameter
was conservatively designed.
2. The exist of compressible silty sand from the pile tip will led to large pile
toe displacement, and thus presented a small concrete compression. These pile
toe and pile head curves (Q-s curves) also illustrate that the 83 m length pile
with 1800 mm diameter was conservatively designed. Moreover, the trend of
Q-s curves from the loading stages of this 83 m pile was discovered relative
linear (close to the elastic deformation line), which illustrated the deformation
of concrete element is the main factor referring to the ultimate bearing capacity,
thus by applying post-grouting technology would not be recommended.
9.5 Super-Long and Large Diameter Piles 271
3. From the load transfer curves (load versus depths), the principle finding was that
the transferred load decreased along the pile length during each applied load,
but the gradients were different, this is due to different pile shaft and the exist
of different soil layers. Also, when condition of the soil being soft and very soft
(clay), and loose to very loose (sand), the transfer rate of axial force will be
smaller.
4. By comparing the shaft and end bearing percentage, it was found that the increase
of the pile length can make contribution to the increase of the shaft resistance and
hence reduce the transferred loads to the bearing stratum. Also, by comparison
of the proportion of shaft and toe resistance under working load, the 83 m pile
with 1800 mm diameter was a shaft dominated pile (the total capacity of pile is
mainly contributed by shaft resistance).
5. From the load-shaft resistance curves, it was found that the shaft resistance of
the most of soil layers developed from zero to a maximum value and then main-
tained when increasing the applied loads. Also, the shaft resistance from the
upper layers developed firstly. The shaft resistance softening and hardening were
discovered from some soil layers. Most importantly, the ‘Mutual Compensation’
phenomenon was discovered, that is when shaft softening occurred from one soil
layer, the shaft hardening of the other soil will occur simultaneously.
6. From the Shaft Resistance Distribution Curves, It Was Found that the Shaft
Resistance of Each Soil Layer Does not Develop Simultaneously. The Shaft
Resistance Distribution Illustrated an ‘R’ Shape.
From a practical engineering point of view, the soil layers around these three piles
were mostly sand, which possesses a relatively high voids; thus, post-grouting tech-
nology is highly recommended to improve the ultimate bearing capacity. However,
it is recommended that shaft-and-base grouting be applied to P66 and P12, while
only shaft grouting be applied to P105. This is because P105 is a shaft-dominated
pile or a pure-friction pile; thus, using base grouting to reinforce the pile toe cannot
effectively increase the ultimate bearing capacity. Further, for P105, because the
transferred load from the pile toe was observed to be relatively small, increasing the
pile length would be inappropriate to increase the ultimate bearing capacity; thus,
increasing the pile diameter is recommended.
9.6 Piles Under Ultimate Load
To determine the pile behavior under ultimate loading, 4 piles were cast. Three types
of tests that lead to failure were considered. The first test involved loading the pile to
failure caused by inadequate concrete strength, the second test involved loading the
pile until failure because of the soil being soft (or the pile being overdesigned for its
capacity) and the final test involved loading the pile to failure because of eccentricity.
The behaviors of these three piles were compared with a pile that achieved the design,
and it was concluded that:
1. For the pile with inadequate rigidity, the pile will suffer from structural failure.
Under particular loads, the concrete with inadequate strength will be crushed, and
the steel reinforcement will try to resist the vertical loading, but will be ‘squeezed
out’ immediately. This structure can still resist extra loading, and soil-pile system
failure will occur soon.
2. For the pile applied with eccentricity load, the pile suffered compression and
tension in different parts. Under a certain loading, the concrete crushes and there
will be a dramatic settlement increase. Later, the tension part will elongate so
the increase of downward settlement may change to small settlement decreasing;
Lastly, the soil-pile system will fail which will lead to increase of settlement
again.
3. For the pile suffering from plunging failure, it is believed that the shaft resistance
will develop under increased loads, and, after the total shaft resistance is fully
developed, the loads will transfer to the pile end, thereby leading to the failure
of the bearing stratum.
4. For the piles suffering from ultimate loading, s-lgt curve is recommended for
analysis because this method can provide information to check where loading has
caused the pile foundation to become unstable. Similar to the previous conclusion,
the double tangent method was conservative and Chin’s method overestimated
the ultimate capacity. It is also suggested that Chin’s method should not be used
for pile capacity suffering from eccentric load.
This book provides research of different types of pile foundations which includes
cast-in-situ piles (grouted pile and non-grouted piles), precast piles (displacement and
non-displacement piles) composite piles (CFRP laminar confined piles and GFRP
bar reinforced piles) and super long and large diameter piles. The research considers
recent popular topics including grouting technique, non-displacement precast pile,
fibre material application, shaft resistance of super-long and large-diameter pile and
failure behaviour of pile foundation. Moreover, various project case studies are intro-
duced including high-rise building project and tunnelling project. This book also
covers all field tests of pile foundation in terms of static load tests with load orienta-
tion consideration (compressive, uplift and horizontal) and dynamic load test as well
as inclinometer tests.
Overall, this book has provided practical knowledge for geotechnical and foun-
dation engineering. It is expected that this book will provide academic guidance for
both foundation and geotechnical engineers. Further, the presented methods and test
data can be used to confirm the suitability of pile soil systems to satisfy the designed
load with an appropriate safety factor. This book may also provide information for
the implementation of new test analysis methods or procedures. Additionally, this
book has provided recommendations for consistent use in geotechnical practice.

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Advanced Topics in Science

and Technology in China 62

More information about this series at http://www.springer.com/series/7887

Full-Scale Field Tests

ISSN 1995-6819 ISSN 1995-6827 (electronic)

Jointly published with Zhejiang University Press

© Zhejiang University Press 2021

Gold Coast, Australia Jialin Zhou

3 Field Tests of Post Grouted Concrete Piles . . . . . . . . . . . . . . . . . . . . . . . . 81

5.3.5 Results of GFRP Bar–Reinforced Concrete Piles . . . . . . . . . 163

8.4.2 FRP Bar–Reinforced Composite Piles . . . . . . . . . . . . . . . . . . . 263

AASHTO American Association of State Highway and Transportation Officials

SRP Structurally Reinforced Plastic

N-value Standard Penetration Test (SPT) Blow Count

NB Average Corrected SPT N-value of Bearing Stratum

Ft Converted Force by Pile Driving Analyser (PDA)

Fig. 2.1 Direct shear apparatus-shear box . . . . . . . . . . . . . . . . . . . . . . . . . . 7

Fig. 2.28 Pile classification chart (Hannigan et al. 2006) . . . . . . . . . . . . . . . 29

Fig. 2.58 Determination of shaft and end resistance based on Chin’s

Fig. 3.18 Results of Q-s curvatures of uplift loaded piles . . . . . . . . . . . . . . 97

Fig. 4.32 Unit shaft resistance of P156 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 136

Fig. 6.3 Reaction system design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 177

Fig. 7.21 Q-s curve of P40 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 217

Fig. 8.30 Allowable load versus double tangent interpretation . . . . . . . . . . 254

© Zhejiang University Press 2021 1

For cast-in-situ concrete piles, this book aims to:

For precast concrete piles, this book aims to:

For composite piles, this book aims to:

1.3 Book Organization

2.1 General Introduction

© Zhejiang University Press 2021 5

2.2 Soil Stratum Determination

2.2.1 Laboratory Tests

2.2.1.1 Direct Shear Tests

Fig. 2.1 Direct shear apparatus-shear box

Fig. 2.3 Determination of

Two perforated grid plates

Bottom half of shear box Upper half of shear box

Fig. 2.4 Components of shear box

Impact data logger

Direct digital shear machine

Shear box container Loading yoke

Using weight to provide

Fig. 2.5 Direct shear apparatus

2.2.1.2 Consolidation Tests

Fig. 2.6 Horizontal movement versus shear stress

Fig. 2.7 Horizontal movement versus vertical movement

UlƟmate Shear Strength

Linear (Peak Shear Strengh)

Fig. 2.8 Normal stress versus shear stress

Table 2.1 Ultimate and peak

Fig. 2.9 Basic theory of element under stress

Fig. 2.10 Void ratio–effective stress relationship (Craig 1983)

The consolidation of soil mass involves a time-dependent change in volume. The

from dissipation of water. This process is known as ‘primary consolidation’. It has

(a) Loading device (b) Consolidation cell (Craig, 1983)

Fig. 2.11 Apparatus of consolidation test

Fig. 2.12 Typical compression versus time (AS 1289.6.6.1, 1998)

Fig. 2.13 The log time method