Civil 713 Steel

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CIVIL AND ENVIRONMENTAL ENGINEERING
Civil 713: 2020

Structures and Design 4
Volume 2
CIVIL 713
STRUCTURES AND DESIGN 4

VOLUME 2 SEMESTER 1 NOTES 2020
These Volume 2 notes contain the

following sections –
1. Overview of steel member design

2. Connections in structural steel
3. Portal frame design and detailing
4. Composite beam design
5. Checking Structures
6. Non-specific design, especially timber framed buildings
Refer Vol 1 notes for introduction to seismic dynamics; structural form and
lateral load resisting systems; reinforced concrete and masonry walls; strut and
tie design in reinforced concrete
Department of Civil & Environmental Engineering
University of Auckland
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CIVIL 713

SEMESTER 1 – 2020
SECTION 1
STRUCTURAL STEEL OVERVIEW AND DESIGN NOTES
Section 1.1: Steel Refresher Notes

copies of the slides presented. Note that the slides up
to General Principles of Connection Design will not be
covered extensively in the lectures, as students are
expected to have covered this material previously.
Section 1.2: Steel Design Review

Design of structural steel systems
Design structural calculations
Review of analysis of steel structures
This page is blank
CIVIL 713: 2020 Steel refresher slides March 2020
713 Structures and Design 4

Refresher on structural steel and
connections
By Associate Professor G Charles Clifton,

The University of Auckland
March 2020
Structural Steel Properties
The following slides give a very brief review of the

basic properties of structural steel
Basic Properties of Steel: 1

For stresses up to fy
• Steel behaves elastically
For stresses in excess of fy
• Steel deforms plastically
• Yield plateau followed by strain
hardening
Strain rate effects on fy
• fy increases with increasing strain
rate
• For blast, impact loading up to
100% increase
• For earthquake loading slight
increase (< 10%)
Page 1 of 66
Basic Properties of Steel: 2

300W, 350W =
structural low
carbon steels
480W = high
strength, low alloy
steel
700Q = quenched and

tempered steel
Modulus of Elasticity,
E is the same for all
types of steel
Structural Steel Mechanical Properties
For New Zealand, Australian, UK/European

Steels:
• chemical composition remains same for given
grade of steel, therefore:
• design yield stress, fy, varies with element
thickness
• design tensile strength is constant for grade
• see next slide for details of Australasian Steels
See more detailed information in NZS 3404 Table C2.2.1
Page 2 of 66
For I-sections, which fy to use in design?

• In tension and compression
– conservatively use the flanges as will be equal to or lower
than the webs
– or calculate weighted average value for flange and web
• In bending
– flanges control section moment capacity, so
– use fy for the flanges
• In major axis shear
– this is carried by the web, so
– use fy for the web
• In minor axis shear
– this is carried by the flanges, so
– use fy for the flanges
• However, for this course, where advised you can use
constant value as appropriate for the grade of steel –ie
fy=300 for Grade 300 steel
Structural Bolts
Two common strengths:

• 4.6  mild steel
– fy = 240 MPa, fu = 400 MPa
• 8.8  high strength
structural
– fy = 640 MPa, fu = 830 MPa Property Class 8.8
• Latter most common in
steel to steel connections
• Both used in steel to
concrete hold down
connections
Also grade 10.9 bolts;
• 10.9  highest strength
structural
– fy = 900 MPa, fu = 1000 MPa
Bolt Modes for Structural Bolts

For property class 4.6
• Snug tight mode, designated /S
– allows bolts to slip into bearing in service
– rotation can occur in service
For property class 8.8 or 10.9

• Snug tight mode, designated /S
• Fully tensioned mode, designated /T
– bolts plastically stretched when tightened
– very high clamping forces
– joints rigid under normal operating conditions
Page 3 of 66
Fully Tensioned Bolts

• First all bolts snug
tightened to pull plates
into direct contact
• Then nut position marked
• Then nut force turned by a
defined rotation
• This plastically stretches
the bolt, generating a
minimum clamping force
• Heavy duty tensioning
equipment is required
– over 1kNm of torque for
M24 bolt
– generates >300kN
tension in bolt
– heaviest generates >
1000 kN per bolt
Design modes for shear transfer through

fully tensioned bolts
In the serviceability limit state:
• /TB  tension bearing mode
– joint will very likely remain rigid
• /TF  tension friction mode
– joint guaranteed to remain rigid
In the ultimate limit state:

• /TB  tension bearing mode
– joint likely to remain rigid under
design loading
– may slip if overloaded eg by severe
earthquake
Types of Welds
• Fillet weld
• Incomplete penetration
butt weld
• Complete penetration
butt weld
• typical electrode
strengths are
– fuw = 410 MPa or
– fuw = 480 MPa
Page 4 of 66
Fillet Welds
• Simple equation for design strength
of fillet welds
• Based on ultimate shear failure
through the DTT
• welds can be weaker than members
connected
• cheapest weld to produce
• if double sided then performs well in
static and earthquake loading
• not so good in fatigue loading
• welding process can influence design
strength
Incomplete Penetration Butt Weld

• Design as for Fillet
Weld
• Good for plate to plate
connection where full
strength connection not
required
• Is always weaker than
plates being connected
• Poor performance in
fatigue
• No ductility capability
Complete Penetration Butt Weld

• Most expensive weld
to make
• Develops full
strength of the
weakest member
being connected
• Good for static and
earthquake loading
• Best weld type for
fatigue loading
Page 5 of 66
Hold Down Bolts into Concrete

• Can be property
class 4.6 or 8.8 or
AISI 4140 bar
• Can be partially
tensioned where
clamping force
against uplift or
rotation is required
• Size ranges from
16mm up to > 100
mm
• Typically galvanized
Structural Forms
Clause 4.2 of NZS 3404
Sway and Braced

Structural Systems
Sway system:
• Depends on bending stiffness and
bending strength of beams and
columns to resist lateral loads or
deflections
• Lateral displacement at ends of
columns is not prevented
(example is a portal frame)
Braced system:
• Does not depend on bending
stiffness and strength of beams and
deflections
columns is effectively prevented
(example is the central propped column in a
propped portal frame)
Page 6 of 66
NZS 3404 recognises three forms of

construction:
• rigid construction (sway or braced systems)
– connections effectively rigid up to nominal capacity of
weakest member connected
– suitable for use in moment framed seismic-resisting systems
• semi-rigid construction (sway or braced systems)
– connections develop dependable, known degree of flexural
restraint under design actions
– connections weaker than weakest member
– suitable for use in moment framed or braced framed seismic-
resisting systems
• simple construction (braced systems only)
– connections allow rotation between members without
developing design moments ,except from eccentric transfer of
shear or axial force
– suitable for use in braced framed seismic-resisting systems
Connection Characteristics
Simple (rotate under design actions without moment)
Semi-rigid (carry moment but weaker than beams)
Rigid (no rotation under design actions)
Moment-Rotation Curves for Different

Types of Connections
Illustration of connection in simple

construction
Simple
Page 7 of 66
Illustrations of connections in semi-rigid

construction
Semi-rigid
Illustration of connections in rigid

construction
Rigid
Assumptions and
Approximations For Analysis
Clause 4.3 of NZS 3404
Page 8 of 66
2 Dimensional and 3 Dimensional

Analysis
• regular building structures may be analysed as
a series of parallel two dimensional frames
• analysis carried out in each dimension
• for seismic and wind loading must consider
response of overall structure
– regular buildings may be analysed as series of 2D
frames for wind and earthquake
Design Actions in Beams of

Multi-Storey Buildings
Floor by floor analysis of rigidly connected floor

beams under vertical loading may be undertaken
in:
• braced frames
• rectangular sway frames with uniform or near
uniform loading
Model used showing variable imposed

loading requirement on continuous beams
Arrangements of Imposed (Live) Loads

for Buildings
• Covered by AS/NZS 1170.1 Clause 3.3
• For continuous beams see previous slide
• For columns see next slide; this invokes
general requirement which is:
– imposed load is applied or not applied whichever
gives the most critical actions on any structural
element
Page 9 of 66
Arrangements of Imposed Loads

for Columns: NZS 3404 Part 2
This applies for

simple, semi-rigid
and rigid systems.
For notes on
application see NZS
3404 Part 2 C4.3.1
Span Length: beam

• for a flexural member the span
length is the distance centre to
centre of the supports
• for rigid and semi-rigid structural
systems can take the design
moment as acting at the column
face
• for negative moment at the column
face this reduces the design
moment compared with the
centreline value
• for positive moment at the column
face it increases the design
moment compared with the
centreline value: this is important
for earthquake loading
where:
Mc,centreline = moment at the column centreline
= design shear force at end of beam

dc = depth of column
Span Length: column

• for a flexural member the span
length is the distance centre to
centre of the supports
• for rigid and semi-rigid
structural systems can take the
design moment as acting at the
column face
• this will reduce the column
design moment for columns
with a high moment gradient
which is of benefit for
earthquake loading
where:
Mb,centreline = moment at the beam centreline
= design shear force in column

db = depth of beam
Page 10 of 66
Simple Construction 1 of 2
• end of members free to rotate
• pinned ended for triangular
structures eg trusses
• beam reaction acts eccentric to
column, generating a design
moment, M*. Eccentricity given by:
– max (100mm or centre of bearing)
for loads into side of column
– face of column for loads into top of
column
• M* does not need to be magnified to
account for second order effects
Simple Construction: 2 of 2
For a continuous column, M*
from eccentricity of loading at
any floor shall be taken as:
a) Divided above and below by
(I/L)each way
b) Having no effect at the floor
levels above and below
c) No increase required for
second order effects
Methods of Structural
Analysis
Clauses 4.4, 4.5 and 4.6 of NZS 3404
Page 11 of 66
NZS 3404 Permits Three Methods

• Elastic analysis (Clause 4.4)
• Elastic analysis with moment or shear
redistribution (Clause 4.5)
• Plastic analysis (Clause 4.6)
More detail on the first two now follows
Elastic Analysis (Clause 4.5) 1 of 2

• Individual members assumed to remain elastic
for analysis
• Second order effects must be considered
• Member can develop design plastic moment
capacity where appropriate
• can be used for serviceability limit state design
and for ultimate limit state design
Elastic Analysis 2 of 2
• analysis may be static or dynamic
• second order effects must be considered by:

a) compliance with exclusion clauses (applies to very stiff
structural systems); or
b) first-order structural analysis with magnification of
bending moments; or
c) second-order analysis (Appendix E to NZS 3404)
• second order effects relate to a loss of elastic stiffness

of a structural system due to applied compression
loading on the members of that system
• covered below
Page 12 of 66
Elastic Analysis With Redistribution

(Clause 4.5)
• Commence with elastic analysis
• Peak moments can be reduced by
redistribution, but:
– limited by the plastic rotation capacity of the
member at the points where design moment is less
than the elastic moment
– may not require full plastic hinge development
– equilibrium between internal and external forces
required
– members must not be limited in strength by local or
lateral buckling where moment is reduced
– for ultimate limit state design consideration only
Moment Redistribution
Requires
1. Rigid supported or continuous beam
2. For the elastic bending moment along the
beam:
a. Magnitude of negative moment at ends greater than
positive moment near mid-span
b. Limits on reduction in elastic support positive
moment to avoid yielding in service
3. Cross section at support where elastic
negative bending moment is being reduced
must be stable under inelastic action
4. Aim to equalise design negative and design
positive moments in a beam span; and/or
5. Equalise design negative moments along
beams in a multi-bay frame
Example for Propped Cantilever
Elastic distribution Fully redistributed elastic Plastic collapse

distribution mechanism
Page 13 of 66
Second Order Effects
Background to Second Order Effects
Why must they be considered

• Loadings Standard requires reduced resistance
and increased deformation of the structure
resulting from lateral deflection of members to
be considered
What are they?
• Second order effects relate to a loss of
stiffness of a structural system due to applied
compression loading on the members of that
system
Two Forms of Second Order Effect

• P- effect: lateral
displacement of the
joints of a structural
system
– depends on system
response
• P- effect: member
deflection away from a
straight line between
the member’s ends
– depends on individual
member response
Page 14 of 66
Critical Parameter Involved

c= elastic buckling load factor
cN* = elastic buckling load set for the system
In words: the elastic buckling load is the set of

applied compression loads that would cause
the structural system to become elastically
unstable even if all members had fy = 
Elastic Buckling Load Factor for Braced

System
• calculated on a column
by column basis
• determined for the
critical column
Elastic Buckling Load Factor for a Sway

System
Page 15 of 66
Significance of Second Order Effects

• they cause an increase in deformation and therefore an increase in
first-order design moment
• critical parameter is c
a) when c≥10, no moment magnification is required
b) when c<10, can use first order elastic analysis and magnify the design
moments from the first order elastic analysis by
c) when c<3.5 for a braced system or 5 for a sway system, then a second
order elastic analysis is required
c=3.5 corresponds to b = 1.4 when cm = 1.0 (uniform moment)

c=5 corresponds to s = 1.2
d) NOTE: c≥3.5 is required by NZS 3404: otherwise a special study
required
Second Order Effects in Real Structures
Most structures exhibit combination of both

P- and P- effects: see example below
Second-Order
Effect
Implementation:
See Flowchart in
Student
Standard
Notes:
1. Second order effects are
applied to members carrying
compression
2. Second order multiplier ≥ 1.0
3. For a cantilever column, sway
multiplier may be > 1.0,
braced multiplier = 1.0
Page 16 of 66
Second-Order Analysis Software

• most elastic analysis programs now include a
second-order option
• this accounts for elastic P- effects
• it doesn’t always account for elastic P- effects
• if it has inelastic analysis options and large
displacement option is used then it accounts
for inelastic effects
For earthquake induced P- , use the specific
provisions of NZS 1170.5
Composite Construction – a
very brief overview
Elements of Composite Floors
Page 17 of 66
Composites – Benefits
Why use composite ?

• Utilise beneficial
properties of each
material to best
advantage
• Stiffness – 3x to 4.5x
• Strength - 1.5x to 2.5x
• Reduction in beam depth
• Lighter weight
• Improved vibration and
fire resistance
Composite vs non-composite
Types of Light
Steel Floor System
Available In New
Zealand
First and last span

onto primary beams,
simply supported
lengths
Middle two span onto
secondary beams, 2 or
3 spans continuous
Page 18 of 66
Stages of Composite
Construction
Stage 1: Steel Frame Erection and Placement of
Decking
• selfweight of steel and deck and construction loads
• restraint only at ends of secondary beams and at
points of attachment of primary beams until
decking in place
Stage 2: Pouring of Concrete
• wet concrete and construction loads
• bare steel strength only
• decking provides restraint for secondary beams
Stage 3: Composite Action
• full dead and imposed loading
• full strength of composite section obtained
• continuous lateral restraint to top flange of beams
Steel Members Subject to

Compression: Summary of
NZS 3404 Provisions and
Additional Considerations
Page 19 of 66
Determination of Section Compression

Capacity : Clause 6.2.1.1
where:
kf = form factor
An = net area of the cross section
fy = yield stress
Form Factor
If kf = 1.0, all of the cross section can reach

yield before local buckling occurs
If kf < 1.0, some elements of the cross section
will undergo local buckling before they yield
Ae = effective area is determined on an element

by element basis
Ag = gross cross section area
Element Slenderness: Flat Plate

Based on effective width concepts.
In practice the stress distribution
on a slender flat plate element is
greatest at the supported edges
and decreases away from these
The effective width concept
determines the effective width
away from points of longitudinal
edge support which can carry the
yield load and ignores the
material that is further away
than that.
See example opposite for
application to a plate with the
two vertical edges simply
supported.
A plate element is slender when
be<b for that element
Ae = bet < Ag = bt for that element
Page 20 of 66
Element Slenderness: CHS

• For a circular hollow section, the effective
diameter concept is used:
– de = effective outside diameter
– do = actual outside diameter
• When de < do, the effective diameter is used in

conjunction with the wall thickness, t, to
determine the effective section properties
Net Area of Cross Section, An

If local loss of area from bolt holes
etc does not exceed a specified
value from Clause 6.2.1.1 this
loss can be neglected.
This allows the area loss from bolt
holes in column splices such as
shown opposite to be typically
ignored and design based on Ag.
Becomes more critical for higher

strength steel because fy/fu
increases as fy increases so the
loss of area that can be
neglected decreases.
Which yield strength to use?

Minimum for any element,
or:
weighted value for each
element of the cross
section
Page 21 of 66
Nominal Member Compression

Capacity, Nc
Ns = nominal section compression capacity

c = member slenderness reduction factor
c = function of (effective length, slenderness

ratio, pattern of residual stresses, member out
of straightness)
Now addressing each of these in turn
Sway and Braced

Structural Systems
Sway system:
• Depends on bending stiffness and
bending strength of beams and
deflections
columns is not prevented
Braced system:
• Does not depend on bending
stiffness and strength of beams and
deflections
columns is effectively prevented
Effective Length: equivalent pin ended

length of the column
• NZS 3404
Section 4.8.3
• See examples
for isolated
columns in NZS
3404 Figure
4.8.3.2
• Sway member
values > 1.0
are used to
determine
second order
effects
Page 22 of 66
Effective Length of Columns in Braced

Frames
Use Eqn
4.8.3.4 and
Figure
4.8.3.3 (a)
Rotational Stiffness at Column Bases
Pinned base has some

rotational stiffness
 =  in theory;  = 10 is
realistic maximum
Fixed base has some rotational

flexibility
 = 0 in theory;  = 0.6 is
realistic minimum
Slenderness Ratio, n
Le = effective length
r = radius of gyration =
I = moment of inertia perpendicular to
buckling plane
kf = form factor (from previous slides)
fy = yield stress
Page 23 of 66
Radius of Gyration
Is a measure of how far the material in a cross section is
spread out from its centroid. See effect below on the
member compression capacity
Pattern and Extent of Residual

Compression Stresses
Determined through
the compression
member section
constant, b
The higher the

compression residual
stress, the higher
the value of b and
lower the member
compression
capacity
Effect of Member Out of Straightness
This based on
L/1000 between
points of
restraint.
Influence
included in the
NZS 3404
equations for c
Page 24 of 66
Determination of Member Compression

Capacity
Given by NZS 3404 Clause 6.3.3
Note that both Le and r are directional, in that:

• Lex and Ley may be different
• for any member other than a square hollow
section, rx and ry will be different
Thus Ncx ≠ Ncy in most cases
Discontinuous Angle, Channel and Tee

Section Compression Members
Covered by NZS 3404 Clause 6.6.
Compression load is eccentrically introduced into the
member cross section, generating combined moment and
compression loads
For members with slenderness ratio L/r > 150 the effect of
the eccentric end moment can be ignored and design
based on compression load alone.
Slenderness ratio to use depends on end connection detail

and is given by Table 6.6
For members with slenderness ratio L/r ≤ 150, the effects

of joint end moment must be determined and design for
combined moment and compression undertaken
Restraints to Compression Members

Restraints are added to increase Nc by changing
the buckling mode
Page 25 of 66
Stiffness of Restraining Elements

Bracing elements must have minimum stiffness
to be effective in changing buckled shape
m = 162EI/L3 is expression for critical stiffness
NZS 3404 Requirements for an

Individual Restraint
Each restraint must carry 2.5% of the maximum
design axial compression force in the member at
the position of the restraint .
This force must be carried back to points of

anchorage
Provided all components in the load path are

made of steel and designed to resist the
restraint forces, no explicit stiffness criterion
must be met
More Restraints than Necessary?

The restraint force can be shared
One Three
restraint intermediate
at the restraints at
centre is the quarter
needed points are
for provided;
Nc ≥Nc*, restraint force
but can be shared
between the 3
restraints
Page 26 of 66
Parallel Restrained Compression

Members
A line of restraints
restraining multiple
compression members in
parallel must be designed
for an increased restraint
force from Clause 6.7.3.
For restraint AB opposite
with 7 or more columns in
parallel:
NRAB* = 0.025N* +
*
6x0.0125N
= 0.1N*
This is the maximum
restraint force required in
parallel restrained members
Importance of Brace Connection Detail

to Compression Member
Entire cross section will move laterally when
buckling under compression loading, as shown
below for an I-section.
Both flanges of the I-section must be effectively
braced.
Method of Bracing: Direct

Bracing to Both Flanges
• Most direct method of restraining the flanges
• Axial bracing member to both flanges
• Strength requirements of NZS 3404 to restrain
2.5% of N* are applied to both flanges; each
restraint has to resist 1.25% of N*
• This shown is below
Page 27 of 66
Bracing to the Centre Line only (no

Flange Bracing)
• As shown opposite
• Brace designed for 2.5%N* and
stops cross section from lateral
movement
• but does not prevent twisting of
the cross section
• will change compression buckling
failure into torsional buckling
failure unless
NOZ > 4NOLy, where
– NOZ = elastic torsional buckling
load
– NOLy = minor axis elastic
compression buckling load
Combined Lateral and Flexural Bracing

• Direct restraint to one flange
• Flexural restraint to the other flange
• Fly-brace or stiffener to inside flange
• Cumulative axial restraint and moment restraint for
member directly connected to girt
Mbr*=0.5x0.025N*(a + d/2)
Nbr*=sum (0.025N*) as per
NZS 3404 restraint
provisions
Steel Members Subject to

Bending: Introduction,
Definitions and Section
Moduli
Page 28 of 66
Section and Member Moment Capacity
Section moment capacity  what the cross section

can deliver
Member moment capacity  what the length of member
between adjacent points of
cross section restraint can
deliver
Definition of Member for Bending

Member  length between supports (Clause
5.3.1.3(a)
Support  element preventing both in-plane
deflection and out-of-plane deflection or
twisting
Partial twist restraint Full twist restraint to

to top flange top flange
Definition of Segment for Bending

Segment  length between adjacent points of
cross section restraint (Clause 5.3.1.2(a))
Restraint (for bending)  element preventing

out-of-plane deflection or twisting of the cross
section
Page 29 of 66
Elastic and Plastic Section Modulii

Elastic section modulus = I/y
• calculated about the elastic neutral axis
Plastic section modulus = first moment of area

of cross section about neutral axis
• calculated about the plastic neutral axis
Calculated for major axis and different about

each
When the axis is not an axis of symmetry the
position of each axis is different
Example: Application to Solid

Rectangular Cross Section
Plastic Modulus for I section
About the About the

major axis minor axis
• the web
contribution
is negligible
Page 30 of 66
Application of NZS 3404 for

In-Plane Bending
The following slides cover application of the design

provisions of NZS 3404 for in-plane bending,
specifically written for application with the Student
Steel Structures Standard
In-Plane Bending: Section Compression

Capacity
This is what the cross section can deliver,
covering:
• local buckling of elements in compression
• loss of area due to eg bolt holes
• shear lag (in very wide flange sections and not
covered further in this course)
Allowance for Local Buckling of

Elements in Compression
Typically
one
flange
and the
web
Page 31 of 66
Section Slenderness Limits

Given by
NZS 3404
Table 5.2
Application
to flat plate
elements
shown
Compact Sections
Cross section can develop
full plastic action before
local buckling occurs
Ze = Zc = min (S; 1.5Z)

where:
S = plastic modulus
Z = elastic modulus
Non-Compact Sections
Cross section can reach
yield in extreme fibre
with some plasticity
developing but cannot
develop full plastic
distribution of stresses.
Ze varies between Z and
Zc, thus:
Page 32 of 66
Slender Sections
These are cross

sections where at least
one element in
compression will
undergo local buckling
before the extreme
fibre reaches yield.
The stress distribution
at design is elastic, and
Ze < Z
Influence of Holes
Local reductions in flange areas of less than
100{1-[fy/(0.85fu)]}% may be ignored
– this covers most bolted splices in grade 250, 300
and 350 members
– allows up to 18% loss of flange area in Grade 300
steel without needing to account for this loss of area
For larger holes calculate Z and S from either:

– the net section or
– (An/Ag) times value for the gross section
– An = ΣAn,flanges + Ag,web
Application of NZS 3404 for

Out-of-Plane Bending
The following slides cover application of the design

provisions of NZS 3404 for out-of-plane bending,
specifically written for application with the Student
Steel Structures Standard
Page 33 of 66
Out-of-plane Bending:
Member Moment Capacity
Design for member moment capacity, Mbx, undertaken on

a Segment by Segment basis.
Note that out-of-plane bending check is only required for
sections bent about the major principal x-axis.
For sections bent about the minor principal y-axis, section
moment capacity only, Msy, is required
Segment is defined as:

1. the distance along a member between adjacent cross
sections which are fully, partially or laterally
restrained; or
2. the distance between an unrestrained end and the
adjacent cross section which is fully or partially
restrained
Clause 5.5: Critical Flange for

Out-of-plane Bending
Definition: critical flange is the flange which in
the absence of any restraint would deflect the
furthest during buckling
Explanation of
Clause 5.5.2:
5.5.2: The critical flange of a segment with both

ends restrained shall be the compression
flange
Reason:
– compression flange is prone to buckle as a column
– tension flange is unconditionally stable
– hence the compression flange is the critical flange
Page 34 of 66
Explanation of
Clause 5.5.3.1:
5.5.3.1: When gravity loads are dominant, both

flanges of a segment with one end
unrestrained shall be considered critical
• gravity load on top, RT,

generates upsetting
couple → top flange critical
but:
• bottom flange is in
compression, therefore
also potentially unstable
Explanation of
Clause 5.5.3.2:
5.5.3.2: When wind loads dominant, critical

flange is exterior for external pressure or
interior for internal pressure
Types of Cross Section Restraints: Full

Section Restraint
Examples from Standard
Full section restraint
(F)
a)lateral deflection of
critical flange is
prevented and
twist restraint to the
section is provided; or
b)lateral deflection to
some point is provided
and
effective twist restraint
to the cross section
Page 35 of 66
Types of Cross Section Restraints:

Partial Section Restraint
Examples from Standard
Partial section
restraint (P)
a)lateral deflection of some
point is effectively
prevented and
partial twist restraint to
the section is provided; or
b)full twist restraint is

provided and
lateral deflection of the
cross section is partially
restrained
Types of Cross Section Restraints:

Lateral Restraint to the Critical Flange
Lateral Restraint (L) Example from Standard
lateral deflection of the
critical flange is effectively
prevented but
there is no twist restraint
to the cross section
If one segment end is
unrestrained the other end
must be F or P restrained,
not L
Unrestrained Cross Section
Unrestrained (U) Example from Standard
no direct restraint to
lateral deflection of
the critical flange and
no restraint against
twist of the cross
section
Page 36 of 66
Restraint:
Actual
Examples:
1 of 2
These from
HERA Report
R4-92
Restraint
Classifications
Restraint:
Actual
Examples:
2 of 2
These from
HERA Report
R4-92
Restraint
Classifications
Member Design Moment Capacity
where:
 = strength reduction factor = 0.9
Ms = section moment capacity
m = moment gradient factor ≥ 1.0
s = slenderness reduction factor ≤ 1.0
If Mb = Ms, then the segment has Full Lateral

Restraint. This is the desired condition for most
floor beams in the completed building
Page 37 of 66
Slenderness Reduction Factor, s
Graphical
Representation
Shown Below;
horizontal axis is
Msp/Moa
Note coupled nature of bending
(EIy) and torsion (GJ; EIw)
Definition of terms in Moa on next
slide
Moa: definition of terms

in equation
Geometric terms defined for an I section:

see last page of student standard for
equations for other types of cross section
• E, G = Elastic and Shear Modulus, respectively

• Iy = second moment of area about y-axis
• J = (uniform) torsion constant:
• Iw = warping
(torsion) constant:
• Le = segment effective length
Segment Effective Length, Le

Le = ktklkrL
where:
kt = twist restraint factor given in Table 5.6.3(1)
and ≥ 1.0
kl = load height factor given in Table 5.6.3(2)
and ≥ 1.0
kr = critical flange rotation in plan factor given
by Table 5.6.3(3) and clause 5.4.2.4 and
≤ 1.0
L = segment length, which is length between adjacent
restraints/supports or between a F or P restraint and the
adjacent unrestrained of a member
Page 38 of 66
Twist Restraint Factor
Effectiveness of the “goal posts” at the end of the

segment in providing effective lateral restraint of to
the critical flange
For typical I sections in buildings;

• kt = 1.0 for both ends F or L or for FU
• kt = 1.05 for one end F or L, other end P
• kt = 1.1 for both ends P
Load Height Factor
If the load is applied above the

shear centre within the segment
length it destabilises the
segment and lowers the member
moment capacity. This is shown
opposite as load RT, with load RB
being applied below the shear
centre
If RT is applied at points of restraint, kl = 1.0 as the restraint resists the

twist
If RT is applied between points of restraint, kl = 1.4 due to destabilising
effect shown above
If RB is applied between points of restraint, it is stabilising and becomes a
point of partial twist restraint.
If RT is applied at the unrestrained end of a segment with one end
unrestrained (eg at the unrestrained end of a cantilever member), kl = 2.0
Rotation Restraint in Plan Factor

• Applies to the critical
flange in plan
• Analogous to fixed and
pinned end supports of
columns in compression
In example opposite,
kr = 0.7 Top view is of critical flange
Page 39 of 66
Member Moment Gradient Factor, m
Why does it have an

influence? Simply put:
• critical flange = flange in
compression
• moment gradient
changes magnitude of
compression in critical
flange
• the quicker this reduces
between points of
restraint, the better
Values of m
1. For segments restrained at both ends
m ≥ 1.0
m is given by Clause 5.6.1.1.1(b)
• Table 5.6.1; or
• Equation 5.6.1.1(2)
Common values which are worth remembering are:

• m = 1.75 for a triangular BMD
• m = 1.35 for simple supports, central point load
• m = 1.13 for simple supports, UDL
2. For segments unrestrained at one end

m is given by Table 5.6.2 (specific cases)
m is given by HERA DCB Issue No 16 page 3 (general case), as
m,5.6.2 = 1.132 m,Eqn5.6.1.1(2) – 0.86
Special Cases:
Clause 5.6.1
Angle and Tee sections:

• Iw ≈ 0 hence GJ governs
Hollow sections:
• J >> Iw hence GJ governs
Narrow rectangular sections bent

about strong axis:
• Iw ≈ 0 hence GJ governs for Mbx
• Zex = Sx = bd2/4 for Msx
Page 40 of 66
Significant Transverse Load

Transverse load = load applied to a member or segment
between supports or restraints and acting transverse to the
longitudinal axis of a member (NZS 3404 definition)
– example: self weight of a floor beam
When can it be ignored in beam member moment capacity
determination?
Answer: when it generates a simply supported design bending
moment not exceeding 10% of the design section moment
capacity, Msx, for a segment or member with full lateral
restraint or 10% of the design member moment capacity,
Mbx, for a segment or member without full lateral restraint
Benefit: for a primary floor beam carrying loads from incoming
secondary beams, the bending moment due to beam
selfweight can be ignored in beam member moment capacity
determination (still need to include it in determining M*).
Means don’t need to determine BMD for point loads + UDL to
calculate m.
Restraining Force and

Moment Requirements: NZS 3404 Clause 5.4.3
• Similar to compression requirements
• Based on resisting 2.5% of maximum force in
critical flanges of the adjacent segments to the
point of restraint
• This is dependent on bending moments in the
adjacent segments and length of segment
over which the restrained flange is the critical
flange
• Can allow sharing of forces between closely
spaced restraints
• Restraints restraining parallel members have
same provisions as for compression members
Restraint Requirements for F or P

Restraint: Top Flange Critical
Page 41 of 66
Restraint Requirements for F or P

Restraint: Bottom Flange Critical
Design for Torsional End Restraint

NZS 3404 Clause 5.14.5
Icr for F section restraint
0.2 Icr for P section restraint
Structural Systems With Inherent

Overall Lateral Stability for Providing
Restraint
Page 42 of 66
Guidance on Restraint Requirements
• Given in HERA Report R4-92: Restraint

Classifications for Beam Member Moment
Capacity Determination to NZS 3404: 1997
NZS 3404 Application of

Shear Design
Relevant Clauses
Clause 5.9: Design of
webs
Clause 5.10:
Arrangements of
stiffeners
Clause 5.11: Nominal
shear capacity of webs
Clause 5.12:Interaction
of shear and bending
moment
Page 43 of 66
Clause 5.11.1 Shear Capacity
V* ≤ Vv is required
where:
V* = design shear force
Vv = nominal shear
capacity determined
from
– clause 5.11.2 for a web
with uniform shear
stress distribution
– clause 5.11.3 for a web
with non-uniform shear
stress distribution
Clause 5.11.2 Shear Capacity of Web

With Uniform Shear Stress Distribution
Stocky flat plate web (Clause 5.11.4):

Vvu = Vw = 0.6fyAw for a flat plate
where:
0.6fy = shear yield stress ≈fy/√3 (Von Mises)
Aw = gross area of web
=dtw for hot rolled section and d1tw for
welded section
Slender flat plate web (Clause 5.11.5):

• depends on whether unstiffened or stiffened
• see next slide
Clause 5.11.5.1 Shear Buckling Capacity

of Slender Unstiffened Flat Plate Web
Vvu = Vb = vVw
where:
v = elastic shear buckling coefficient
dp = d for hot rolled section and d1 for welded

section
Page 44 of 66
Clause 5.11.5.2 Shear Buckling Capacity

of Slender Stiffened Flat Plate Web
Vvu = Vb = vdfVw
where:
v = elastic shear buckling
coefficient for stiffened
plate
d = tension field
contribution
f = flange plastic hinge
contribution
Clause 5.11.2 Shear Capacity of Web

With Uniform Shear Stress Distribution
Circular Hollow Section Web:
This is treated as effectively
uniform.
CHS divided into quadrants.
Side quadrants take 82% of
shear and top and bottom
quadrants 82% of moment
Shear Capacity of Flat Plate With Non-

Uniform Shear Stress Distribution
General requirement is given by Clause 5.11.3:
for a rectangular flat plate:
Page 45 of 66
Interaction of Shear and Bending
Clause 5.12.2
Simplified interaction
method
Can be conservative if
shear is high, but OK for
almost all applications
EBF active link design uses
proportioning method:
• webs carry shear
• flanges carry moment
and any axial load
Block Shear
Where:
fu = nominal tensile strength of the steel supporting the bolt
group
Ant and Aev are calculated as per the figures above.
and is used with  = 0.85 to give the design block shear capacity
NZS 3404 Application of

Combined Actions
The next set of slides covers the NZS 3404

application of combined actions.
The details are from DGV1 section 6.6 and the

commentary to NZS 3404 section 8
Page 46 of 66
Covers Members Subject to Bending and

Axial Load
Must satisfy section capacity and member

capacity:
• Section capacity  capacity of the critical cross
section along the member
• Member capacity  capacity of the member
between adjacent points of support/restraint
What defines the member length over which

combined actions are checked?
Definition of Member: Clause 8.1.2.1:

Combined Bending and Compression
A member comprises:
a) The length between adjacent points of
support/restraint; or
b) The length between a point of
support/restraint and the adjacent point of
support (for bending) or restraint (for
compression); or
c) The cantilever length from the point of
support/restraint for a free-standing
cantilever
Cross section at point of support/restraint is
supported for bending and restrained for
compression
NOTE: A member may contain several segments
Definition of Member: Combined

Bending and Compression: Examples
Column case 1 comprises one member, running

from base to knee
Column case 2 comprises two members:
– the first runs from the base to girt no 2
– the second runs from the knee to girt no 2
Page 47 of 66
Definition of Member: Clause 8.1.2.1:

Combined Bending and Tension
A member comprises:
a) The length between adjacent points of
support; or
b) The cantilever length from the support for a
free-standing cantilever
Cross section at point of support is supported for
bending and for tension
NOTE: A member may contain several segments
Member With Full

Lateral Restraint
This is very important for beam-columns bent about the x-axis,

which are potentially subject to flexural-torsional buckling
Member has full lateral
restraint when all
segments along the
member have full lateral
restraint
A member expected to
become inelastic shall
have full lateral restraint
Full lateral restraint means
that section moment
capacity governs for x-
axis bending.
It means only in-plane
combined actions need
be checked for the x-axis
direction
Member Without
Full Lateral Restraint
• For the member
capacity checks, both
in-plane and out-of-
plane capacity must be
determined
• This can significantly
reduce the combined
actions capacity
compared with the in-
plane check, due to
interaction of flexural
torsional buckling and
y-axis compression
buckling
• It means that both in-
plane and out-of-plane
combined actions
checks must be made
Page 48 of 66
Significant Axial Load (Force): Clause

8.1.4
Combined actions checks are required when the
level of axial load on a member is significant.
For a member with full lateral restraint bent
about the x-axis:
• N* ≥ 0.05Ns for I and channel sections
For a member without full lateral restraint bent
about the x-axis:
• N* ≥ 0.05Ncy
For a member bent about the y-axis
• N* ≥ 0.05Ncy
Means most portal frame rafters don’t need
combined actions checks
Use of Alternative Design Provisions:

Clause 8.1.5
The clause gives specific details on which

sections are eligible for design actions checks.
The criterion for allowing combined actions

checks is simple:
• if the cross section can develop plasticity while
resisting combined actions alternative
provisions can be used
• solid rectangular sections can also use these
provisions
Design Actions: Clause 8.2

The design actions must incorporate second-
order effects as determined from section 4
(this is required in the check for bending
anyway)
For the section capacity check, the design
actions are those applying at the location of
the cross section being checked
For the member capacity check, the design
actions are the maximum actions within the
member being checked
Page 49 of 66
Format for Equations
All NZS 3404 combined actions equations calculate the reduced

moment capacity in the presence of axial load.
For example, equation 2 from the background slides is rearranged
and expressed thus:
Section Capacity:
Clause 8.3
• 8.3.2: uniaxial bending about x-axis

– general provision is linear interaction
– alternative provision based on plastic capacity of
cross section
• 8.3.3: uniaxial bending about y-axis
cross section
• can use alternative provisions for rectangular plate
• 8.3.4: biaxial bending
cross section
Member Capacity for In-Plane Behaviour:

Elastic Analysis: Clause 8.4.2
• 8.4.2.2: bending and compression

– takes no account of the beneficial effect of non-
uniform moment gradient along member
– alternative provision allows for both plasticity and
non-uniform moment gradient
• 8.4.2.3: bending and tension
– section capacity governs so cross references back to
section capacity checks
Page 50 of 66
Member Capacity for In-Plane Behaviour:

Plastic Analysis: Clause 8.4.3
• Member must be able to form a plastic hinge at the

critical cross section
– plastic hinge at ends of member : clause 8.4.3.2.1
– cross section can develop full plastic action (element
slenderness; limits on axial compression)
• when these are met

section capacity
governs, using
alternative
provisions
• Clause 8.4.3.4 is
same as 8.3.2.2 or
8.3.3.2
Member Capacity for Out-of-Plane

Behaviour: Clause 8.4.4
• 8.4.4.1: bending and compression

– takes no account of the beneficial effect of non-
uniform moment gradient along member
– alternative provision allows for both plasticity and
non-uniform moment gradient and generates much
higher capacities
• 8.4.4.2: bending and tension
– linear interaction but in this case the tension resists
lateral buckling of the compression flange in bending
so is beneficial
Biaxial Bending Capacity:

Clause 8.4.5
• Member under compression: equation is non-

linear interaction of bending reduced as
appropriate by compression, thus:
• Member under tension: equation is non-linear

interaction of bending reduced as appropriate
by compression, thus:
Page 51 of 66
Flow charts in
commentary
Clause 8.4.6: Stocky Angles

Eccentrically Loaded in Compression
• For L/ry < 150 must consider moment from
eccentric transfer of compression at member
ends when loaded through one leg
• comply with 8.4.6
• interaction of non principal axis compression
and bending
General Principles of
Connection Design
Page 52 of 66
Introduction to Bolts and

Welds; covered earlier in
these slides
Connection Design: No Inelastic
Demand or Fatigue Loading
1. Load path to be as simple and direct as
practicable
2. Determine internal forces generated in the
members being connected
• recognise primary torsion and other actions
3. Incoming force to be transferred into
components parallel to it
4. Provide for reactions when component forces
change direction
5. Design connection components and connectors
for design actions
6. Avoid use of single sided fillet welds in bending
Connection Design: Fatigue
Loading
practicable
change direction
not to exceed their fatigue endurance limit
6. Avoid weld details critically dependent on skill
of welded to achieve very high quality
Page 53 of 66
Connection Design for Inelastic
Demand: 1 of 2
Most common causes of inelastic demand are
severe earthquake or severe fire

practicable
change direction
5. Design actions based on system response
Connection Design for Inelastic
Demand: 2 of 2
6. Suppress connector only failure modes through
detailing and overstrength design
7. Don’t mix bolts and welds to carry the same
design action
8. Fillet welds must be double sided and balanced
for design actions including overstrength where
required to suppress connector failure
10.Detail connections to sustain inelastic rotation
of connected members
11.Ensure that all materials are suitably notch
tough for their in-service condition
Load Paths Through

Connections
This is the MOST IMPORTANT ASPECT of steel

connection design
Page 54 of 66
Determination of Load Path

Through Connections
• Most important part of connection design
– as illustrated from the Christchurch earthquake series
• Must understand basic principles
• Introduction to important principles follows
– more details are in the notes
Provision for Transfer of Transverse

Forces
Transverse forces must
enter that part of the
member (cross
section) which lies
parallel to the
incoming force
• if this can’t be done
then significant stress
concentrations will
occur
• requires use of
stiffeners when this
force is large
Use of Stiffeners in Transfer of

Transverse Forces
• See examples opposite
• Top case doesn’t need
stiffeners because
incoming line of
transverse force is
parallel to the web
• Middle case needs
stiffeners because
incoming line of
transverse force is
perpendicular to web
and needs to pass
through flange into
web
• Bottom case needs to
transfer force directly
to flanges
Page 55 of 66
Allowance for Component Forces

• When an applied
internal force changes
direction, component
forces are introduced
into the connected
elements
• The effect of these
must be designed for
• Example opposite is
portal frame knee
joint
Importance of Load Path: Seismic Hold

Down Connection for Thin Walled Silos
• Seismic actions
generate high tension
forces in hold down
bolts as they resist
uplift
• These tension forces
must be transferred
into the walls of the
silo without
overloading the walls
• Load path is complex
Welded Moment
Resisting Connections
Page 56 of 66
Bolted Moment End Plate Connection:

Load Paths and Failure Modes
See notes for design procedures; more detailed are in

SCNZ Steel Connect and HERA Report R4-142
Page 57 of 66
Three Types of Connections

Simple (rotate under design actions without moment)
Semi-rigid (carry moment but weaker than beams)
Rigid (no rotation under design actions)
Moment-Rotation Curves for Different

Examples of Simple
Connections
Web Plate
• Carries vertical loading
• Failure mode is web plate
in-plane yielding and bolt
hole elongation
• Bolt and weld failure is
suppressed
• Very limited axial load
capacity
• High rotation and tension
pull-in capacity
• No design moment
capacity
Page 58 of 66
Flexible End Plate

• Carries vertical loading
• Failure mode is end plate
out of plane bending
• Bolt and weld failure is
suppressed
• Reasonable compression
capacity
• High rotation and tension
pull-in capacity
• No design moment capacity
Examples of Semi-rigid
Connections
Sliding Hinge Joint

• Rigid elastic behaviour to
defined moment
• Then can rotate elastically
by sliding at base of joint
• Can accommodate high
rotation demands with
minimal degradation
Page 59 of 66
Sliding Hinge Joint general view
Action of the Sliding Hinge Joint in Severe Earthquakes
Structure sways to right
Structure sways to left
Page 60 of 66
Examples of Rigid
Connections
Welded Moment Connection

• Beam to column
example shown
below and opposite
Bolted Moment End Plate

• Connection can be
designed to be weaker
or stronger than the
beam
• Failure mode is by
endplate and or column
flange out of plane
bending
• Bolt or weld failure
suppressed
• Complex load paths and
design procedure
Page 61 of 66
Moment Resisting Beam to Column

Connection into Concrete Filled Tubular
Column
Moment Resisting Beam to Column

Connection with Flange Bolted External
Diaphragm Plates
Examples of Splices
Page 62 of 66
Beam Moment-Resisting Splice

• Flanges resist
moment-induced
and direct axial
forces
• Web resists shear
and moment due
to eccentric shear
transfer
• Bolt and weld
failure suppressed
Column Moment-Resisting Splice

• Flanges resist
moment-induced and
direct axial forces
• Web resists shear and
moment due to
eccentric shear
transfer
• Bolt and weld failure
suppressed
• Compression forces
can be resisted by
end bearing
Appropriate Choice of Cost-

Effective Connections
Page 63 of 66
Appropriate Choice of Connections

Connections represent:
• Up to 40% of design time and
• Between 15% and 55% of erected steelwork cost
and
• Significant impact on speed of construction
Important factors to consider are:
• What type is needed for structural performance
• Connection to be assembled in shop or on site
• How skilled is labour
• How accessible is the connection
• What degree of supervision is available
Designing for Cost-Effective Fabrication
1. Consider ease of erection in design

2. In a flooring system comprising primary and
secondary beams
1. detail secondary beams for cutting to length and
bolt holing only
2. weld cleats to primary beams only
3. For moment-resisting steel frames
1. detail beams for bolt holing only
2. detail columns to receive welded components
Economical Choice of Connections

Two main principles:
1. Maximise shop welding and maximise site bolting
2. Prefabricate off-site as much as possible and
when on site
prefabricate at easy working level as much as
possible
Four lesser principles:

1. Use simplicity in framing and connection design
2. Use symmetry where possible
3. Use repetition where possible
4. Use industry developed standard details were
available
Page 64 of 66
Yieldline Theory
Required when flat plates are loaded in tension out-of-

plane
Examples of application: 1 of 2
Close-up of baseplate
Baseplate of sign showing yieldline showing
yieldline failure pattern separation at bolt line
Examples of application: 2 of 2
Close-up of baseplate
showing yieldline from
column base rotation
Low damage hybrid timber-steel

low damage moment frame
specimen on shake table
Page 65 of 66
Design Based on
Equivalent Tee Stub
Three failure modes

1. complete flange
yielding
2. bolt stretching with

flange yielding
3. bolt stretching no
flange yielding
Effective Length
of Tee Stub
Determined from tables for real applications, eg
Page 66 of 66
Civil 713 Structures and Design 4 - 2020 - Section 1.2 Steel Design Review
1.2.1 Design of Structural Steel Systems –
Concept of Design - To produce a solution to a problem which meets

technical criteria, is cost effective, available, practical
to produce and satisfies the needs of the end user
Meeting Technical Criteria - Means the solution should exceed the minimum
requirements under serviceability and ultimate load
conditions expected during the life of the structure – refer
to NZ Standards, Codes of Practice, Text Books, Local
custom
Cost effective - Means the solution achieves value for money in both final
operating and maintenance costs throughout the life of the
structure
Available - Means readily available on the current market without

substantial cost or delay in production and delivery
Practical to Produce - Means the designer has understood the practical features
of the design in producing the solution within the
environment and achieving the intent of the engineer
Satisfying the Needs of the End User - Means meeting specific criteria of the Client
with regard to the specific use of the article or product
notwithstanding having to meet NZ Building By-
laws, Standards, Codes of Practice and Common Law
This course in Structural Steel Design –
1. Sets out to address the Design of Steel Components of Buildings to meet

Student Standard criteria
2. Addresses the Design of Connections in Structural Steel
3. Briefly covers Building Systems, Corrosion Protection, and
Construction Techniques
Section 1.2 – Page 1

Civil 713 Structures and Design 4 - 2020 - Steel Design Review
1.2.2 Design Structural Calculations –
a. In this course within your assignments, you will be asked to perform design
calculations for various structures. The intent is to liken the experience to producing a
set of calculations for a design office and NOT answering a series of questions.
Accordingly, as in a design office, the calculations must be self explanatory, complete
and tell the whole story. The student must put her/himself into the position of not
knowing anything about the structure and for the calculations to be read by someone
who knows very little about it. Significant percentage of the marks will be allocated for
layout and clarity of calculations as this (as you will see below) is an important part of
your communication skills in the design office. Some students use these calculations
as examples of your work in a job interview. Outstanding, well drawn up calculations
can be an extremely good example of what the employer can expect from an
interviewee
b. Calculations are a story to be told, in sequence and brimming with clarity. The main
requirements for calculations are to describe the basis of any design, the thought
processes behind the design (of the whole structure, the elements of the structure and
the connections), to confirm the documents referred to, for future referral by a
draftsperson in drawing the structure and connections, for writing a specification or for
checking purposes, later alterations to the structure (either during construction or
much later when changes to the building are required by a Client
c. Are generally carried out on squared paper where sketches to scale can be made
d. Generally commence (after the title page and contents (including page numbers) with
a (to reasonable scale) sketch of the general arrangement (GA) of the structure to “set
the scene”. Referrals then to “Beam B23” or “connection brace BR12 to Column base
C5” should be clearly seen on the GA. In addition use subtitles regularly to allow the
checker to follow your progress through the calculation. Eg “Design Beam A-B”,
“Summary of design actions”, “Design for strength”, “Moment”, “Shear”, “Bearing”,
“Serviceability” etc
e. Using MathCad, Spreadsheet or Computer printouts for calculations can be fraught

with difficulties principally because so much is hidden from view which means doing a
quick check is almost impossible. As Item b above, calculations are a story to be told
and understood by others later on. They are a line of communication which needs to
spell out the intent and process of the mind of the engineer in achieving a result for a
particular application. In this course Mathcad or spreadsheet will only be accepted if
substantial attempt is made to describe what is being done, from where values have
been obtained and the formulae involved
f. Do not use a figure from a previous page or section without referral to the page
number from where you obtain the number – eg φMb = 23.4 kNm > M* = 19.8 kNm
(see p 12). This allows you or a colleague at a later date to easily determine from
where that information comes from. Too often glaring errors are made by extracting a
figure from a previous page, a computer printout or a chart in a document, which is the
wrong figure
g. In addition, equations from a design guide or a NZ Standard should be referred to,

again to make sure the correct section of the Standard or design guide is being
used, eg Combined bending & axial ref NZS3404 (or the student standard) Eqn
8.4.4.1 etc

h. Each section of the calculation must be complete – that is, it reaches a firm conclusion
– eg in the design of a beam you may “Try” various sizes and eventually “Adopt” the
final size – that is, close the section with “Adopt 310UB40 for Beam05 (same for all
frames and levels 03 to 05)” et
i. In computer analysis let the computer sort out the load combinations – students
should set out the basic G, ϕQ, E and W cases and allow the computer to do the rest.
Summaries of the basic loads should be clear in your calculations and should be
referred to when doing the computer output review mentioned in (k) below
j. Computer printout need normally not be fully included in the calculations but referred
to as an appendix. However the basic input needs to be clear and a summary of the
output should be shown from which specific items or diagrams (such as bending
moment or shear force) can be referred to from time to time within the calculations.
Check thoroughly the values used in the input – any significant errors here make a
nonsense of the subsequent design (refer Invercargill Stadium Construction problems)
k. Any computer output should be reviewed carefully for accuracy (within the
calculations) by doing quick hand checks. Many errors in design are made by
believing what comes out of a computer or calculator is correct when it may be far
from correct – It is essential that computer printouts are checked for correctness by
selecting say 2 critical combinations and for each, carry out ΣH = 0 and ΣV = 0 and
checking that the effective
simply supported bending moment between the ends of a beam match the figure
obtained by hand – for example a vertical ULS UDL of 10 kN/m on a beam (within a
frame say) 5.6 m long should reflect in the computer output with an effective SS BM of
39.2 kNm. If the end moments from the frame analysis in the computer output show -
12.3 kNm (LHS) and -18.9 kNm (RHS) and the midspan moment as any figure very
much different from + 24.6 kNm, then something is definitely wrong. Note that a
symmetrically loaded and geometrically symmetric frame will automatically show ΣH =
0 so this is not really a check at all for accuracy of input and analysis
l. Summarize design actions on the element being designed at the start of each section.
This sets the scene and provides an “opening sentence” to this part of your
calculations. It shows you have addressed all combinations and you are preparing to
select which combinations should be addressed in design. Sketch the local BMD and
SFD. This clarifies in the reader’s mind the beam or column and its full design
actions. Its only when you transfer the design actions from the computer output to the
calculations sheet that you really understand the various actions being imposed on the
member being designed
m. Sketches - Each section which shows calculation of a connection should be complete

in itself and end up with a scaled sketch of the connection showing everything needed
by the draftsperson. Another engineer may be checking the final drawing against the
calculations and it must be clear the intent of the design engineer. All dimensions
need to be shown such as edge and spacing distances to bolts. You must be clear
where dimensions start and finish. A cardinal sin is to place on the sketch something
which is incorrect from the calculations. Double check this – it is nonsense to
calculate an M24 8.8/S bolt is needed and show on the sketch an M20 bolt! An
incompetent computer driven draftsperson may insert a sketch from a previous job on
the drawing and not for example change the cleat thickness or the bolt size. You will
not be praised for getting it wrong in incorrectly transferring design information onto
your sketch or not double checking it is correct on the final drawing. Major problems
can arise on site from errors in drawings with remedial work needing doing and or a
failure resulting. All welding needs to be shown and be complete. Don’t split
sketches, eg a sketch of the flange plates and a separate sketch for the web plate in a

simple beam-beam connection. This can create errors in transfer of information. In
addition only when combining both into a final sketch do you find that interference may
occur between bolts. If you have shown a sketch earlier which is superseded by a
final sketch scrawl “superseded” across the first sketch to ensure a draftsperson does
not pick up incorrect information from an earlier versio
n. Obtaining information from a sketch or photocopied drawing can create a problem.

The rule is don’t scale – but if there is no alternative and you need to scale a
dimension, you have to take extra care. First establish what scale the drawing drawn
at. Photocopies may not scale well and be correct – if you know two or three
dimensions as factual on the drawing you can scale that and use that as a base for
measuring others. At the end of the day nothing beats visiting the site and taking site
dimensions
o. In your assignments, clarify the comparison of “strength to actual” within the

calculations – for example φMb = 342 kNm > M* OK leaves the checker with some
information but not enough to give the whole story. You as the designer (as you write
this) remember or know that M* = 132 kNm but that may be buried further back in the
calculations and may be obscured by several M* ‘s being considered for this design. It
is much better to write φMb = 342 kNm > M* (132 kNm) OK. This gives you much
more information and you can see that if appropriate a recycle may be worth
contemplating. It is only when you see these comparisons that you may double check
that the M* is the correct figure extracted from previous pages and that there may be
more work needed for cost effectiveness
p. Recycling means looking at a stronger or weaker component to satisfy the end

condition. Do not recycle by rewriting calculations on continuing pages – provided you
spread out the original calculation a recycling process using forms of brackets Try 4
M20 8.8/S Bolts {4 - M24 8.8/S} [6 M20 8.8/S] is a progression that can be carried
through a calculation quite easily provided you have left room. Saves time, paper and
provides you with double checks as you recycle through. Often in that process you
may find an error in your original calculation which makes the original selection OK
q. Don’t reach a result then not comment on it – if in bearing, the capacity of the web in
crushing is say 980 kN don’t just leave it at that, the line should read for example
“Bearing Capacity = 980 kN> R* = 632 kN Accept as OK in crushing”
r. Making an assumption for self weight is best as early as possible and is probably a
nuisance to add in at a later occasion – note that in SAP the self weight of the beam
allocated will be automatically added. For a bridge beam whether or not you select 80
or 110 kg/m for the self weight of the beam is not going to make much difference to
the final result and can be reviewed near the end of the calculations as to the actual
weight and its impact (if any) on your results
s. With combination M/EI diagrams which require calculations under serviceability (for
point loads), it is worth looking at splitting into components and using standard formula
or easy moment area equations to achieve part displacements and add up the parts to
achieve the whole
t. In Floors the area reduction factor to Q ALWAYS applies (including during the
assessment of seismic weight). Ensure you read and thoroughly understand the Notes
in A3.1.1 NZS1170 Student version which provide restrictions to the use of ψa. The
area reduction factor ψa applies differently to each elemental design particularly to skip
loading and seismic weight

u. All preliminary sketches should be drawn to a scale and in proportion – only this will
allow the student to understand the significance of size – it is nonsense to design for
three bolts in a vertical row in a horizontal 150UC – there is not enough room and a
scaled sketch would clarify this from the outset
v. Many students are attracted to the “to be conservative” comment when judging
whether or not a design action can be carried or not. Be careful – while engineers
need to be cautious when applying design actions, the rules are sufficiently robust and
are generally based on historical performance. Don’t be unnecessarily “over-
conservative” to the extent of costing your client extra money
1.2.3 Review of Analysis of Steel Structures (in a bit more

detail than in the refresher slides)
1. Simple Steel Structures –
Simply supported statically determinate beams, trusses and columns require

very little analysis and can normally be analysed by hand. Load combinations
for strength design and serviceability calculations can be carried out after
design actions have been determined from simple analysis. Examples of these
are window lintels, non continuous crane beams, diagonally braced structures,
cantilever posts, simply supported floor beams etc
2. Continuous Steel Structures –
a) Continuous beams can be simply analysed by hand or computer. For simple

loadings, hand analysis is quick. Remember to check overall V  0 as a final
check on accuracy of your analysis. Examples of these are continuous crane
rails, beams continuous over many spans. Note – skip floor loading need only
be applied when considering the gravity combination of loads. As soon as
wind, fire or seismic actions enter into the combination assume the reduced Q*
applies to all spans.
b) Simply supported pitched or parallel chord roof trusses can be simplified by
assuming pinjoints and analyzing the structure by truss analysis. If point or
uniformly distributed loads are applied to parts of the truss between supports,
then combined actions must be checked, that is the member attracts an axial
load (from truss action) and some design bending action due to the local load
being applied between supports. Examples of these are wind trusses in the
plane of the roof and the many styles of roof trusses where depth of structure
is available.
c) If trusses are constructed in adjacent spans, take care with the assumption of
them being simply supported as actual construction and continuity of the
trusses through the supports may in effect allow them to act as continuous and
this can create differing design actions.
d) Floor trusses or heavily loaded bridge trusses can be similarly analysed
assuming pinjoints, but substantial care needs to be taken when detailing the
connections to ensure this assumption is reasonable. Transfer of heavy loads
through connections where, because of construction difficulties, eccentricities
may occur, can prove a problem to the intersecting members with local
moments due to these eccentricities being significant

3 Assuming Full Fixity in Truss/Beam Design –
Full fixity at beam or truss ends is usually very difficult to achieve in practice
and should not be assumed unless the connections and load paths through the
supporting structure are fully understood and designed for. Any small rotation
of the supporting member or the connection will render “full fixity” null and void.
Structural failures have occurred because designers assumed fixity when in
reality none was achieved
4. Braced Steel Structures (triangulated structures or frames braced against

sidesway)
a) These are termed “Simple Construction” in Cl 4.3.4 NZS 3404 and are
analysed assuming all members are pin connected. This also applies to
structures which rely on other (separate) structure for lateral support such
as a lean to frame, mezzanine floor structure or the externals of a multi
storey office building where all lateral loads are carried by the central
reinforced concrete core
b) Beams or bracing transfer axial loads through “pin-jointed” connections to
columns which need to be designed for the additional moments induced by
the eccentricity of the connection
c) The eccentricity to be allowed in design for a normal side cleat connection
is 100 mm from the face of the column or the actual eccentricity whichever
is the greater
d) If the beam passes over the column with a column cap the eccentricity may
be applied at the face of the column
e) For a column continuous (above and below) through the connection, the
moment induced by this eccentricity can be assumed shared by the
columns above and below the connection, in proportion to the values of the
/ of each column above and below the connection. This moment is not
carried any further into the floor, beams or column actions at any other level
f) Refer to Section 1, Page 6 herein for taking account of second order effects
principally due to (caused by the local deformation of member from its
original shape)
g) Having considered second-order effects as above, then the element design
of the members in compression become easier. The effective length factor
for all columns becomes 1.0 and no consideration for extended effective
lengths as Student Code 4.8.3 requires is necessary
5. Steel Sway Framed Structures –
a) Framed steel structures subject to sway require analysis to determine

design actions. A first-order analysis does not make any allowance for
second order effects such as p-delta (either due to sway effects ∆ or
flexible columns ). These effects are allowed for by carrying out a
second order frame analysis (by computer) or by taking the values of
moments from a first order analysis and magnifying the moments to allow
for these effects

b) Most computer structural analysis programs these days allow for second
order effects in which case analysis (at least for structures subject to non-
seismic loads combinations) can be determined and no moment
magnification for second order effect is necessary
c) For sway structures subject to seismic load combinations a first order
analysis is required and the second-order effects taken account of via the
Loadings Standard rules NZS1170 Cl 6.5
d) For sway structures subject to non-seismic load combinations, either carry
out a first order frame analysis and determine second order effects
(including moment magnification) as below, or perform a second order
frame analysis direct in which case moment magnification is unnecessary
e) Frames under the Design Action including Seismic Loads come under
special design criteria for analysis, second order effects and ductility
demands. These criteria given in Chapter 12 of the Steel Standard are not
part of this University Course but must be taken into account in all
Structural Steel Design in the Design Office
f) Having considered second-order effects as above, then the element design
of the members in compression become easier. The effective length factor
for all columns becomes 1.0 and no consideration for extended effective
lengths as Student Code 4.8.3 requires is necessary
If you need to calculate second order effects by hand calculation to no sway and sway
frames please make reference to the SESOC Design Guide (Simplified Design
of Steel Members) Section 7 and the flowchart in the NZS 3404 commentary.
Web Crushing & Buckling – refer Student Standard Clauses

5.13 – 5.14 inclusive
Webs in areas where high reaction loads can bring substantial compression loads on
the web require checking for elastic buckling and crushing and may require stiffening –
see Figure 5.13.4 of NZS 3404. Note that the sketches shown on Student
Code Fig 5.13.1.1 show a concentrated applied load imposing vertical gravity
loads on the beam and thereby possibly overstressing the beam. In the case of
a bridge abutment or support condition this effect is upside down – that is the beam
is being supported (via a reaction) on a bearing plate and the applied loads are
therefore upwards. Some students and designers seem to invariably get this
wrong …
Example of a Stiffener Design at an End Support
Check 610UB101 supported by a 100 mm bearing plate
carrying an ULS reaction R* of 630kN for local effects of
web crushing and buckling. Design any stiffner/s if needed
ϕ = 0.9 and RB = 1.25 bbf.tw.fy
Check Bearing (5.13.2 and 3) R* < ϕ RB
fy = 300MPa, .tw = 10.6mm

6mm bbf = bs + 2 * 2.5 tf (if
available)
tf = 14.8mm 2.5 tf = 37mm Then bbf = 100 + 10 + 37 =

147mm

ϕ Rb = 0.9 * 1.25 * 147 * 10.6 * 300 = 526kN < R* = 630kN Fails in Bearing
Check Buckling (5.13.4) Web acts as a column – use Section 6
αb = 0.5 kf = 1.0 d1 = d2 = 572mm kf = 1.0 ϕ = 0.9
.
√ 148 ; Table 6.3.3(2) αc = 0.279; bbb = 100 + 10 + 37 + 572/2 =
.
433mm
ϕRbb = 0.9 * 1.0 * 0.279 * 433 * 10.6 * 300 = 346 kN < R* = 630kN Fails in Buckling
Both actions require a stiffner – Available distance from edge of web to edge of flange
is 109 mm
Try an 80 by 10 stiffner each side of the web welded as shown
Check maximum outstand of web (5.14.3) 137 > 80 OK
Check bearing – bearing area increased by 2 * 80 * 10 = 1600mm2
Then ϕ Rsb = 0.9 * 1.25 * (147*10.6+1600) * 300 = 1066kN > R* = 630kN OK in

Bearing
Check buckling (5.14.2) Combined action of additional stiffner as a cruciform column -

use 5.14.2.2 to find length of web effective when the combined section is acting as a
column (in buckling)
.
Effective maximum length of web given by 169 on one side
The other side (distance to the edge) is 60mm – bb = 169 + 10 + 60 = 239mm
Radius of gyration of the combined section about the centerline of the web is given by
where 10 170.6 4.14 6 (ignoring the web)
And 1600 239 10.6 4133 r = 31.6mm Le = 0.7 X 572 = 400mm
√ 13.9 From Table 6.3.3(2) αc = 0.993 (αb = 0.5 kf = 1.0)

.
Then ϕRsbb = 0.9 * 1.0 * 0.993 * 4133 * 300 = 1110 kN > R* = 630kN OK in Buckling
ADOPT an 80 by 10mm flat bar stiffner welded to each side of the web centreline of
the 100mm bearing plate (Weld 5mm fillet full profile)
Section 1.2– Page 8

Members Subject to Combined Actions of Bending & Axial Load–
Eccentrically Axially Loaded Columns, Portal Frame Columns, End
Wall Columns carrying axial load and lateral loading (wind)
• Chapter 8 of the Student Code

• Steel members in combined axial load and bending termed beam-columns,
where the principal action of carrying the load is by a combination of flexing
and compression
• Steel members in combined bending and axial load must be analysed for the
individual capacities in each direction and then apply the Interaction Formula
• there is a BIG difference in capacity between a beam-column having full lateral
restraint in bending and one not having full lateral restraint in bending; see
section 1.1, page 48 for the reasons why.
Check Section Capacity φ Ns . Clause 6.2.1
Check Effective Length kle, and αc about both axes, and therefore determine φNcx and φNcy
Clause 6.3.3
In Combination with Bending in one direction –
a) Major axis bending – without full lateral restraint –
N* M*
  1.0 Cl 8.4.4.1
N cy M bx
weak axis buckling
b) Minor axis bending – without full lateral restraint
N* My *
  1.0 Cl 8.4.4.1
N s M sy
Both Section Capacity
Biaxial Bending –
Is created when columns are subjected to bending moments in two directions

or when a column is subjected to an eccentric axial load, that is load is applied
offcentre in both x-x and y-y directions. BOTH conditions shown below must apply -
Mx* My* N*
   1 .0
 M sx  M sy  N s
Where all denominator capacities are Section Capacities

AND
1.4 1.4
 M x *  M y *
     1.0 Cl 8.4.5.1
M rx  M ry 
Where Mbx (bending capacity) and Msy (section capacity) are modified as
 N*
M rx  M bx 1   Cl 8.4.4.1
 N cy 
and
 N*
M ry  M sy 1   Cl 8.4.4.1
 N cy 
Taking effective column length factor ke as 1 for sway frame columns and
as 1.0 or less for braced columns

CIVIL 713

SEMESTER 1 – 2020
SECTION 2 – Structural Steel Connections
Section 2.1 Review and New Material

2.1.0.General introduction 1
2.1.1. Bolted Connections 2
2.1.2. Welded Connections 12
2.1.3. Minimum Design Actions 21
2.1.4. Web Plate Connections 25
2.1.5. Moment End Plate Connections 29
2.1.6. Column Baseplate Design, Pinned 31
2.1.7. Design of Anchor Bolts in Shear 35
2.1.8. Splices 39
Section 2.2: Welded and Bolted Moment Resisting Connections

(extracts from a HERA report so don’t start at section 1)
• Welded Moment Resisting Connections
• Bolted Moment Resisting Endplate Connections
• MEP design example
• References
Section 2.3: SCNZ Design Procedure for MEPS: Moment End
Plate Splice: Flush
Section 2.4: Design of Moment Resisting Column Baseplate

Connections
(extracts from HERA Steel Design and Construction Bulletin Issue No 56)
This page is blank
713 Structures and Design 4 – 2020 – Section 2.1 Steel Connections General
________________________________________________________________________
Section 2.1 Structural Steel Connections – Review Civil 313

material and new material on connections.
SPECIAL NOTE - Most students in the Class will have done Civil 313. These notes
assume familiarity with the material delivered in Section 9 and 10 of the Civil 313
Coursebook, a summary of which is in the refresher notes in section 1 of this volume
and on CANVAS. For those that don’t have this information please refer to the Civil
313 notes. Note also that connection design was not covered in the 2019 Civil 313
lectures to the extent that is presented in the Civil 313 notes.
2.0 CONNECTIONS GENERAL INTRODUCTION
Connections – join members together into structural systems.
Shop Connections – made in the fabrication shop
Site Connections – made on site
Need - Change of Direction

Framing in
One member to another – (Purlin – Rafter)
• Importance of security and safety
• Importance of Design
• Importance of clarity in what the Engineer wants
• Importance of ensuring the Engineer gets what he/she wants
Apply the two CD’s – Concept – Design – Communicate – Diligence
Factors governing the selection of the type of connection
1. Must the connection be rigid (no slip) for operational reasons?

2. Can the connection be allowed to slip under severe conditions?
3. Is the connection to be carried out in the field or in the shop?
4. Do you want a high quality appearance or are aesthetics secondary?
5. Have you the room to fit the type of connection being considered?
6. What degree of supervision have you available?
7. Is the connection to be carried out by skilled or semi-skilled labour?
8. Is the connection accessible?
In general, connections made in the fabrication shop should be welded and items of
structure fabricated to such a size that they can be transported on the tray of a truck
and erected easily, or that they can be laid in a galvanizing bath.
The shop welded connection gives the cheapest solution for a connection with
minimum room used and high quality appearance. Shop supervision can be of high
quality (specification) requiring highly skilled and ticketed labour under the eyes of a
competent shop foreman. Prepared sections can be inspected easily and ultrasonic
or other subsurface inspection techniques can be cost effectively used if and where
necessary
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Section 2.1 - Page 1
________________________________________________________________________
The field connection (ie done on site) is the area of most contention – which is best:
bolts or welding?
The following table summarizes the advantages and disadvantages of field

connections in welded and bolted joints :
BOLTED WELDED
Labour required Semi-skilled Skilled

Weather Protection Nil Temporary
Scaffolding Sometimes required Usually required
Temporary fixing Usually Always
Speed of erection Faster Slower
Immediate Corrosion Protection None Required
Damage to galvanized elements None Substantial
Inspection time Little Varies depending
on degree
Inspection Labour Semi-skilled Technician
Cost of Inspection Low High
Overall Cost Usually cheaper Expensive
Finished appearance Clumsy Excellent
Size of connection Large Small
Replacement of unsatisfactory
Work Easy to unbolt Difficult
The above table would tend to favour bolting in almost all situations, but each
problem must be considered on its own merit - the general rule is to weld in the
shop and bolt on site. Some types of connections, such as splices between lengths
of structural hollow section columns, are usually welded on site.
2.1 BOLTS - Three property classes of bolts available in New Zealand. These
used to be referred to as grades and this is still a commonly used term.
Mild Steel Commercial Bolts – Property Class (PC) 4.6 where the 4
represents a hundredth of the tensile strength (400 MPa) and the 0.6 the fraction of
the Yield Strength to Tensile Strength (240 MPa) – Available painted black, hot dip
galvanized or stainless steel
High Strength Structural Bolts – PC 8.8 – (830 MPa and 660 MPa
being the tensile and yield strength respectively) and PC 10.9 (1000 MPa and 900
MPa being the tensile and yield strength respectively) – Generally available in hot
dip galvanized. PC 8.8 bolts for structural purposes have heavy heads and
oversized nuts so they can be dependably fully tensioned to fracture of the bolt, as
do the PC 10.9 bolts. This enables them to be used in the fully tensioned mode.
There are three modes of tightening specified for bolts and these affect the way the
item is designed and affects the design strength of the connection
a) Snug tight (/S)

b) Tension Bearing (/TB)
c) Full tensioning (or tension friction), (/TF)
________________________________________________________________________
________________________________________________________________________
All property classes are used in a snug tight mode where the reason the bolt is there
is to transfer load as a dowel (in shear) by bearing between the bolt surfaces and the
cleat or web – snug tightening is the full procedure for this mode, sufficient for
structural purposes
Full tightening is only used with PC 8.8 or PC 10.9 bolts where the full tightening
procedure squeezes the mating plates together and forms a friction force effectively
transferring the load by friction between the faces and not in bearing - /TB allows
some slip in the joint and offers a higher design load than /TF which guarantees no
slip in the joint at serviceability loads and therefore has a downgraded design
capacity. Examples are in Cranes, Bridges, longspan splices and structures where
dynamic loads are present and that any slip would harm the structure, cause
misalignment and can lead to fatigue failures. Friction connections provide the best
load distribution across the joint, but are very expensive due to the additional labour
for construction and inspection and the consideration of the quality of the joint
interfaces. Splices in eg portal frame rafters must have /TB bolts to prevent rotation
under operational loads.
SHEAR & BEARING MODES –
If tightened correctly slip only occurs at full design

load in /TB mode and joint remains rigid (ie no slip)
at design load in /TF mode
Snug tight mode is the most economical using PC

8.8 bolts (where some slip is allowed) because they
give about 2.5 times greater capacity than PC 4,6 at
only about 50% increase in cost. PC10.9 bolts are
more expensive but give higher design capacity
Snug-tight /S (as defined in the NZ Standard) is the

tightening of a bolt achieved by a few impacts from
an impact wrench or the full effort of a person using
a standard podger spanner (which has a given
length for various diameters of bolt). It is the full
procedure for a structural bolt in snug tight mode or
the intermediate level of tightening prior to the full
tensioning in the /TF and /TB modes
Fully Tensioned Modes /T (as defined in the NZ Standard) where only high
strength friction grip bolts (to AS1252) should be used. Designers select /TB for
bearing type joints (some slip may occur) or /TF for fully rigid joints. Both modes rely
on the friction between the mating surfaces to transfer the load initially – in the
bearing mode it is assumed the bearing surfaces eventually come into play at the
ultimate limit state and the design is similar to /S design; in the friction mode
designers check the serviceability limit state for no-slip and the ultimate limit state in
bearing. In addition care must be taken at the design and construction stages to
ensure the surface condition of the interfaces matches the coefficient of friction
assumed at the design time
________________________________________________________________________
________________________________________________________________________
Fully tensioning involves initial snug tightening, then either continuing tightening with
an impact wrench using a load indicating washer, or part turn of the nut method
where the nut is turned one half a turn after proper snug tightening – refer to the
supporting information on CANVAS.
There is also the requirement in the fabrication and erection standard that for any
bolts/nut assemblies the nuts shall run freely and shall be checked by running the
nut along the bolts by hand before being used in a connection. This captures and
rejects bolt/nut assemblies with poor surface finish that will undergo plastic torsion
during bolt tightening instead of achieving plasticity in tension.
Stress Areas for Bolts to be used in Design -
Nominal
Diameter Pitch Areas (mm2)
(mm) (mm)
Ac As Ao
df p Core Tensile Shank
Stress
12 1.75 76.2 84.3 113
16 2 144 157 201
20 2.5 225 245 314
24 3 324 353 452
30 3.5 519 561 706
36 4 759 817 1016
The maximum shear force a bolt can sustain is dependent on the core (Ac) or the
shank (Ao) area depending on the position of the shear plane, while the maximum
tension force depends on the tensile stress area (As). These figures are repeated in
the student Code Table C7.2 (Page 109). This table needs a note added that these
are Tensile Stress areas to round bars “threaded to AS1275”. This is an ISO metric
cut thread which is the standard thread type used in steel construction.
Bolt Strength under Limit State

Bolts in Simple Shear
Refer Student Code of Practice Section 9.3.2
A bolt subject to an ultimate design shear force V* shall satisfy –
V* ≤ φ Vf ≤ φ x 0.62 x fuf x fr x (nnAc + nxAo)
Where φ = the Strength Reduction factor for Bolts in Shear – Table 3.3 of the
Student Code (φ = 0.8)
Vf = the nominal shear capacity of the bolt
fuf = minimum tensile strength of the bolt refer Table 9.3.1 for PC 4.6
and PC 8.8 Bolts
fr = a reduction factor given in Table 9.3.2.1 to account for long lap
connections >300 mm where the load transfer into each bolt may be different.
The outer bolts tend to carry more load than the inner bolts and unbuttoning
________________________________________________________________________
________________________________________________________________________
can occur – so for long rows of bolts > 300 mm such as in bracing, lap and
gusset joints a reduction factor is applied
nn = no. of bolts with threads intercepting the shear plane
Ac = core area of the bolt
nx = no. of shear planes without threads intercepting the shear plane
Ao = Shank area of the bolt
Tests have shown that the average shear stress at failure (fvf) is 62% of the tensile
strength of the bolt (fuf) and that any tension in the bolt due to tightening has no
effect on its shear capacity. That is /S and /T modes of tightening on the same PC of
bolt has no effect on the shear strength
Explain Double/Single Shear – Explain Clause 9.3.2.1 with diagrams
Bolts in Bending
Normal bolted connections with zero distance between plies are not considered
acting in bending – however holding down bolts which anchor steel baseplates to
say concrete foundations may attract bending depending on the confidence the
designer has in the concrete and mortar/epoxy packing between the baseplate and
the foundation – refer to section 2.7 of this section’s notes under Baseplate and
Connection Design. A few specialist connections such as the Sliding Hinge Joint with
Asymmetric Friction Connections use bolts in bending as part of the connection
behaviour
Local Effects in Bolted Connections

Ply in Bearing – For thin webs or cleats the failure mechanism for a bolt in shear is
not the bolt but may be the crushing of the thin web or cleat against the bolt face.
This generally manifests itself when large high strength bolts are required and the
designer forgets to check the thin cleat or web of the beam for bearing
The bearing area from a bolt of dia. df on a drilled hole in a cleat thickness tp is df x tp
Tests show that bearing stresses can be achieved well in excess of the yield stress
of the material such that the Code Value for the bearing capacity of a ply is given by
V*b ≤ C1 x φ x Vb = C1 x φ x 3.2 x df x tp x fup (Code Eq 9.3.2.4(1))
Where C1 = 1.0 for connections where seismic action is not being considered
φ = 0.9 (Table 3.3(1)) as the bearing is in the steel
fup = the ultimate tensile strength of the ply material (eg 430 MPa for
grade 300 steels, typically)
This is only achievable of course if there is sufficient edge or pitch between bolts
In order to ensure a bolt failure occurs with ovalling of the bolt hole a good rule
of thumb for cleat / ply thickness is Cleat Thickness NOT > 50% of bolt diameter
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This is a requirement for any cleat in a beam to column connection.
That is the maximum cleat thickness for an M20 bolt is 10mm
Ply failing in bolt tearout - If the edge distance is short or the pitch too low, the
cleat, web or plate may fail in shearing – often called a tear-out failure
In this instance –
V* b ≤ φ x ae x tp x fup (Code Eq 9.3.2.4(2))
Where ae is termed the “edge distance” and is the distance from the edge of an
adjacent hole or the edge of the cleat/plate plus half the bolt diameter. That
definition caters for both nominal sized (circular) holes and slotted holes, but for
nominal sized holes it is typically taken as from the edge of the adjacent hole or the
edge of the cleat/plate to the centre of the hole. The difference has no effect on
behaviour for nominal sized bolt holes. Nominal sized holes are 2mm greater in
diameter than the bolt diameter for up to M24 and 3mm greater in diameter than the
bolt diameter for larger bolt diameters
Reconciling the two equations, the second equation comes into play only with short
edge distances or pitch – ie when ae < 3.2 df
Minimum edge distances are recommended in the Code – refer Table 9.6.3 – for all
conditions other than those pertaining to connections required to sustain seismic
actions – 1.25 to 1.75 times the bolt diameter.
Minimum and maximum Pitch recommendations are also made
This would indicate that if minimum pitch and edge distances are adopted that
bearing failure due to tear-out should always be checked
Bolts (or threaded round bar) in Tension –
The capacity of a bolt in tension is governed by the “stress area” which is larger
than the “core” area, fails over a finite length which strains to ultimate and does not
necessarily occur at the points of deepest thread
Design Tensile Bolt Force given by
N*tf ≤ φ Ntf = As x fuf
Where – φ = 0.8 for Bolts in Tension
Threaded rod or round bar is exactly the same and must NOT be designed for the
gross area in tension of the bar – use the tension “stress area” of the equivalent bolt
diameter to determine the capacity of the threaded rod.
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Bolts in Combined Shear & Tension –
Normal elliptical interaction diagram and formula given by
2 2
V *   N* 
  +   ≤ 1.0
 φV   φN 
V* and N* are the applied actions

φV and φN are the Bolt capacities under Shear and Tension
φ = 0.8 for Bolts in Shear and Tension
Recommended Bolt sizes –
M12 – Generally PC 4.6 – Light Timber to steel or timber to timber connections

M16 – PC 4.6 or M12 PC 8.8 – For purlin & girt applications and lightly loaded cleats
and splices. For purlins and girts with slotted holes where the bolts must be fully
tensioned to prevent slip, the PC 8.8 is particularly important and M12 is the typical
size used.
M20 PC 4.6 – For baseplates (HDG) and lightly loaded structural connections
M20 PC 8.8 – for most structural applications usually in snug tight mode, usually the
minimum requirement for most structural steel projects
M24 & M30 PC 8.8 – for larger HD Bolt assemblies
In /TB or /TF modes use M20 unless circumstances create an unavoidable selection
of M24/8.8/T since anything M24 and above require heavier than standard
equipment or equipment in top condition. However for a large job such as Eden
Park for instance an M24/8.8/TF or B was the standard item on site and fabricators
building larger buildings will have the equipment to tighten much larger sizes (up to
M42 PC 10.9 which have an installed bolt tension of over 1000 kN)
Availability of Steel in Plate or Flat Bar (NOT rolled sections)
Flat plate is manufactured and available in varying widths and lengths of sheet –
refer to supplementary material on CANVAS. The sheets are a given thickness and
elements of connections or curved plate for a tank, say, can be cut to size and
manufactured and welded to suit.
Steel is also available in flat bar – similar to plate it only comes in given thicknesses
and widths (Refer Data Sheet Chart of Flat Bar). Flat bar is used extensively for the
manufacture of steel cleats in connections where cutting small pieces of steel from
sheet plate is high labour content
Eccentrically Loaded Bolt Groups –
Calculation of the load in the most remote bolt in a group of bolts receiving an
eccentric moment – same example as in CIVIL313 – refer those notes and example
below
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Proof of the formula that the load on
any bolt in a bolt group is given by
M × rmax
Vmax =
∑r2
Where
Vmax is the load on the bolt

perpendicular to the radial line
between the bolt and the centroid of
the group
M is the applied eccentric
moment on the Bolt Group
rmax is the distance from the bolt in question to the centroid of the bolt
group
∑ r2 is the sum of all the r distances for all the bolts in the bolt group
M = the sum of all the forces in each bolt times the radial distance to
the centroid
M = ∑ Vn × rn
In addition there is a linear relationship between the bolt force and the distance from
the centroid, the closer the bolt is to the centroid the smaller the load for a given
moment
Vn Vn max
=
rn rn max
Where Vn & rn apply to the load in any bolt and Vnmax & rnmax represents the load
and radial distance to the most remote bolt (the load we want to find)
Substituting for Vn
∑V × rn × rn Vn max × ∑ rn2
M = =
n max
rn max rn max
M × rn max
Then Vn max =
∑ rn2
Now ∑r n
2
(
= ∑ x2 + y2 )
Where rn is the polar distance from the centroid to any bolt
And x and y are the x and y distances from the centroid to any bolt
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Everything on the right hand side of the equation is known. Therefore for any bolt
group, given the applied moment, and knowing the centroid of the bolt group
(assuming ALL bolts are the same size), calculating the sum of all the x2 and y2 for
each bolt in the bolt group, the load in the most remote bolt in a line perpendicular to
the radial line from the centroid can be found
10 kN horizontal
70 70 310
120 kN vertical
75 75 75 100
170
75 125
180
70
42.4 kN
16.2.1 kN 49.1 kN
39.2 kN 16.2+12 kN
170
39.2+1kN
70
Cleats – [Note these Design Notes relate to simple building connections with
normal “commercial” level loads. It is emphasized that heavy Industrial or
Bridge Loadings can create substantial local and fatigue effects which must be
taken into account in the Design of ALL elements of a connection – reference
is made to HERA Limit State Design Guides Chapter 10, Hogan & Thomas
‘Design of Structural Connections’, and the behaviour of eccentric cleats in
compression in HERA Report R4-142]
Cleats are the pieces of steel welded to steel

beams or columns which when drilled for
bolts provide the connection between
members. The term “ply” refers to the
material being connected by a bolt whether it
is the cleat or the web or flange of the beam
– the plies must be considered on their merits
– it is nonsense to design a cleat for 12 mm
thickness required because of bearing on the
bolt when the web of the beam is only say 6.1
mm and may fail
Cleats should be designed using standard

thicknesses of flat bar or plate – flat bar is
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cheaper to fabricate as it only needs one cut to produce the item whereas if plate is
selected then at least two cuts are necessary. If specifying flat bar select available
widths from the supply charts (refer CANVAS charts) for example 130 by 16 Flat bar
is available but not 110 by 16 – so if the cleat in the top left hand detail needs about
70 – 80 wide and 8 thickness you can choose 75 by 8 or 90 by 8 flat bar as the cleat
size.
Moreover it is important to recognize that structural steel while in transit and in

construction/erection requires considerable robustness. It may be that purlin cleats
for a roof rafter can be designed as 3 mm cleats but if they all bend and become
twisted during lifting, in transit or during erection having to straighten them is a huge
problem to the contractor in cost of remedial work and time on the project
Therefore, be realistic with selection of the thickness of cleats – the majority of the
cost of a cleat is in the high labour content of cutting and welding – the thickness is
generally a very small part of the cost – in flange or web plates match (or select
greater than) the thickness of the flange or web – use minimum 5mm or 6 mm
thickness for small cleats and 10 or 12 mm minimum for reasonably important
structural cleats
Design cleats to transfer the design actions where applicable – whether in tension,
shear, bending and whether or not seismic forces are applied
Pure Tension – Cleat or Ply should be checked for bearing and bolt “tear-out” as
previous notes. Take the net area (cleat area less area of bolt holes) for stress
calculations – Refer Cl 7.2 Student Standard
Pure Shear – Cleat or ply should be checked for bearing and bolt “tear-out” as
previous notes. Take an effective area for shear calculations through the bolts as
shown –
Calculate Shear Capacity via Student Standard 5.11.4 for the applicable ply
thickness
The welds on cleats welded to columns are sized to develop the design tension
capacity of the cleat to ensure that under inelastic rotation from earthquake or fire
the welds won’t fracture and will enable the cleat to plastically deform in-plane .
Where failure can occur along two surfaces, one in shear and one in tension,
as shown in the figure on the next page, a BLOCK SHEAR failure may govern.
This requires checking for the combined ultimate strength shear failure along
the average area of the shear plane and the ultimate strength tensile failure
along the net area of the tensile plane. See the slide on page 46 of section 1
for the details.
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Examples where block shear failure will need to be checked.
Combined Shear and Tension – Check can be made via
2 2
 N * V *
  +   ≤ 1.0
 φN t   φVv 
Pure Bending – Cleat required to carry a bending moment. Calculate the section
moment capacity of the steel cleat bending about its strong axis by one of two ways.
Either (conservatively) use the elastic section modulus in bending ignoring bolt
𝑏𝑏𝑑𝑑2
holes) using 𝜑𝜑𝑀𝑀𝑠𝑠 = 𝜑𝜑𝑓𝑓𝑦𝑦 6 where d is the full depth of the cleat and provided the
edge distance to the extreme bolt holes is > 1.5df. Or use the plastic section moment
capacity, bd2/4, but including the reduced Z due to bolt holes. Normally the
reasonable thickness of cleat or ply to length of compression, distances at the
compression face and lateral support from connected components prevents the plate
from buckling. However students should understand that if “high” moments are
present due to unusually heavy design actions on a particular component then a
more rigorous analysis is necessary and the plastic section modulus reduced by bolt
holes must be used.
Combined Shear and Bending – Section 5.12 of the Student Standard applies –
that is combined effects need only be calculated if either
• the design moments on the section exceeds 75% of the allowable
moment capacity or
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• the design shear force on the section exceeds 60% of the allowable
shear capacity of the section
It is unusual for either of these to be applicable in which case combined action need
not apply – If either of these occur then the interaction formula in Cl 5.12 applies
namely –
V * 1.6 M *
+ ≤ 2.2
φVv φM s
Connections subject to Seismic Loading – this is not covered in Year 4 lectures

and all assignments, problem sheets and the examination assume that Section 12 of
the Steel Standard 3404 is ignored. However in the design office students must take
full allowance for the design of connections when seismic action apply – Refer for
example to alternative Minimum Design Actions Cl 12.9.2 and specific connection
clauses such as 12.9.5 and 12.9.6
Welding to Cleats – Generally cleats should not be intermittently welded to

structural steel although a design of the connection would normally show that very
little weld is necessary to transfer the design actions. Crevice corrosion can create
early deterioration in structural steel and for this reason alone it is usually better to
weld all round (full profile weld) with a light seam weld or a small fillet weld. Refer to
the welding section of the notes and the Student Steel Standard or NZS 3404. This
is especially the case with modern welding equipment where the cost of depositing a
weld is low and the speed at which this can be done when not stopping and starting
is high.
Proprietary Fasteners – Chemsets and Wedge Anchors are terms applied to

proprietary items available through suppliers for holding down bolt systems. These
are appropriate where existing concrete foundations, beams or walls require drilling
to insert new HD bolt assemblies or shear connections. Chemsets provide a
chemical connection where the drilled hole is filled with a threaded rod and epoxied
into place. Wedge anchors rely on the mechanical action of wedging to provide the
anchoring. Suppliers such as Hilti, Steelmasters, and Hylton Parker among many
have vast ranges of fasteners for all applications and their websites are worth
exploring. With proprietary items the product is only as good as the quality of
installation and careful supervision may be necessary in order for you to be satisfied
you are getting what you have specified”
Note that NZS 3101 Section 17 has comprehensive provisions for determining the
design capacity of fasteners in tension and shear failure, including into uncracked
and cracked concrete.
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2.2 WELDED CONNECTIONS
Section 9.7 of the Steel Standard and the Student Steel Standard
Successful welding requires adequate skills, well maintained equipment and a

proper care of welding rods. It also requires specialized equipment, job quality
control and trained personnel
It is therefore better to carry out as much welding as is practicable in the
fabrication shop where conditions and supervision are excellent rather than on site
where weather conditions, dust and confined access can make things difficult for
welding operations
The aim of good design is to provide the necessary structural performance for
the lowest completed cost – to achieve this attention must be given to
• Economical design & detailing
• Good welding procedures and correct process selection
• Responsible inspection techniques
The major sections emphasized are : -
 Minimize weld volume – don’t use a larger length or size of weld than
necessary
 Detail downhand welding where possible (overhead or vertical welding
is far less productive)
 Clean and simple detailing to enable welds to be placed simply and
therefore efficiently
 Detail maximum welding in the shop
 Use fillet welds as much as possible
 Where the full strength of a connected member is expected to be
transferred it is preferable to use vee butt welds
NOTE – That the quality of cutting plates is important, as on it depends the fit-up of
the parts and the root gap. If this is too large or variable, poor welds will result
necessitating back-gouging or rewelding from the reverse side (if there is access!) .
It is good practice to ensure that, if back gouging is necessary, access is adequate
for it to be done – if not give consideration to a backing strip
Properties of Weld Metal
The Code allows for different electrodes (Table 9.7.3.10(1)). E48XX is now the
preferred Grade for all steelwork and the lesser strength grade E410 is not
commonly used. The E48 refers to the nominal tensile strength of the weld metal.
fuw (nominal tensile strength) = 480 MPa

Fabricators will have a range of higher strength grades and Steel Construction New
Zealand has design guidance matching these to specific steel grades.
BUTT WELDS
Two forms are permitted
a) Complete penetration – develops the full strength of the metal being joined –
no need to calculate unit stresses, nor attempt to determine its size
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b) Incomplete penetration – used where less than full strength is acceptable –

these welds are permitted to carry shear and compression loads only and
have low ratings for fatigue duties. Design as for fillet weld with throat
thickness shown
Area Weld = lw x tw
Ultimate Shear Stress = 123 MPa
Ultimate Compression Stress = 228 MPa
Design Throat
Thickness t
The design throat thickness (DTT) for incomplete penetration butt welds are
determined in the same way as for fillet welds and the welding standard, AS/NZS
1554.1., gives DTTs for a common range of prequalified incomplete penetration butt
welds.
However, a balanced, double sided incomplete penetration butt weld can perform
very well in static and earthquake loading.
FILLET WELDS
Most fillet welds when joining steel components either in parallel or at 90° to each
other are specified as equal leg length fillet welds.
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Design of Fillet Welds -
Leg Length
Failure is through the Throat at an assumed 45 deg angle

for an equal leg length weld. That is conservative for welds
loaded parallel to one leg.
Throat Thickness tt = Leg length / √2
The Steel Code limits failure in shear across the plane to 60% of nominal tensile
stress (NTS) of the weld metal
Then Vw = 0.6 x fuw x tt x kr
where fuw = 480 MPa, typically and kr is a reduction factor for weld length
Note that kr only comes into effect after a weld length of 1700 mm which is highly
unusual for a structural weld – in which case in most instances we can ignore kr
φ factors in welding vary depending on the category of use
GP – General Purpose – essentially for non structural purposes such as seal welds
to keep out moisture – non structural – φ = 0.6
SP – Structural purpose – for welds under seismic or fatigue loading forming the
main structural load carrying path – φ = 0.8
SP has smaller permitted imperfections, which is therefore more reliable.
However this is achieved by greater supervision which is more costly but

appropriate for seismic or a wide range of fatigue loading. Very high levels of fatigue
loading require more stringent quality control and this is covered in a separate
welding standard. This is a more severe condition on welded steelwork than severe
earthquake loading.
Then V* ≤ φ Vw
Refer to information on CANVAS for design capacities of various leg length fillet
welds
Minimum and maximum sizes of fillet welds – Refer Code 9.7.3 and Table 9.7.3.2 for
minimum sizes and sketches for maximum sizes when welding lap plates
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Preferred sizes for structural purposes - 5mm and 6mm in the Fabrication Shop –
use 8 mm only if required structurally. On SITE use a minimum of 6 mm as the
confidence you can expect on site of getting a 5 mm fillet weld is low if specified as
such
Weld symbols and designation –
• Butt welds do not attract a size as the size and shape are governed by the
industry standard for varying thicknesses of flat bar or plate
• Vee, double vee or bevel butt welds have separate designations
• Additional symbols are required for a backing strip to a vee butt weld, or a
sealing run on the reverse side of the vee butt weld
• Fillet welds (always shown like a tick, vertical line then diagonal line sloping
away to the top right) require a size to be specified
• Supplementary symbols indicate weld all round, shop versus site welds (flag
always extends to the right) and lengths of intermittent fillet welds
See below for the standard weld symbol details.
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Simple Connections are straight forward
Multiply length of weld by φ Vw in kN/mm and this gives the shear force or tension
force able to be carried
For complex weld groups special analysis may be necessary or simplify by splitting
the design actions into how they are carried by the member being connected and
designing the weld group as an elastic assemblage under the design actions (Clause
9.8.1.1) or as an extension of the connected member (Clause 9.8.1.2).
For example – weld UB to endplate – say fillet weld all round to same size or
determine the flange forces from bending and axial load, sizing the flange to
endplate welds for these and determine the web forces from vertical shear and axial
load, sizing the web to endplate welds for the web forces
Total Length carries Total Length carries

vertical Shear Force Flange Forces
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In general, use fillet welds for attachments where strength of the connection is not
critical or where the strength of the provided weld is well in excess of the
requirement – examples such as splice plates, cleats for web side plates,
connections to pinned baseplates should all be in fillet welds
If the connection is a moment connection requiring often the full strength of the
joining unit to be developed then it is better to specify bevel or vee butt welds
In general, vee butt welds to UC and UB have a rear face to gain access to for
inspecting and cleaning the back face of the weld and ensuring a good weld is
achieved. Often called back gouging, the operator welds from one side, then grinds
out the back of the weld to inspect and tidy up the back face and runs further weld or
often a sealing run along that back face.
For the instance where an RHS is to be joined to a beam or a flat where a moment
connection is required, or where two RHS members are joined in a frame where it is
required to act as a knee joint then vee butt welding should be specified. However
because of the nature of the RHS , no back gouging can be achieved – so in this
instance we specify a vee butt weld with a backing strip which prevent flare and
oxidation occurring on the back face creating defects in the weld – refer sketches
Welding Rod with

clear access to vee
UB flanges ground
to achieve correct
profile for vee butt
weld Square
Hollow
Section
Rectangular
Hollow
Section
Applied Moment
RHS in Cantilever
M*
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The advantages of fillet welding are
• Economically attractive
• Minimum edge preparation
• Easy fit up
Note: there is a significant cost penalty for specifying higher weld size than
necessary
eg. increasing weld size from 6 to an 8 mm weld gives a 33% increase in strength
but a 78% increase in weld area, however both can be made with 1 weld run.
Increasing from 8mm to 10mm generally requires 3 weld runs which incurs a
considerable cost increase.
Intermittent Fillet Welds
Used when forces carried by the welds are small and ONLY in a mild non-corrosive
environment
Intermittent welds are NOT permitted in either of the following;
a. Where exposure to weather could cause an increase in corrosion

b. When member is subject to vibrations and fatigue is a consideration
If the designer selects intermittent fillet welding then Rule 9.7.3.8 and 3.9 of NZS
3404 applies (not given in the Student Code)
i. Clear spacing of consecutive welds must not exceed 16t (where t is

the thickness of the thinner member being connected) for compression
elements and up to 24t for tension elements. Maximum 300 mm in
both cases
ii. At the ends of a built up member or a fully stressed tensile member,
the length of the first weld must not be less than the width of the
component
Intermittent welds are slow and costly and in many cases it may be simpler and
cheaper to specify a light seam weld full length. They are also sources of crevice
corrosion and fatigue crack initiation. With increased use of machine and automated
welding intermittent welds are becoming less used.
Distortion –
Differential heating of the parent metal and subsequent shrinkage may produce
distortion. This can be avoided by using the minimum number of weld connections,
the minimum size of weld and balancing where possible welds each side of a
member. The rest is up to the fabricator who may be able to control distortion by
presetting, suitable jigging, use of correct welding sequences or straightening after
welding
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DETAILING -
3
??
??
6
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SUMMARY – WELDING
• Design with welding details in mind

• Do not specify oversize welds
• The weld all round policy should be avoided if possible as it can lead to
unnecessary cost. However much “all round” welding achieves a waterproof
connection minimizing onset of corrosion in exposed steelwork
• Keep the number of pieces to be welded to a minimum – use simple detailing
• Aim for as much shop fabrication in the shop as possible
• Use fillet welding in preference to vee butt welding
• Where practicable provide adequate access to welding
• Be receptive to alternative methods or proposals – recognize the value of
consultation with the fabricator
• Standardise joint details to reduce variety
• Regularly observe welding in the shop or on site to ensure the specification is
being met – it is a bit late when 100 tonne of steel arrives on site, is half
erected and you find an error in the welding – I realize it’s the contractor’s
problem but the error holds up progress, the remedial work is never as good
as the original and the contractor may try to recover his/her costs from other
areas
• Use nondestructive testing judiciously. NDT of welds is disruptive to flow of
work and adds considerably to the cost. Much of the cost will be avoided if
NDT is restricted to critical joints and carried out on a random basis
• Avoid turning of members to weld on the other side
HERA and other publications give properties of weld groups of varying shapes
2.3 MINIMUM DESIGN ACTIONS ON CONNECTIONS
• NZS 3404 Code Clauses 9.1.4, 12.9.2 and the Student Code requires Minimum
Design Actions (MDA) to be applied to all connections in steel. This ensures that all
connections have some minimum capacity, that they operate in a ductile manner
and/or are overstrength to the connecting members in order for predicted mechanisms
to occur with confidence
• Cl 9.1.4 refers to those connections subject to non-seismic actions only. These are
connections which have little relevance to the main horizontal load carrying
mechanism. Generally these are simple connections transferring load (be it gravity,
seismic or wind) from one part to another. For example a gravity beam between main
frames, a beam and post system relying on other parts of the structure for lateral load
carrying capacity, or the struts perpendicular to a main frame transferring longitudinal
wind or seismic forces to the main braced structure
• For those connections within parts of the structure which are the main seismic
resisting system in that plane, these are termed “connections subject to earthquake
load” and in this instance Clause 12.9.2 refers. This requires more stringent rules
depending on the category of the structure
• In this instance designers need to determine whether the connection is associated or
not with the main seismic resisting system even though seismic may not be the
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governing design action. If the frame is subject to seismic loads and forms part of the
seismic or lateral load resisting system then the connections are deemed to require
MDA under 12.9.2
• Under 9.1.4 the MDA for a connection is expressed as a percentage of the member
design capacity. Note minimum Member Design – ie if a selection of a member is
made for reasons other than design capacity (eg deflection) then the MDA only needs
to be a percentage of the minimum member design capacity and not the capacity of
the member chosen
• In addition under 9.1.4, the MDA only applies to the principle action of the
connection – eg a splice in a horizontal beam does not need to be designed to carry
the MDA for its axial carrying capacity since the principle reason for the beam is to
carry design moments and shear forces. The same applies to a pinned baseplate, the
connection need not have an MDA considered as a percentage of the moment carrying
capacity of the member since the column base is not required to take moment. A
column which may only come into tension occasionally from wind uplift would not
require the end connections to be designed for an MDA in tension
• Under 12.9.2 (copy enclosed) the MDA is expressed as a percentage of the design
section capacity of the member in place
Important Note : - As this course does not deal substantially with seismic design of frames,
the assignment asks the student to ignore sections of the code and is a simplified version of
what would be normally considered in the design office.
Civil 714 covers seismic design of steel systems and connections
MDA under Cl 9.1.4 – Connections not subject to earthquake loads - Design connection for
the greater of the design action of the member or the following percentages of the member
design capacity (minimum size of member under the ULS)
Principal Design Action BM SF Ends of Axial members

Tension or Compression
Connections in Rigid
Construction 50% 15% 30%
Note for rod bracing in tension – 100% of the tensile carrying capacity of the member
For Splices (Cl 9.1.4) 30% 15% 30%
MDA under Cl 12.9.2 – Connections subject to earthquake loads – Design connections for
the greater of the design action of the member, or the elastic design action (under seismic) or
the following percentages of the design section capacity of the member in place
For Seismic 30% 15% 50%
Note Splices in columns have special requirements depending on category of building

structure
Note also combinations of above should be considered as acting simultaneously
EXAMPLES : -
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1. A simply supported gravity beam between two frames spans 4.2 m and carries an
ULS UDL of 14.6 kN/m. The beam is not subject to design actions under seismic.
The design has resulted in a 250UB26 steel beam being selected for the duty.
Determine the design actions on the end connections.
M* = 32 kNm (midspan) V* = 31 kN
MDA – Design end connections for the design actions or 15% of the shear capacity of
the member. Since the principle action of the beam is bending and shear and not axial
the MDA due to axial can be ignored. In addition as the end connections are
considered simply supported MDA due to moment is zero
15% SF Capacity = 0.9 x 0.15 x 0.6 x 300 x 248 x 5 = 30.2 kN < Design Action so
Design Action = 31 kN governs
2. A secondary horizontal beam in a structural system requires a splice at a point where

the design actions are
i. M* = 12.3 kNm
ii. V* = 10.9 kN
iii. N* = 15 kN
The beam which is not subjected to seismic design actions, has been designed as a
310UB40, is laterally supported at 2600 mm centres and a uniform BM can be
assumed as acting at the point
Calculate the final design actions on the connection …
As the splice is not subject to earthquake loads MDA are based on

30% Moment, 15% Shear and 30% axial. Since the beam is not principally an axial
member the MDA due to axial can be ignored
15% SF Capacity = 0.9 x 0.15 x 0.6 x 300 x 252 x 6.1 = 37.4 kN Governs as this is >
Design Action (10.9 kN)
30% BM Capacity given by 𝜑𝜑𝜑𝜑 = 𝜑𝜑. 𝛼𝛼𝑆𝑆 . 𝛼𝛼𝑀𝑀 . 𝑓𝑓𝑦𝑦 .ze
∝𝑆𝑆 = 0.713 For a 310UB40 over 2600 mm
∝𝑀𝑀 = 1.0 for uniform bending moment
Then BM capacity = 0.9 x 0.713 x 1.0 x 300 x 395E3 = 76 kNm
30% BM Capacity = 23 kNm > Design Action (12.3 kNm) so MDA governs
Design Splice for the combined actions of V* = 41.5 kN, N* = 15 kN and M* = 38

kNm
3. A standard Column baseplate to a portal frame requires designing. Seismic actions do

not govern the design of the frame nevertheless the frame is part of the horizontal
seismic resisting element of the structure. The portal frame is considered pinned
based and has no uplift actions under any design condition. The following design
actions apply –
i. M* = 0
ii. V* = 46 kN
iii. N*c = 284 kN
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iv. Elastic Response Seismic N*c = 196 kN
Determine the appropriate final design actions on the connection …
As the splice is subject to earthquake loads MDA are based on the design actions, the
elastic response seismic action, or 30% Section Moment capacity, 15% Shear capacity
and 50% section axial capacity of the final member designed (310UB40). Since the
column base is assumed pinned the upper limit elastic seismic action is zero, therefore
the MDA due to moment can be ignored. The column is not principally a tension
member so that MDA due to tension can be ignored
15% SF Capacity = 0.9 x 0.15 x 0.6 x 300 x 304 x 6.1 = 45 kN V* = 46 kN Governs

as the MDA is < Design Action
50% of the axial section capacity – 0.5 x 0.9 x 0.952 x 5210 x 300 = 670 kN
Design baseplate for the combined actions of V* = 46 kN and N*c = 670 kN
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2.4 WEB PLATE CONNECTION
The most common simple shear connection used is the web plate (WP) connection for beam
to beam or beam to column. However while it is described as simple the understanding of the
load path through the connection and the flexibility of the supporting member can raise
significant issues which students need to carefully understand.
Consider the LH sketch - A shear force can be transferred from beam to column (say) through
a single bolt as shown in the sketch. The shear force passes from the web of the beam
through into the single bolt into the cleat which is welded to the column face. The design
action must be passed through each element, that is, web, bolt, cleat, weld, column
P P
M' M'
e e
Moment carried by Moment carried by

support support
It is clear that a pin (zero moment) occurs at the centroid of the bolt in which case the cleat
and the weld must be designed for a small moment induced by the shear force multiplied by
the eccentricity between the bolt and the column face. The column itself is also subject to the
moment caused by the shear force this time multiplied by the eccentricity from the centerline
of the bolt to the centerline of the column. This is expressed in structural terms by the
moment diagram also shown. The cleat, the weld and the column can be assumed in
structural terms to be relatively rigid and should be designed to accommodate the moments
induced without severe rotation or displacement. If they aren’t, the connection would not
work and the supported beam would move unacceptably downwards at that point
Turn now to the connection between a beam and another beam as shown to the right of the
previous sketch. If the connection was designed for a single bolt as in the column support,
the supporting member, cleat and welds would all need to be capable of carrying the moment
caused by the eccentricity of the connection. In general unless the beam is restrained in some
way this can’t occur and the supporting beam rotates and produces unacceptable and
unwanted rotations and displacement of the supported beam. The solution is to make the
bolted connection a bolt group and design the bolt group to carry a moment due to an
eccentricity. This is expressed in structural terms by the bending moment diagram shown
above
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Because the supporting member on the left can rotate it is assumed a flexible connection
therefore design these by assuming the pin is at the centerline of the support beam and the
cleat and bolt group require designing for shear plus a moment caused by the eccentricity.
P
M'
Moment carried by
bolt group
The former assumes the bolt group connection or the supported member cannot rotate and
can carry the moment generated by the eccentricity of the connection
The latter assumes some rigid support and the connection may only be designed as a pinned
connection – note weld carries moment and in excess of cleat tension capacity
When designing any connection you must determine which item is supporting which and
whether or not flexibility can be attributed to either. In this instance eccentricities occur and
one or other of the connection items need to be designed to ensure the connection remains
intact
For example – Design a connection beam to beam (flexible support) between 310UB40
beams, intersecting at right angles, to carry an ultimate shear force of 114 kN (derived from
either the worst design action under ULS ;load combinations or the minimum design action)
140 mm
114 kN
CL
Design for V* = 114 kN and

M* = 114 x 0.14 = 16 kNm (Support must not twist therefore the bolt group
and cleat must allow to carry the eccentric moment induced)
Try 2 Bolt connection V* per bolt = 114/2 = 57 kN M* = 16 kNm
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Check Geometry and layout of Cleat – Generally the first bolt is ~100 mm down from the top
flange – Cleats cannot interfere with the root radius of the web/flange interface and there
must be a clearance from the edge of the cleat and the bolt (generally 1.5 df ). In addition (for
say a splice) if part of the webplate is welded to the web of the beam, room must be given for
the fillet weld to be constructed – (see sketches page 20)
For example - Clear Distance between root radii for 310UB40 = 284 – 23 = 261 – Try 210
length cleat with bolts edge distance 35mm giving 140 mm between two bolts. Distance
from top of flange to centerline of bolt = 152 – 70 = 82 OK
2 x H* x 0.07 = 16 kNm Therefore H* = 114 kN

R’ = √( 1142 + 572 ) = 127 kN/bolt
Try M20 8.8/S Bolt since there is no specific requirement for a “no slip” connection
øV = ø x 0.62 x fuf x Ac = 0.8 x 0.62 x 830 x 225
= 93 kN < 127 kN NG
Note: - Inserting a third bolt in between the two bolts makes no difference to the horizontal
force H* but does reduce the direct shear ( R = 120 kN still NG!)
Therefore, either, go to 2 M24 8.8/S bolts, increase depth between bolts, or a four bolt
connection – There is no room to increase the length of the cleat (cramped for room at the
flanges) and a four bolt connection would increase the eccentricity of the connection
Try 2 M24 8.8/S Bolts – Edge distance 1.5d = 35mm OK; Pitch = 140mm > 2.5d = 60 OK ;
R* → still 127 kN; 1 M24 8.8 Bolt carries 133kN > 127 kN OK
Design the Cleat – As recommended by HERA - check the moment capacity of the cleat
bd 2
using the elastic section modulus (ignoring bolt holes) = 73.5E3 mm3 for an 10 mm
6
cleat –then ϕ M = 0.9 × 300 × 73.5 E 3 =19.8kNm > M* = 16.0 kNm OK
M*
Since ≥ 0.75 require to check combined bending and shear
φM
Check Shear – Effective length of area for shear equivalent to
140 + 35 – 1.5 x 26 = 136mm Area = 136 x 10 = 1360 mm2
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Then ϕVv = 0.9 × 0.6 × 300 ×1360 = 220kN ≤ V * = 114kN
This exceeds 60% of allowable therefore check Combined actions
114 1.6 ×16

+ =
1.81 < 2.2 Acceptable
220 19.8
ADOPT - 210mm depth by 10mm thick mild steel cleat as shown, drill for 2 M24 8.8/S
Bolts at 140 centres
Need to check also the fillet weld to the UB
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2.5 MOMENT END PLATE CONNECTIONS
Beam End Connections for Beam-column Joints requiring transfer of horizontal axial loads
(if present), and end moment and shear forces
These notes give a general descriptive overview of the behaviour of moment end plate
connections. The detailed design procedure is given in HERA Report R4-142, the relevant
pages of which are in the following section of these notes.
The connection is required to transfer a moment from the beam to the column. Bolts about
the tension flange resist the axial tension force in the flange of the incoming beam induced by
M* and any N*t plus bolt prying forces, which are incorporated into the design procedure
directly (see extracts from HERA Report R4-142 in these notes). Any axial force is
transferred through the flanges in direct proportion to their areas or equally for a symmetrical
beam
Generally for negative support moments only (tension in the top beam flange), the top bolts
carrying this tension force. They need not be designed for tension AND shear as the bottom
bolts at the compression flange generally would carry the shear forces V* required to be
transferred by the connection. This should however be confirmed by calculation at all times.
In addition and where appropriate, design checks must be carried out to ensure the design
action for a positive bending moment (placing the bottom bolts into tension) is covered
The design procedure is given in HERA Report R4-142 and in the SCNZ Steel Connect
Part 1.
Geometry –
The endplate is typically equal to or greater than the thickness of the column flange being
bolted to, although in large building structures using three plate column sections, flange
thicknesses can be high
Refer to the MEP8 Drawing (Ex R4-100) – This general arrangement can be considered
typical for most applications whether or not there are 2 or 4 bolt groupings top and or bottom
and whether or not the bolts are included within the steel beam or straddle the flange as
shown. The following guidelines apply -
1. Generally 2 bolts above or below the tension flange or 4 bolts in tension grouped
symmetrically about the tension flange
2. Generally two or four bolts about the compression flange as for tension
3. Where two bolts they are generally outside the tension flange but not mandatory as
inside the tension flange provides better appearance and takes less room
4. While these are not shown more often than not horizontal stiffeners to the column are
common or recommended whether or not they are needed
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NOTE - The design processes given in year 4 are for static applied actions only and do not
address overstrength actions and seismic detailing
Bolts Design – Prying – Only occurs in steel to steel connections, when bolts are in tension
and plate deformation occurs. Prying occurs in mode 1 or mode 2 failure of the equivalent
Tee stub so occurs in most cases. It does not occur to a significant extent in moment end plate
connections onto concrete, because concrete is a softer material than steel. It is a complex
phenomenon and the earlier design requirements for these connections used different
procedures with different levels of prying force depending on the type of connection and the
geometry.
Prying is shown conceptually below. This shows a Tee stub endplate in mode 2 failure, where
there is separation at the bolt line. The applied tension force is 2F, with two bolts clamping
the Tee stub to the foundation to resist the applied force. However, the deformed shape of the
endplate generates a prying force Q between the endplate tips and the supporting surface,
meaning that the tension force in the bolts for equilibrium is increased from 2F to (2F+2Q)
where Q is the additional force due to prying on one bolt.
The prying force can be high, but is now incorporated directly into the mode 1 and mode 2
failure conditions and so the increase in bolt force due to prying does not need to be expressly
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considered. What instead happens is that the bolt design tension capacity is used to calculate
the applied tension that can be resisted and the prying action decreases this. See the
introduction to yieldline theory at the end of the steel refresher notes and the details in the
HERA Report R4-142 extracts following in section 2.2.
Final drawing and details of a typical connection look as shown below ……….
6 M24 8.8/TB
BOLTS
2 edges
(Typ) 5
60
70
10 mm Web plate
460UB75 680 x 225 x 25
to one side of web
Endplate
Plug Weld (HERA)
5 60
100 x 5 mm Web
stiffners t & b
each side
Flanges and face
of stiffner 140
310UC97
2.6 COLUMN BASEPLATE DESIGN (Pinned) –
Design connection for Ultimate design actions or the minimum design actions calculated.
Required to transmit axial compression loads onto (generally) concrete foundations (ie don’t
crush the concrete), to transfer axial tension forces (if any) through to the foundation via
tension bolts and shear forces via bolts in shear into the footing
See below for a typical baseplate and HD bolt detail – steel column is welded to a baseplate
which may or may not extend beyond the dimensions of the steel column
Usually 2 or 4 bolts - Usually 4.6/S HD bolts and often Hot dip galvanized - Often welded
onto spacer bars. For special reasons like moment connection not pinned) or extreme tension,
8.8/S bolts may be used but requires care if welded to spacer bars or hot dip galvanized.
Baseplates need to be robust (they should not be damaged during transport and erection) and
therefore depending on the size of the column minimum thicknesses can be 16 or 20 mm. Be
aware of the scale of all parts of the connection - it is inappropriate to have a 10 mm thick
baseplate with M24 holding down bolts
Don’t be too generous with size of plate – extend 15 – 30 mm beyond flanges of column is
common (UNLESS something interferes with this extension – see below for NO outstands)
Column usually fillet welded to baseplate all round – minimum shop fillet weld of 5 mm fillet
weld often well in excess to transfer load
Required to check compression stress on the concrete (concrete in bearing), thickness of the
plate and number of Bolts (to transfer shear forces)
For baseplates required to carry tension forces, yield line theory is required – this is covered
in section 2.4 of these notes.
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Example of final details for a baseplate design and fabricated to achieve no outstands
HD Bolts can be positioned

within the reinforced concrete
foundation prior to pouring in
the exact position (and height)
required by the steel erection
subcontractor. Often the bolts
(if in mild steel) can be welded
into a group with flat bar
joining the bolts in an exact
position which improves the
chance of getting the group in
the correct place. However this
practice may prove a problem
with interference with rebar –
relies on contractor preference
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Example of Baseplate Design for 530UB82 subject to ULS Design actions N* = 600 kN and
H* = 85 kN in a seismic frame. MDA refer p 44 treat as a splice in a seismic frame.
30% of axial load section capacity of 530UB82 = 30% of 2557 kN = 767 kN
15% of shear capacity of 530UB82 = 15% of 821 kN = 123 kN
OR – lateral load minimum design action of .05 * max ULS axial force = 30 kN
MDA governs design for N* = 767 kN and H* = 123 kN; d = 528 mm b = 209 mm
Select Baseplate Dimensions 600 by 225 – Then referring to Fig 10.40 for
Plate Thickness Design -
N = 600 mm, 0.95D = 502 mm, m = 49 mm
B = 225 mm, 0.8B = 167 mm, n = 29 mm
n’ = 0.25 √bd = 83 mm l = n’ = 83 mm
2 × 767,000
t bp = 83 × = 17.1mm ADOPT 20 mm baseplate ←
0.9 × 300 × 225 × 600
Check Compression on Concrete – this is conservative and assumes some flexibility in the
baseplate. If the baseplate is substantially stiff and does not flex the engineer may adopt the
bearing being applied across the entire baseplate (Engineer’s discretion and judgement). In
baseplate design the method shown here is conservative. Assume 300 dispersion through the
thickness of the baseplate to the concrete and in calculating the dispersion area under the web
don’t double up from the full area adopted under the flanges. Under Section 16 of NZS3101
Concrete Standard, when the supporting surface is wider on all sides than the loaded area the
design bearing strength can be raised by a factor of √(A2/A1) NOT > 2 where A1 is the loaded
area at the surface and A2 is the wider and larger area taking a 1:2 dispersion through the
concrete, provided the structure below the baseplate allows that
41 435 41
tbp
q = t + 2√3 x tbp
530UB82 – Flange Area = 209 x 2 x (13.2 + 2√3 x 20)

= 209 x 2 x 82 = 34500 mm2
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Ignores spread at ends of flange – usually limited by side of baseplate

Web Area = 435 x 1 x (9.6 + 2√3 x 20)
= 435 x 78
= 34300 mm2
Ignores spread under flange – ie length 435 = 517 (length of web) – 41 (half spread through
baseplate) x 2 – refer sketch previous page
Total Bearing Area A1= 68,800 mm2 In this instance it is expected that the foundation is
much wider than the baseplate and that a factor of two can be applied to the ULS bearing
capacity
Assume 20 MPa Concrete (28 day strength) – f’c = 20 MPa
From Cl 16.3.1 NZS 3101 (2006) - not included in the Student Code for Design of Concrete
Structures
Bearing Capacity of Concrete = ø x 0.85 x f’c x 2 x A
ø = 0.65 Then = 0.65 x 0.85 x 20 x 68,800 x 2
= 1520 kN > N* = 767 kN OK
Check Weld – Length of fillet weld (530UB82) ~ 2x209 + 2x502 + 4x100=1822 mm

Capacity of Weld – GP E41, say, 5 mm shop weld = 0.522x1822 = 951 kN > combination of
ULS axial load (767 kN) and ULS shear force (123 kN) OK. Note this assumes there is no
bearing of the steel column onto the baseplate and ALL the vertical and shear load passes
through the weld. This is conservative but usually OK without unduly penalizing the cost of
the connection. To ensure 100% contact between the steel column and the baseplate one
would have to machine the end of the column – a costly process
Design holding down bolts as an Anchor Bolt – see Notes following -
Preliminary Design - Transfer Shear in Four Bolts V* = 123 kN – Try 4 - M20 – 4.6/S Bolts
øV = 0.8 x 0.62 x 400 x 4 x 225

= 179 kN > 123 kN OK BUT …
see Design of Anchor Bolts for transfer of shear through into the concrete foundation
Construction Details – Holding Down Anchor Bolts to be poured in-situ with the foundation
block can be sent to site as single units with a fishtale end or in a unit fixed to a cage by bolts
or welds
For design of a moment-resisting column baseplate see the notes from HERA DCB No
56 given in section 2.4 of these notes.
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2.7 DESIGN OF ANCHOR BOLTS carrying Shear Forces for Steel Structures
Refer to SESOC Design Guide No 1 by John Scarry published in SESOC Journal Sept 2009
and included in CANVAS Resources
Because the design of anchor bolts (or holding down bolts) straddle the material standards
between steel and concrete, neither standard addresses the issues of bolts in tension and in
shear within concrete
This short section addresses the requirement to carry shear forces via long bolts, from a
pinned steel baseplate into the concrete foundation, a very common occurrence but one which
is not well covered in design offices
Many problems have arisen from this interface where incorrect design procedures have been
used, inadequate anchorage or concrete edge distances are specified, inaccurate placing of
holding down bolts necessitating modifications on site, all of which can provide a situation
where the capacity of the connection comes into question
The SESOC Design Guide written by John Scarry is comprehensive and provides substantial
information on the design, installation and site problems which require addressing in the
Design Office
Size of Holes in Steel Baseplates – Normal drilling for structural steel bolts (in connections
for instance) allow for a 2 mm increase in hole diameter over the nominal fastener diameter
(22mm dia hole for an M20 bolt). In baseplates, in recognition of the lower tolerances of
concrete construction a 6 mm increase in hole diameter is allowed provided a special plate
washer (minimum thickness 4 mm) drilled for 2 mm oversize hole is used under the nut. If
any baseplate has a larger diameter hole than 6 mm (say due to changes required on site to
accommodate site conditions) then a specifically designed washer shall be provided thick
enough to resist and transfer all actions through the bolt will be provided and site fillet
welded to the baseplate in its final position
Anchorage in the Concrete (bolt in tension) – Refer to the SESOC Standard – it is not the
intent of these notes or this course to refer to or provide guidance in designing the
embedment of a tension bolt, threaded rod or details into the concrete except to say that it is
good practice to ensure that all bolt embedment is within the reinforcement cage of the
foundation and has sufficient side cover
Detailing – It is important that designers make clear by sketches within their calculations of
the intent of the design. Holding down bolt assemblies are critical as the steel fabricator
provides the steel bolts often placed and held in groups of 2 or 4 and the concrete contractor
requires to set them out and place them within the foundation in their correct position,
orientation and level, suspended within the cage of reinforcement prior to pouring. The
dimensions, lengths, anchorage all need specifying clearly so that the drafts-people have it
clear on the drawings and the Contractor and subcontractors all understand the requirement
Grout, Mortar and Drypack – In the construction of concrete foundations, the top surface
of the Concrete can be rough, may not be level and almost certainly not at the precise
elevation required by steel erection. For this reason, the concrete foundations are finished to
a level some 20 – 50 mm below the level of the baseplate. Erection of the structural steel
requires precision and the base of the columns are set up on steel shims. The varying
thickness gap between the underside of the baseplate and the concrete surface is then filled
with non shrinking grout, mortar or drypack which provides the surface through which
vertical and horizontal loads may pass and achieves a correct height within steelwork
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tolerances for the structure. The design engineer is required to make a decision as to whether
or not the material between the baseplate and the concrete can transfer the shear loads. This
decision sets the basis for the design of the anchor and comes from experience, a knowledge
of the material and the confidence of achieving the quality needed on site to assume that
Minimum Design Actions for Baseplates – Baseplates are not specifically mentioned in
MDA requirements. In NZS 1170 under “Structural Robustness” a requirement for minimum
lateral resistance of 5% of the axial load of any connection of compression tension members
should be allowed. A baseplate should be regarded as a splice (be it between different
materials) in which case normal splice requirements for MDA. In this course we consider the
MDA for a baseplate to be the greater of the max design ULS actions on the connection, or
the MDA splice requirements for a column (9.1.4.1 (v) ad 12.9.2.2) or in the case of lateral
forces a lateral connection force 5% of the axial load
Bolt Strength – Care must be taken when specifying Anchor bolts especially in detailing.
The SESOC standard provides excellent guidance on selection. To summarize: PC 4.6 bolts
are most common with no restrictions on bending the bolts and welding attachments (for
spacers). PC 8.8 (fy < 650 MPa) may be used but has restrictions on bending and welding.
Both these grades can be hot dip galvanized. Grades of Anchor Bolt material with Fy > 650
MPa should not be hot dip galvanised
Design of Anchor Bolts in a Pinned Base for Shear only – Shear transfer by Dowel
Action through the Bolts – Horizontal shear forces can be transferred from the baseplate
into the foundation by a physical shear key penetrating the foundation or by friction between
the baseplate and the grout/foundation assuming that a normal vertical compressive force is
present to achieve the friction. The former is an expensive solution and the latter can be risky
depending on the confidence the designer has of the vertical compression force being present.
Usually designers allow for the holding down bolts to secure the column base and transfer the
shear forces assuming zero friction at the base. The bolt is required to transfer a horizontal
force from the column baseplate (by bearing on the hole) through to the embedment within
the grout or concrete foundation (refer to the attached sketches). Depending on the resistance
of the grout to the bolt, the bolt experiences a level of bending dependent on the length of the
lever arm between the horizontal reactions.
Refer Sketch on the next page -
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IMPORTANT NOTE – When carrying out calculations for anchor bolts always include
sketches like these within your calculations to remind you and the engineer checking your
calculations what model you have used in determining the bending in the anchor bolt
Note: - If the steel baseplate has oversize holes then there is no bearing between the baseplate
and the bolt and the dimension p moves up into the washer thereby increasing the lever arm
p a
LeverArm = +g+
2 2
– When the mortar can resist shear, g = 0

– f’ is the minimum value of strength of the mortar or the concrete whichever
is appropriate
– df is the nominal diameter of the fastener
Vi * = 2 f u d f p = 2.5 f ' d f a
Solving for the unknowns p and a the Total Moment demand on the bolt is given by
a p
M i* = Vi *  + g + 
2 2
The bolt is assumed to be held rigidly within the concrete and its assumed that a full plastic
moment can occur within the bolt at that point. At the nut/washer interface, bearing
vertically occurs between the washer and the baseplate thereby achieving some fixity and its
assumed that the full elastic moment of the threaded section at the top is achieved
Accordingly it is accepted that 40% of the total moment demand can be attributed to the top
and 60% moment carried by the bolt at the bottom with the bending moment in the bolt
looking like …..
The moment capacity of a threaded bolt is given by…
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φ .M f = φ . f y .ze
where
1.5π 3
ze = .d s and d s = d f − 0.9383 p
32
where p is the pitch of the threads – Refer to Bolt Data (Technical Data Sheets)
eg an M20 4.6/S HD Bolt has a moment capacity given by
p = 2.5mm; Then; d s = d f − 0.9383 p = 17.65mm
z e = 810mm 3 − Then − φM f = 0.8 x 240 x810 = 0.16kNm
which isn’t much ………… Accordingly when designed properly for this condition the bolt
size often doesn’t meet the requirements and more bolts or bigger bolts are required
The interaction equation for combined bending and shear is given by …
V* M*
+ ≤ 1.2
φVn φM s
Example - Analyse the previous four bolt baseplate connection with 4 M20 4.6/S HDG HD
Bolts as shown required to transfer the maximum ULS shear Force of 123 kN. Assume in the
first instance that the grout is incapable of carrying shear forces and then assume full strength
is available through the grout (assume f’ of grout and concrete is 25 MPa)
Case 1 – Assume no strength in grout V* per bolt = 123/4 = 30.75 kN

30750 30750
Depth of stress block in baseplate is given by p = = = 1.92mm
2. f u .d f 2.400.20
Depth of the stress block in the concrete is given by
30750 30750
a= '
= = 24.6mm
2.5 f c .d f 2.5 x 25 x 20
Moment Arm (incl 25 grout) = 0.9 + 25 + 12.3 = 38.2 mm
V* = 30.75 kn/bolt and therefore worst M* = 60% of 30.75 *0 .0382 = 0.70 kNm
Design Shear strength given by φV = φ 0.62 f u Ac = 0.8 x0.62 x 400 x 225 = 44.6kN
Design Moment Strength of the Bolt (see above) = 0.16 kNm >> 0.70 kNm !!
M* is already well in excess of the strength in moment so we need more bolts, higher
strength bolts or need to provide some shear capacity within the grout
Case 2 – Assume grout is capable of carrying full shear, and try high strength bolts
As above Moment Arm now = 0.9 + 12.3 = 13.2 mm – M* now = 0.24 kNm > 0.16 kNm
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Try M20 8.8/S HD HDG Bolts – Moment Capacity given by -

p = 2.5mm; Then; d s = d f − 0.9383 p = 17.65mm
z e = 810mm 3 − Then − φM f = 0.8 x640 x810 = 0.41kNm
Design Shear strength given by φV = φ 0.62 f u Ac = 0.8 x0.62 x830 x 225 = 92.6kN
30.75 0.24
Moment Shear Interaction given by + = 0.91 < 1.2 OK
92.6 0.41
Adopt 4 - M24 8.8/S HDG Anchor Bolts for this baseplate
600 x 225 x 20Fl 530UB82

Drill for 4M20 8.8/S
HDG HD Bolts
90
TYP 250
5
Baseplate on 25 mm high strength

grout see specification
2.8 SPLICES –
Required for site connections –
• Ensure maximum welding of parts of the structure in the fabrication shop and
transport units to site sufficiently large to fit onto the tray of a truck or easily
transported without major traffic problems – generally lengths should not exceed 25
m but longer lengths can be accommodated at night and with pilot vehicles.
Generally keep width to 2.4 metres but standing upright on the back of a low loader
can get wider units
• Ensure site connections are relatively easy to build and access is straight forward
• Generally site connections should be bolted to minimize the cost but if the site bolting
cannot achieve the appearance or structural requirement of the connection then site
welding may be necessary
• Drawings should detail site connections and show their expected position by
dimensioning in each frame – otherwise the fabricator may choose for themselves
where to put the site connection and this may not be where the designer wants it
• For splices in beams and columns in non-residential construction minimum weld size
is 5mm shop fillet weld and minimum cleat size is 10 – 12 mm thickness. Minimum
bolt size is 16 – 20mm diameter
Design for the greater of
a) Design Action or
b) Minimum Design Requirements
________________________________________________________________________
________________________________________________________________________
Most common is the welded/bolted splice in beams and columns
• Place usually at points of low demand eg points of contraflexure

• Place to allow ease of fabrication, transportation and erection
• Web plates carry and transfer the shear forces
• Top and bottom plates carry the moment transfer by compression and tension forces
through the splice plates attached to the flanges
• Use 8.8/TB Bolts for the flanges – fully tightened in accordance with Friction Grip
standards. Normal 8.8/S snug tight bolts perfectly OK for the web plate connection
Example : -
Design Rafter Splice to a 460UB67 to transfer the following Design Actions –
M* = 70 kNm at the Splice Position V* = 60 kN N* = 15 kN compression
Beam is fully restrained throughout its span

Flange Plates carry BM in tension and compression plus proportion of axial load
Web Plate carries Shear and proportion of Axial Force
Design bolts in flange -
Moment – Design for either M* = 70 kNm or Minimum Design Action -
MDA = 0.3 x ø Ms = 0.3 x 0.9 x 300 (Fy = 300 MPa) x 1480E3 (Sx)
= 119 kNm Governs ←
Axial Force - Shared flange to web in ratio of 2Af to Aw
= 2x190x12.7:428x8.5 = 0.57:0.43 (Flange to web)
Flange Force = 119 / (454 – 12.7) = 270 kN or 70 / (454 – 12.7) + 0.57x15/2 = 218 kN
N*f = 270kN governs - Try M20 Bolts – Ac = 225 mm2
1 M20 8.8/TB Bolts in single shear carries – 0.8 x 0.62 x 830 x 225 = 93 kN
1 M16 8.8/TB Bolts in single shear carries – 0.8 x 0.62 x 830 x 144 = 59 kN
________________________________________________________________________
________________________________________________________________________
4 – M20 8.8/TB Bolts carries 372 kN > 270 OK
ADOPT 4 M20 8.8/TB Bolts to each Splice Plate ←
Select Size of Flange Plate –
Need –
1. Room between bolts across the beam and making due allowance for room for a nut,
and a spanner against the web – see sketch – dimensions are 70 mm/ 90 mm/ or 140
mm depending on the width of the available flange and edge distances
2. Edge (side) distances should be ~ 1.5D = 30 mm for a 20 dia bolt
3. Room for a weld between the side of the splice plate and the flange of the beam, that
is select a width of flat bar wider or narrower than the flange width to at least allow
for a weld – say minimum 10 mm
4. End distances check for tearout
5. Thickness may be governed by bearing
Re Item 3 above ; Eg a 250 UB has a 146 mm wide flange – don’t use a 150 mm splice plate
as it doesn’t comply with item 3 above – use 130 wide flat bar. Similarly for a 200UB with a
134 Flange – use a slightly greater width of splice plate say 150 mm to allow room for the
weld. Designer can also think of downhand welding during fabrication which means a wider
flangeplate on the bottom and a narrower flangeplate on the top
In this instance – 460 UB has a 190 by 12.7 thick flange – try 150 wide splice plate –
Drill for 4 M20 Bolts – check yield or failure criteria under tension force
Area steel less bolt holes = (150 – 2 x 22) x t = 106 t
Yield - Thickness of Plate needed = 270,000 / (0.9 x 300 x 150) = 6.7

mm
Failure – Thickness needed = 270,000/(0.9 x 0.85 x 1.0 x 430 x 106) =

7.7 mm
As a check, compare this with 460 UB Flange thickness = 12.7 mm – as long as this
thickness is of the same order as the flange, no errors have been b made? In this instance say
10mm cf 12.7 mm all OK
Check bearing on bolts – Max ULS V* in one bolt = 270/4 = 68kN/bolt
9.3.2.4.1 - ϕVb = ϕ3.2dftpfup = 0.9x3.2x20x10x430 = 248 kN > 68 kN OK in bearing
Check tearout for a 45 mm end distance on the plate or 35mm on the flange – V* = 68
kN
9.3.2.4.1 - ϕVb = ϕaetpfup = 0.9x45x10x430 = 174 kN > 68 kN OK in tearout (also OK on

35mm and 12.7 flange
________________________________________________________________________
________________________________________________________________________
ADOPT 10 mm x 150 Flat bar top & bottom Splice Plates ←
45 90 35 150 x 10 Flat
150 3 sides
6
NOTE – this is a preliminary sketch & incomplete – for a FINAL sketch see page 56
Weld – Length of weld available = 150 mm + 300 mm = 450 mm – required to transfer 270
kN
φVw of 5 mm fillet weld – 0.522 kN/mm - (E41XX GP Weld – see chart)
Then φVw (Weld group) = 0.522 x 450 mm = 235 kN < 270 kN – NG
NOTE: Welding plate to plate the fillet weld must be 2mm less than the plate thickness - see
p20
Try 6 mm weld » 0.626 kN/mm – or lengthen weld and plate to 200 mm and try a 5
mm weld – 6mm fillet weld OK against 10mm thick flatbar
φVw (6 mm Weld group) = 0.626 x 450 mm = 282 kN > 240 kN OK
OR φVw (5 mm Weld group) = 0.522 x 550 mm = 287 kN > 270 kN - OK
ADOPT as shown above with a 6 mm fillet weld ←

Web Plate & Bolts – PRELIMINARY SKETCH SUPERCEDED – see final sketch p 56
________________________________________________________________________
________________________________________________________________________
NOTE – this is a preliminary sketch & incomplete – for a FINAL sketch see page 46
Maximum size of plate (between fillets) = 400 mm – BUT NOTE ability to achieve a 5 fillet
weld between the edge of the webplate and the flange say 30mm top & bottom
Try 330 mm length of web plate and, say, 200 wide flat bar (cheaper than plate)
Required to transfer V* = 60 kN ↓ and N* = 0.43 * 15 = 7 kN →
However Minimum D A 0.15Vv = 0.15 x 0.9 x 0.6 x 300 x 454 x 8.5 = 94 kN Governs ←
The beam is not principally an axial load carrying member so the MDA for axial load can be
ignored – Combine MDA in shear with actual N* Design Action
Try 3 bolt connection at 130 mm apart with a 35 edge distance
Eccentricity between c of g of bolt group and c of g of weld group which produces an out of
balance moment which can be shared between the weld group and bolt group, taken all by the
bolt group or taken all by the weld group – these notes take the full eccentric moment on
BOTH bolt group and weld group (belts and braces) – note that if the full moment due to
eccentricity is taken by one group only then the eccentricity may change due to the design
action being in a different place from the centroids
Weld Group Centroid - If xb is the distance to the centre of a weld group
Then 530 x xb = 2 x 100 x 50 + 330 x 100 - That is xb = 81 mm
Then Eccentricity of Connection (distance from centroid of bolt group to centroid of

weld group) = 55 (bolts to end of beam) + 81 + 10 (tolerance between beams) = 146 mm
Then M* of bolt group and connection = V* x 0.146
________________________________________________________________________
________________________________________________________________________
= 94 x 0.146 = 13.7 kNm
This moment can be carried by the top and bottom bolts 260 mm apart
Horizontal Force generated by the Moment = 13.7 / 0.26 = 52.7 kN
Sharing V* = 94 kN vertically between three bolts and sharing an axial load N* of 7

kN between the three bolts (say 3 kN) we get a load distribution between the bolts as follows
Resultant Top Bolt = √( 55.72 + 31.32 ) = 64 kN
1 – M16 8.8/S bolt carries 0.8 x 0.62 x 830 x 144 = 59 kN

NG
1 – M20 8.8/S bolt carries 0.8 x 0.62 x 830 x 225 = 93 kN
OK
ADOPT – 3 M20 8.8/S Bolts in Web Plate ←

Web of 460UB is 8.5 mm try say 12 mm webplate (a conservative rule of thumb for the
thickness of the web plate is to duplicate the area of the web of the UB as an initial check (eg
460 x 8.5 / 330 = 11.8 mm)
Check bearing on web from bolts – V* = 55.7 perp to edge of web (web thinner than plate)
9.3.2.4.1 - ϕVb = ϕ3.2dftpfup = 0.9x3.2x20x8.5x430 = 210 kN > 56 kN OK in bearing
Check tearout in web for a 50 mm end distance – V* = 55.7 kN
9.3.2.4.1 - ϕVb = ϕaetpfup = 0.9x50x8.5x430 = 164 kN > 56 kN OK in tearout (also OK on 40

mm and 12 webplate
Check strength of webplate - the webplate is required to transfer the Shear across the
section as well as the out of balance moment
Check Plate to carry 13.7 kNm – Use φM = 0.9 fy Zx of vertical plate
Where zx Plate width b & depth d = bd2/6 – (Elastic Design of the plate gross
dimensions – no need to take account of bolt holes if edge distance > 1.5 df also Code 5.2.7)
Then φM = 0.9 x 300 x bwp x 3302 / 6 = 13.7 E6 (in Nmm units)
bwp = 2.8 mm < 12 mm OK
Check Shear in the Web Plate - (See Code 5.12.3)
0.75 φM = 0.75 x 0.9 x 300 x 12 x 3302 / 6 = 44 kNm >> M* = 13.7 kNm
The φV = 0.9 x 0.6 x 300 x [330 –( 3 x 22 )] x 12 = 513 kN > V* = 94 kN OK
________________________________________________________________________
________________________________________________________________________
By inspection and recycle could go to 10 mm webplate
ADOPT 330 by 10 mm Web Plate ← - See final sketch with dimensions
Consider the Weld Group
Assume 5 mm fillet weld
φVw = 0.522 kN/mm
Then φVH = 0.522 x 100 mm = 52.2 kN
φVV = 0.522 x 330 mm = 172 kN
The Weld Group is required to carry a vertical shear force of 94 kN, and (being
conservative because of the stiffness of the weld group compared with the bolt group), a
Moment due to eccentricity of the connection of 13.7 kNm ( equivalent to a pair of forces 330
apart = 41.5 kN ) and a horizontal axial load of 7 kN (0.43 * 15 kN)
The vertical weld is capable of carrying all the shear
That is – φVV = 172 kN > 94 kN OK
Capacity of the two horizontal welds -
To carry 7 kN (Axial Force) requires 7 / 0.522 = 14 mm total weld or 7 mm each side
Remaining 93 mm (100 – 7) has a φM of
φM = 2 x ( 93 x 0.522 ) kN x 165 mm (moment arm from c of g)
= 16 kNm > 13.9 kNm OK
ADOPT a 5 mm fillet weld group as shown to web plate ←

NOTE if 5 mm size of weld is not enough could try a 6 mm fillet weld – if this still does not
meet the criteria then increase the size (length) of plate and of course the weld length to
achieve what is required
For final sketch see below …
NOTE : The only problem with this sketch is the possible clash between the
inside pair of bolts in the flange and the extreme edge bolts of the web and the
ability of a steelworker to tighten or indeed get a nut on one of the bolts. This
can be remedied by reducing the dimension of 165 and pulling the web bolt
away from the underside of the flange. Discussion with a steel fabricator
would help here
________________________________________________________________________
________________________________________________________________________
FINAL SKETCH FOR 460UB67 BEAM SPLICE
________________________________________________________________________
HERA Report R4-142:2007
Extracts from this report for the Civil 713 lecture
notes on connections
Section 2.2: Welded and Moment

Resisting Connections
The following material comprises extracts from this

HERA report with changes made to the design
example to incorporate the influence of the slab
participation factor in the connection design example.
The first section of this report on Eccentric Cleats in
Compression is not included in these notes as it is
not relevant to Civil 713.
Authors of the HERA Report:
Dr. Charles Clifton, HERA Senior Structural Engineer

Nandor Mago, HERA Finite Element Analyst
Raed El Sarraf, HERA Structural Engineer
Section 2.2 Page 1

3. Design of Columns for Moment-Resisting Connections
Table of Contents for these CIVIL 713 lecture notes

3.1 Introduction
3.2 Column Stiffeners and Panel Zone Design for Welded Moment (WM)
Connections
3.3 Column Stiffeners and Panel Zone Design for Moment-Resisting Endplate (MEP)
Connections
4. Determination of Column Flange and Web Tension Capacity for Bolted Moment-
Resisting Endplate Connections
4.1 Scope of Connections Covered
4.2 Preliminary Assessment of Column Flange Thickness Required
4.3 Determination of Design Tension Actions on Each Bolt Row
4.4 Determination of Column Flange Tension Adequacy
4.5 Calculation of Column Web Tension Adequacy
4.6 Flange Backing Plate Design and Detailing
5. MEP Connection Design Example

5.1 Choose Endplate, Bolts, Weld Details between Endplate and Beam
5.2 Check Column Flange Requirements
5.3 Design of Tension/Compression Stiffeners
5.4 Column Panel Zone
5.5 Check on Column Net Cross-Section Capacity under the Combined Action
References
Note: because the first sections of this report are not included in these notes, the
equation and figure numbers don’t start at 1.
Section 2.2 Page 2

3.1 Introduction
This section provides additional connection design issues required to be considered by

users of HERA Report R4-100:2003 “Structural Steelwork Connections Guide” [1].
These connections are:
• Beam to column welded moment (Designation: WM)

• Beam to column moment-resisting bolted end plate (Designation: MEP)
The connection procedures and details presented in R4-100 [1] are designed to resist
the actions transmitted through the member they support (e.g. the beam in a beam to
column connection). Depending on the connection type, some components of the
connection will be determined by both the supported member and the supporting
member. An example is the horizontal tension/compression stiffeners that will typically
be required for the column supporting a beam through a WM connection.
The approach that has been taken in R4-100 [1] is to provide complete design
guidance for the supported member and for all connection components directly
connected to it, in the case of the WM, MEP and STP connections. This leaves the
designer to complete the remaining connection design, by designing the connection
components that are part of the supporting member. The aim of this section is to
provide design guidance for these connection components not covered by [1].
The design procedures used in developing the WM connections have not changed
significantly from the first edition of R4-100 to the current edition. However, the design
procedures for the bolted moment endplate connections MEP have undergone
significant revision.
The first edition of [1] used the design model presented in NZS 3404 [2] Appendix M,
for MEP connections. However, that procedure resulted in very thick endplates and is
restricted in scope to connections with 4 bolts symmetrically placed about the beam
tension flange. The current edition uses the design model from the Steel Construction
Institute publication Joints in Steel Construction Moment Connections, SCI P207 [9].
This has allowed the use of gusseted endplates and three bolt rows about the beam
tension flange, both of which allow thinner endplates and/or greater moment capacities
to be developed. Extensive FEA on representative connections designed to these
provisions has been undertaken [10, 11] to confirm their performance.
3.2 Column Stiffeners and Panel Zone Design for Welded Moment
(WM) Connections
WM connections require consideration of horizontal tension/compression stiffeners and

the shear strength of the web panel zone. Figure 3.1 shows the plan and elevation of
the WM connection detail with two incoming beams, with the horizontal stiffeners and
panel zone doubler plate reinforcement shown.
Section 2.2 Page 3

Tension/comp ression stiff eners
SE CTI ON B -B
bf
Throat thickness equ als
Doubler plat e plate thickness
Leg (D ) Top weld fillet , bottom we ld

or in comp lete penetration butt (icp)
Leg
A Doubler plat e (in front o f web)

Tension/comp ression
stiffe ners
db
If require d
(D )
B B
EL EVA TI ON
Leg Leg
or (D)
Leg (D )
Leg
SE CTI ON A -A
Stiffen er Stiffen er
Column web
Doubler plat e
Leg
Figure 3.1 Welded moment (WM) connection showing stiffeners and doubler plate.
Notes:
1. This connection shows one doubler plate, fitted on the facing side of the column panel zone.
2. Weld details between the beam and column are not shown; obtain these from HERA Report R4-100
[1].
3.2.1 Design objectives for these components
These are as follows:
1. To transfer the out-of-balance beam flange forces into the column web, without
causing a local buckling or crippling failure in the column.
2. When the beam being connected is subject to inelastic action, to ensure that the
stiffeners do not yield prior to the yielding region developing in the beam.
3. To suppress local buckling when the stiffeners are subject to compression.
4. When the connection is subject to inelastic seismic action, to allow yielding to

dependably develop in the column web panel zone region, but only after the yielding
region(s) in the incoming beam(s) has (have) developed.
Section 2.2 Page 4

5. For a welded I-section column, to ensure that the tension capacity of the column web
to column flange weld within the connection region is sufficient to prevent an
unzipping failure under load.
3.2.2 Tension/compression stiffeners
1. Calculate required maximum and minimum width of each stiffener based on:
bs,max ≥ bs ≥ bs,min (3.1)
bs,max = (bfc-twc)/2 (3.2)
bs,min = min{(bfb-twc)/2;(0.9bfc-twc)/2} (3.3)
where:
bfc = width of column flange

bfb = width of beam flange
twc = thickness of column web
2. Calculate the area of each pair of stiffeners required, using NZS 3404 Clause
12.9.5.3.1 (a) - (c).
More simply, use NZS 3404 Equation 12.9.5.3(4) to determine the required area of
each pair of stiffeners. That equation is reproduced herein as equation 3.4.
⎛ f yb ⎞
As,pair ≥ (bfbtfb - twctfb) ⎜ ⎟ (3.4)
⎜f ⎟
⎝ ys ⎠
where:
fyb = yield stress of beam
fys = yield stress of stiffener
tfb = thickness of beam flange
Equation 3.4 is based on overstrength moment action being developed in the beam.
When this isn't the case, a less conservative answer for stiffener area is obtained from
NZS 3404 NOTE (2), Equations 12.9.5.3(1) to 12.9.5.3(4).
3. Calculate the minimum stiffener thickness necessary to suppress local buckling in

compression from equation 3.5.
⎛b ⎞⎛⎜ f ys ⎞⎟
ts,min≥ ⎜⎜ s ,min ⎟⎟ (3.5)
⎝ C1 ⎠⎜⎝ 250 ⎟⎠
where:
C1 = 8 for the incoming beam being a category 1 or 2 member (see Note 1 from
Clause 12.9.5.3 of NZS 3404 [2])
C1 = 15 otherwise
bs,min = minimum stiffener width from (1) above
Section 2.2 Page 5

4. Select appropriate stiffener width and thickness to satisfy (1) - (3). Use an
appropriate flat bar where possible to minimise the fabrication cost.
5. Calculate fillet weld size for a double-sided fillet weld between stiffener and column
flange at the end of a stiffener adjacent to an incoming beam. This must develop the
design section capacity of the stiffener (i.e. incorporating φ = 0.9), generating a
design force per unit length of weld given by:
0.9 bs ts fys
v*w,s,cf = (3.6)
2 bs
Equation 3.6 assumes that the full width of stiffener contributes to the weld. For
stiffeners close fitted into the column flange/web junction, this is an appropriate
assumption.
6. Select fillet weld leg length, tw,s,cf, such that:
φvw ≥ v*w,s,cf
For consistency with [1], if tw,s,cf > 12mm, use a complete penetration butt weld.
Values of φvw are tabulated in section 9.2.4 of [12].
7. For the welds between the stiffener and column flange at the end of a stiffener
remote from an incoming beam, use tw,s,cf = 5mm. (An example is shown in Figure
3.3 for an MEP connection).
8. Calculate fillet weld size for a double-sided fillet weld between the stiffener and the
column web.
This is given by NOTE (3) to Clause 12.9.5.3.1 of [2]. It is based on the design section
capacity of the stiffener, thus generating a design force/unit length of weld between the
stiffener and the column web or doubler plate given by:
0.9 bst sfys

v*w,s,(cw or dp) = (3.7)
C2 d1c
where:
C2 = 1.0 when 2 beams frame into the column at the connection (see eg. Figure
2.1)
C2 = 2.0 when 1 beam frames into the column at the connection (see e.g. Figure
2.3).
d1c = clear distance between column flanges
9. Select fillet weld leg length, tw,s,cw, such that φvw ≥ v*w,s,cw. Values of φvw are tabulated
in section 9.2.4 of [12].
10. When a doubler plate is to be fitted, the weld on the doubler plate side must be an
incomplete penetration butt weld to allow the doubler plate to fit close to the stiffener,
so that the doubler plate to stiffener weld can be made. This is illustrated in section
Section 2.2 Page 6
A-A of Figure 3.1. For the incomplete penetration butt weld, the design throat
thickness is given by NZS 3404 Clause 9.7.2.3(b). The requirements of (b) (ii) can be
used without needing to refer to AS/NZS 1554.1 [13].
For category SP welds, the depth of preparation for the incomplete penetration butt
weld on the doubler plate side of the horizontal stiffener, dwicp, is given by:
⎛ v * ⎞
dwicp ≥ ⎜⎜ w + 3 ⎟⎟ (3.8)
⎝ 0.48 fuw ⎠
where:
vw* = design load from equation 3.7, in N/mm run
fuw = nominal tensile strength of weld metal
Note dwicp must be specified for this weld on the drawings, rounded up to the nearest
mm. See Figure 3.1 where it is shown as (D) alongside the incomplete penetration butt
weld symbol.
Equation 3.8 is devised by satisfying φvw ≥ vw* for an incomplete penetration butt weld
design throat thickness from NZS 3404 Clause 9.7.2.3(b)(ii)(A) for a single V weld. It
incorporates E48XX grade electrodes and category SP welds, with φ = 0.8.
If dwicp > ts, then a complete penetration butt weld between the horizontal stiffener and
column web is required.
11. Specify stiffener and weld details on the drawings or in the specifications.
3.2.3 WM connections: web panel zone region
3.2.3.1 When using a hot-rolled I-section column or a welded I-section column in

which the column web to column flange welds can develop the design
tension capacity of the column web
• For a BHP welded column (WC) or welded beam (WB), this applies for members
with twc ≤ 16mm.
• Refer to DCB Issue No. 57 [4], page 18, section 2.1 for the background as to why
different provisions apply to welded I-section columns, depending on the capacity of
the column web to column flange weld.
3.2.3.1.1 Factors affecting the need for a doubler plate
For columns of non-seismic-resisting systems, it is unlikely that a doubler plate will be

required.
For columns of seismic-resisting systems with one beam rigidly connected into the
column flange about the major axis, it is also unlikely that a doubler plate will be
required.
For columns of seismic-resisting systems with two beams rigidly connected into the
column about the major axis (e.g. as shown in Figure 3.1), it is likely that a
doubler
Section 2.2 Page 7
plate or plates will be required. If the column is a UC type section, the thickness of
doubler plate reinforcement required will be greater than if the column has a depth to
width ratio similar to a UB. An initial estimate for a UC type column is that the doubler
plate thickness required may be close to that of the column web.
See the design example in section 5.4 for the case of two beams framing into a UB
type column cross section, showing the doubler plate required in that case.
Preferred thickness range for a doubler plate is 5-16mm, for the reasons stated in DCB
No. 57, page 24, section 4.5.
3.2.3.1.2 Design and detailing requirements
Design adequacy is checked to NZS 3404 [2] Clauses 12.9.5.2 and 12.9.5.3.2,
including the slab participation factor.
The background to panel zone design and detailing is given in DCB Issue Nos. 77 and
57. Most of the recommendations are in Issue No. 57, pp. 25-26, with revisions in No.
77, pp 21-23. For ease of use, the design recommendations and detailing requirements
from this general coverage are repeated below.
These requirements should be read in conjunction with Figure 3.1, which shows the
web panel zone region of a WM connection identifying the items covered in this
summary.
1. The doubler plate(s) are fitted between the already welded horizontal stiffeners. It is
(they are) fitted as close against the column web as possible.
The depth of the doubler plate is the clear distance between horizontal stiffeners
less 1mm to allow for fit-up.
The width of the doubler plate is the clear depth between the ends of the column
root radii or welds between column flange and column web; for a hot-rolled column
this is (dc-2tfc-2r1c), where r1 = root radius.
The minimum doubler plate thickness is 5mm; maximum thickness is the lesser of
the column web thickness or 16mm.
2. The weld between the inside face of each stiffener and the column web is an
incomplete penetration butt weld to accommodate the doubler plate. See section
3.1.2(10) for the sizing of this weld.
3. The welds between the top and bottom of the doubler plate and the horizontal
stiffeners, shown in section A-A of Figure 3.1, are sized as follows:
For connections in category 1, 2 or 3 systems, the weld is designed to develop the

design shear yield capacity of the doubler plate, which is given by:
v* w,dp, t and b = 0.9x0.6fy,dptdp (3.9)
Section 2.2 Page 8

where:
fy,dp = yield stress of doubler plate

tdp = thickness of doubler plate
The fillet weld leg length, tw,dp, t and b is chosen such that φvw ≥ v* w,dp, t and b. Values of φvw
are tabulated in section 9.2.4 of [12].
For connections in category 4 systems or in systems not governed by load

combinations including earthquake forces, this weld is sized to resist (V*p - φVc,web) ,
where V*p = design panel zone shear force obtained from the design moment, M*, and
φVc,web = design shear capacity of the web alone.
When grade 250 material is used for the doubler plate and E48 or stronger weld metal
is used, the requirement to size the top and bottom welds to develop the design shear
yield capacity of the doubler plate results in the following standard weld/plate sizes.
tw = 6 mm for tdp = 6 mm
4. For the welds between the sides of the doubler plate and the column flange/web
junction (shown in section B-B of Figure 3.1), it is often most cost-effective to
square cut the doubler plate to fit against the end of the column root radius or flange
to web weld and then to make a butt weld, between the two; an example is shown in
Figure 77.54 of DCB Issue No. 77. The throat of the butt weld should equal the
doubler plate thickness, as shown in the weld note in Section B-B of Figure 3.1. The
weld cross section will look like Figure 4.18(d) of AS 1101.3 [14], however no plate
edge preparation is required.
5. Check the doubler plate slenderness limit:
d p ,max ⎛ f yp ⎞
⎜ ⎟ ≤ 82 (3.10)
t p ⎜⎝ 250 ⎟⎠
where:
dp,max = the maximum clear dimension of the doubler plate (i.e. clear length or clear
depth)
tp = thickness of doubler plate
fyp = yield stress of doubler plate
6. If equation 3.10 is not satisfied, then the doubler plate must be plug welded to the
column web with a single plug weld placed at the centre of the doubler plate, as
shown in Figure 3.1. This plug weld must have a solid diameter given by equation
3.11.
f yp t p
d wp = 5.1 (3.11)
fuw
where:
Section 2.2 Page 9

dwp = diameter of the plug weld

fuw = nominal tensile strength of plug weld metal
For a Grade 250 doubler plate and Grade 480 weld metal, the diameter of plug weld
required from equation 3.11, rounded up to the nearest 5mm, is given by Table 3.1.
This must be specified on the drawings, shown as (D) alongside the plug weld symbol
in Figure 3.1.
Table 3.1 Diameter of single plug weld to Grade 250 doubler plate
Doubler plate thickness Doubler plate fy Plug weld diameter
(mm) (MPa) (mm)
5 280 15
6 280 20
8 260 25
10 260 30
12 260 35
16 250 45
7. The overall panel zone slenderness check, which replaces NZS 3404 Equation
12.9.5.3(6), is given by:
⎛ d c − 2 t fc ⎞⎛ f yp * ⎞
⎜ ⎟⎜ ⎟ ≤ 82 (3.12)
⎜t + k t ⎟⎜ 250 ⎟
⎝ wc 1 p ⎠⎝ ⎠
where:
dc, tfc = column depth and flange thickness
twc = column web thickness
fyp* = as determined from NZS 3404 Clause 12.9.5.3.2.
k1 = 0.25 if the doubler plate is not plug welded to the column web
= 1.0 if the doubler plate is plug welded to the column with a single, central
plug weld of diameter satisfying equation 3.11 or Table 2.
3.2.3.2 When using a welded I-section column in which the column web to
column flange welds cannot develop the design tension capacity of the
column web
3.2.3.2.1 Ensure that the column web alone can satisfy NZS 3404 [2] Clauses 12.9.5.2
and 12.9.5.3.2 without doubler plate reinforcement.
3.2.3.2.2 Ensure that the design tension capacity of the column web to column flange
welds within the connection region and for 100 mm above and below the
incoming beam is not less than the design tension capacity of the column
flange to beam web welds. The column web to column flange welds should be
double-sided fillet welds or double-sided incomplete penetration butt welds or
a complete penetration butt weld.
3.2.4 Welded columns: requirements for column web to column flange welds
within the connection region
These are as given by either section 3.2.3.1 or section 3.2.3.2.2 above.
Section 2.2 Page 10

3.3 Column Stiffeners and Panel Zone Design for Moment-Resisting

Endplate (MEP) Connections
MEP connections require consideration of horizontal tension/compression stiffeners
and the shear strength of the web panel zone. Figure 3.2 shows the plan and elevation
of a MEP, with these components included.
Figure 3.2 MEP component design checks (from [9])

Notes:
1. Column flange bending covers the stiffening option of horizontal stiffener resisting tension, flange
backing plates or both
3.3.1 Scope and objective of design procedure

Figure 3.2 shows the component design check locations for an MEP connection. These
components comprise some or all of: horizontal stiffeners, flange backing plates and
column doubler plates. The design checklist for MEP connections is given in Table 3.2,
along with the location of the design procedure for each item. The procedures in R4-
100 Part 1 are incorporated into the connection details in R4-100 Part 2; those in this
report require designer checks to be made.
The objectives of this design process are:
Section 2.2 Page 11

1. To transfer the out-of-balance beam flange forces into the column web without
causing a local buckling or crippling failure in the column, especially in the column
flanges adjacent to the beam tension flange.
2. When the beam being connected is subject to inelastic action, to ensure that the
stiffeners and column flange do not yield prior to the yielding region developing in the
beam.
3. When the beam being connected is subject to inelastic action, to ensure that bolt
failure alone does not limit the transfer of tension force from endplate to column
flange.
4. To suppress local buckling when the stiffeners are subject to compression.
5. When the connection is subject to inelastic seismic action, to allow yielding to
dependably develop in the column web panel zone region, but only after the yielding
region(s) in the incoming beam(s) has (have) developed.
6. For a welded I-section column, to ensure that the tension capacity of the column web
to column flange weld within the connection region is sufficient to prevent an
unzipping failure under load.
Table 3.2 MEP Component Design Checks
Zone Ref1 Checklist Item Procedure Located In
Tension a Bolt tension R4-100.1 and section 4.3
b End plate bending R4-100.1
c Column flange bending Section 3.3.3
d Beam web tension R4-100.1
e Column web tension Section 3.3.3
f Flange to end plate weld R4-100.1
g Web to end plate weld R4-100.1
Horizontal Shear h Column web panel shear Section 3.3.6
Compression j Beam flange compression R4-100.1
k Beam flange weld R4-100.1
l Column web crushing Section 3.3.4.2
m Column web buckling Section 3.3.4.2
Vertical Shear n Web to end plate weld R4-100.1
p Bolt shear R4-100.1
q Bolt bearing (plate or flange) R4-100.1 and section 3.3.7
Note for table 3.2:
1 See Figure 3.2 for the position of these.
3.3.2 Design actions for column capacity checks
3.3.2.1 Design moment and shear
3.3.2.1.1 Non-earthquake load combinations
The design moment, M*, and shear, V*, are determined from analysis.
3.3.2.1.2 Earthquake load combinations for category 4 MRSFs
Section 2.2 Page 12

For category 4 (elastically responding) moment-resisting steel framed seismic-resisting

systems (MRSFs), the design moment, M*, and shear, V*, for the connection are as
determined from analysis.
Note that these design actions may be less than the beam moment capacity, making
the connection weaker than the beam. The MEP 50/25, MEP 70/35, MEP-G 50/25 and
MEP-G 70/35 connections, from R4-100 Part 2 [1], are examples where the connection
is weaker than the beam.
Connections designed to R4-100 and this Report will exhibit overall elastic behaviour
up to at least 90% of the design moment capacity being reached. Figure 4 from [11]
shows an example of a MEP-G 100/50 limited ductile connection, loaded well into the
inelastic range, that remains elastic up to 94% of the design connection capacity.
In the event of the MRSF being subjected to stronger earthquake action than the
design level event, very minor inelastic demand will occur in the connection region,
which will easily be able to be withstood. This is further facilitated by preventing the bolt
failure mode limiting the connection strength, as stipulated in section 4.4.2.2.
MEP connections designed to these provisions will exhibit a dependable level of

ductility corresponding to that associated with nominally ductile response (μ = 1.25 from
[15]). Sp = 0.9 in accordance with NZS 3404 Clause 12.2.2.1 (Amendment No. 2).
3.3.2.1.3 Earthquake load combinations for category 3 and 2 MRSFs
These are designed as weak beam/ strong column systems, with any ductility demand
occurring in the beam for category 2 MRSFs and typically in the beam for category 3
MRSFs.
The connection must be designed for the over-strength actions from the beam and for
full moment reversal.
The MEP-G 100/50 Limited Ductile connections from R4-100 Part 2 [1] are designed on
this basis for all components given therein. The design moment, M*, is calculated from:
φM con
M*= φ oms (3.13)
0.9
where:
φoms = overstrength factor from NZS 3404 Table 12.2.8(1) for the category 2 beam,
incorporating the slab participation factor where required for the joint panel
zone by Clause 12.10.2.3 (NZS 3404 Amendment No. 2).
φMcon = design moment capacity from R4-100 Part 2
The design shear, V*, is calculated from step 5.3, section 6.2 of HERA Report R4-76
[16].
Also, the bolt failure mode must not limit the connection strength at any bolt row, as
stipulated in section 4.4.2.2.
Section 2.2 Page 13

3.3.2.2 Modification of design moment by axial load
As shown in Figure 3.2, the beam framing into the MEP connection may be subject to
design axial load, N*. This can be accounted for by applying the axial force at the
plastic neutral axis of the beam and calculating the moment induced by this axial force,
M*N, acting about the point of compression contact between beam/end plate and
column.
For the typical case of a beam non-composite in the negative moment region and
comprising an equal flange I-section, the plastic neutral axis is at db/2 from the base of
the compression flange and M*N is given by:
M*N= N* (db-tfb)/2 (3.14)
M*N must be added to M* when designing the connection from R4-100 and this Report,
so that:
1. If N* is tension, the components in the tension zone (see Figure 3.2) are designed
to resist (M*+M*N) and the components in the compression zone are designed to
resist the sum of the bolt actions from applying M* alone.
2. If N* is compression, the components in the tension zone are designed to resist M*

and the components in the compression zone are designed to resist the sum of the
bolt actions from applying (M*+M*N).
3. The column panel zone is designed to resist M*.
3.3.3 Column flange and web requirements
3.3.3.1 Column flange width required
The width of the supporting column flange must be sufficient to carry the bolt layout
specified in [1] plus meet the edge distance requirement of NZS 3404 Clause 12.9.4.4
for seismic-resisting system connections or Table 9.6.2 for non-seismic-resisting
system connections. Using the seismic requirement of 1.5df, this requires a minimum
column flange width, bfc ≥ sg + 3.0df. The bolt gauge, sg, is given in section XVIII of R4-
100: Part 2 [1].
As a simpler and slightly more conservative alternative, check that bfc ≥ bi, where bi =
the endplate width given for the incoming beam from the MEP connection tables in [1].
3.3.3.2 General requirements
Critical actions are bolt tension, column flange bending and column web tension within
the tension zone Table 3.2 and Figure 3.2. They are covered by section 4 herein.
Specifically:
• Bolt tension and column flange bending are covered by sections 4.2, 4.3, 4.4 and
4.6. This covers both the unstiffened column flange capacity and the various
stiffening options available.
Section 2.2 Page 14

• Column web tension is covered by sections 4.3 and 4.5.
3.3.4 Horizontal stiffeners
3.3.4.1 Horizontal stiffeners resisting tension
These are designed as part of the column flange and web tension zone checks. If those
checks show that a pair of horizontal stiffeners is required, they are designed to section
4.4.3.
3.3.4.2 Horizontal stiffeners resisting compression
Critical actions are column web crushing and column web buckling, as shown in Figure
3.2.
The design process is as follows:
1. Calculate the required maximum and minimum width of each stiffener, based on:
bs,max ≥ bs ≥ bs,min (3.15)

(b − t )
bs,max = fc wc (3.16)
2
⎧ (b − t ) (0.9 bfc − t wc ) ⎫
bs,min =min ⎨ fb wc ; ⎬ (3.17)
⎩ 2 2 ⎭
where:
2. Calculate the area of each pair of stiffeners required from the following:
(i) For connectors subject to load combinations including earthquake loads, the need
for a compression stiffener is checked from NZS 3404 Clause 12.9.5.3.1 (a) and
the design undertaken to Clause 12.9.5.3.1 (c).
(ii) Those clauses apply directly to the design actions obtained from section 3.3.2
herein. Where sections 3.3.2.1.1 or 3.3.2.1.2 apply, apply N*fb as specified in Note
(2) to Equations 12.9.5.3 (1) to 12.9.5.3 (4) of NZS 3404, where N*fb is the
compression force N*c generated from section 4.3 herein.
3. Calculate the minimum stiffener thickness necessary to suppress local buckling in
compression, from equation 3.18.
⎛b ⎞⎛⎜ f ys ⎞⎟
ts,min ≥ ⎜⎜ s ,min ⎟⎟ (3.18)
⎝ C1 ⎠⎜⎝ 250 ⎟⎠
where:
C1 = 8 for the incoming beam being a category 1 or 2 member
C1 = 15 otherwise
bs,min = minimum stiffener width from (1) above
Section 2.2 Page 15
4. Select appropriate stiffener width and thickness to satisfy (1) - (3). Use an
appropriate flat bar where possible to minimise the fabrication cost.
3.3.4.3 Welds between horizontal stiffeners and column
The stiffener dimensions and grade to resist tension actions are determined from
section 4.4.3; dimensions and grade to resist compression actions are determined from
section 3.3.4.2. Their location is shown in Figure 3.3.
For reversing loading, the largest values are used.
The welds between the horizontal stiffeners and column are determined as follows:
1. Calculate fillet weld size for a double-sided fillet weld between stiffener and column
flange at the end of a stiffener adjacent to an incoming beam. This must develop the
design section capacity of the stiffener, generating a design force per unit length of
weld given by:
0.9bs t s f ys
v*w,s,cf = (3.19)
2 bs
Equation 3.19 assumes that the full width of stiffener contributes to the weld. For
stiffeners close fitted into the column flange/web junction, this is an appropriate
assumption.
2. Select fillet weld leg length, tw,s,cf, such that:
φvw ≥ v*w,s,cf (3.20)
For consistency with [1], if tw,s,cf > 12mm, use a complete penetration butt weld.
Values of φvw are tabulated in section 9.2.4 of [12].
3. For the welds between the stiffener and column flange at the end of a stiffener
remote from an incoming beam, use tw,s,cf = 5 mm.
4. Calculate fillet weld size for a double-sided fillet weld between the stiffener and the
column web.
This is given by Note (3) to Clause 12.9.5.3.1 of [2]. It is based on the design section
capacity of the stiffener, thus generating a design force/unit length of weld between the
stiffener and the column web or doubler plate given by:
0.9 bs t s f ys
v*w,s, cw = (3.21)
C 2 d1c
where:
C2 = 1.0 when 2 beams frame into the column at the connection (see e.g. Figure
3.1)
Section 2.2 Page 16

C2 = 2.0 when 1 beam frames into the column at the connection (see e.g. Figure
3.3).
d1c = clear distance between column flanges
5. Select fillet weld leg length, tw,s, cw, such that φvw ≥ v*w,s,cw. Values of φvw are
tabulated in section 9.2.4 of [12].
6. When a doubler plate is to be fitted, the weld on the doubler plate side must be an
incomplete penetration butt weld to allow the doubler plate to fit close to the
stiffener, so that the doubler plate to stiffener weld can be made. This is illustrated
in section A-A of Figure 3.3. For the incomplete penetration butt weld, the design
throat thickness is given by NZS 3404 Clause 9.7.2.3(b). The requirements of (b) (ii)
can be used without needing to refer to AS/NZS 1554.1.
For category SP welds, the depth of preparation for the incomplete penetration butt
weld on the doubler plate side of the horizontal stiffener, dwicp, is given by:
⎛ v * ⎞
dwicp ≥ ⎜⎜ w + 3 ⎟⎟ (3.22)
⎝ 0.48 fuw ⎠
where:
vw* = design load from equation 3.21, in N/mm run
fuw = nominal tensile strength of weld metal
Note dwicp must be specified for this weld on the drawings, rounded up to the nearest
mm. See Figure 3.3, where it is shown as (D) alongside the incomplete penetration butt
weld symbol.
If dwicp > ts, then a complete penetration butt weld between horizontal stiffener and
column web is required.
Section 2.2 Page 17

7. Specify stiffener and weld details on the drawings or in the specifications.

Tension/compression stiffeners
SECTION B-B
bf
Doubler plate Throat thickness = plate thickness
Leg (D) Top weld fillet, bottom weld
or Leg incomplete penetration butt (icp)
5
5 A
Tension/compression Doubler plate (in front of web)

A
stiffeners
db
B B
5
5 (D) If required
ELEVATION
Leg (D) Bottom weld fillet, top weld

or
Leg incomplete penetration butt (icp)
Leg
(D)
Leg
SECTION A-A
Stiffener Stiffener
Column web
Doubler plate Leg
Figure 3.3 Moment end plate (MEP) connection showing stiffeners and doubler plate.
3.3.5 MEP to welded columns: requirements for column web to column flange
welds within the connection region
As described in section 4.5.3, over the tension zone of the connection, the column web
to column flange connection must be able to develop the design tension capacity of the
column web. This must extend above the beam tension flange for the distance Leff
determined from Tables 4.1 or 4.2, as appropriate.
For connections subject to reversing action, this will apply adjacent to both flanges of
the incoming beam, effectively requiring full strength welds between the column web
and column flange over the full depth of the connection and for the distance Leff above
and below the beam.
For connections not subject to reversing action, this requirement should extend to
100mm beyond the beam compression flange.
3.3.6 MEP connections: panel zone region
Section 2.2 Page 18

3.3.6.1 Factors affecting the need for a doubler plate
Principal factors influencing the need for doubler plate reinforcement are described in
section 3.2.3.1.1 for moment-resisting frames with WM connections. These factors are
the same for moment-resisting frames with MEP connections, including accounting for
slab participation.
These factors are useful in making a preliminary assessment of the extent of doubler
plate reinforcement required.
3.3.6.2 Design and detailing requirements
Design adequacy is checked to NZS 3404 [2] Clauses 12.9.5.2 and 12.9.5.3.2.
The background to panel zone design and detailing is given in DCB Issue Nos 77 and
57. Most of the recommendations are in issue No. 57, pp 25-26, with revisions in No.
77, pp 21-23. For ease of use, the design recommendations and detailing requirements
from this general coverage are repeated below.
These requirements should be read in conjunction with Figure 3.3, which shows the
web panel zone region of a MEP connection identifying the items covered in this
summary.
1. The doubler plate(s) are fitted between the already welded horizontal stiffeners. It
is/they are fitted against the column web.
The depth of the doubler plate is the clear distance between horizontal stiffeners
less 1mm to allow for fit-up.
The width of the doubler plate is the clear depth between the ends of the column
root radii or welds between column flange and column web; for a hot-rolled column
this is (dc-2tfc-2r1c), where r1 = root radius.
The minimum doubler plate thickness is 5mm; maximum thickness is the lesser of
the column web thickness or 16mm.
2. The weld between the inside face of each stiffener and the column web is an
incomplete penetration butt weld to accommodate the doubler plate. See section
3.3.4.3(6) for the sizing of this weld.
3. The welds between the top and bottom of the doubler plate and the horizontal
stiffeners, as shown in section A-A of Figure 3.3, are sized as follows:
For connections in category 1, 2 or 3 systems, the weld is designed to develop the

design shear yield capacity of the doubler plate, which is given by:
v* w,dp, t and b = 0.9x0.6fyptdp (3.23)

where:
fyp = yield stress of doubler plate

tdp = thickness of doubler plate
Section 2.2 Page 19

The fillet weld leg length, tw,dp, t and b is chosen such that φvw ≥ v* w,dp, t and b. Values of φvw
are tabulated in section 9.2.4 of [12].
For connections in category 4 systems or in systems not governed by load

combinations including earthquake forces, this weld is sized to resist (V*p - φVc,web) ,
where V*p = design panel zone shear force obtained from the design moment, M* (see
sections 3.3.2.1.1 or 3.3.2.1.2), and φVc,web = design shear capacity of the web alone.
When Grade 250 material is used for the doubler plate and E48 or stronger weld metal
is used, the requirement to size the top and bottom to develop the design shear yield
capacity of the doubler plate results in the following standard weld/plate sizes.
tw = 6 mm for tp = 6 mm
4. For the welds between the sides of the doubler plate and the column flange/web
junction (shown in section B-B of Figure 3.3), it is often most cost-effective to
square cut the doubler plate to fit against the end of the column root radius or flange
to web weld, and then to make a butt weld between the two; an example is shown in
Figure 77.54 of DCB Issue No. 77. The throat of the bevel butt weld should be equal
to the doubler plate thickness, as shown in the weld note in section B-B of Figure
3.3. The weld cross section will look like Fig. 4.18(d) of AS1101.3 [14], however no
plate edge preparation is required.
5. The doubler plate slenderness must satisfy:
d p ,max ⎛ f yp ⎞
⎜ ⎟ ≤ 82 (3.24)
t p ⎜⎝ 250 ⎟⎠
where:
dp,max = the maximum clear dimension of the doubler plate (i.e. clear length or clear
depth)
6. If equation 3.24 is not satisfied, then the doubler plate must be plug welded to the
column web with a single plug weld placed at the centre of the doubler plate, as
shown in Figure 3.3. This plug weld must have a solid diameter given by equation
3.25.
f yp t p
d wp = 5.1 (3.25)
fuw
where:
dwp = diameter of the plug weld
fuw = nominal tensile strength of plug weld metal
For a Grade 250 doubler plate and Grade 480 weld metal, the diameter of plug weld
required from equation 2.25, rounded up to the nearest 5mm, is given by Table
3.3.
Section 2.2 Page 20

This must be specified on the drawings, shown as (D) alongside the plug weld symbol
in Figure 3.3.
Table 3.3 Diameter of single plug weld to Grade 250 doubler plate
Doubler plate thickness (mm) Doubler plate fy (MPa) Plug weld diameter (mm)
5 280 15
6 280 20
8 260 25
10 260 30
12 260 35
16 250 45
7. The overall panel zone slenderness check, which replaces NZS 3404 Equation
12.9.5.3(6), is given by:
⎛ d c − 2 t fc ⎞⎛ f yp * ⎞
⎜ ⎟⎜ ⎟ ≤ 82 (3.26)
⎜t + k t ⎟⎜ 250 ⎟
⎝ wc 1 p ⎠⎝ ⎠
where:
dc, tfc = column depth and flange thickness

twc = column web thickness
fyp* = as determined from NZS 3404 Clause 12.9.5.3.2.
k1 = 0.25 if the doubler plate is not plug welded to the column web
= 1.0 if the doubler plate is plug welded to the column with a single, central
plug weld of diameter satisfying equation 3.25 or Table 3.3.
3.3.7 MEP connections: bolt bearing capacity in the column flange

As stated in section IX.B.2 of R4-100: Part 2, the connections developed in R4-100
include a check on the bolt bearing capacity using an assumed tfc = ti, endplate and Grade
250 steel.
If tfc < ti, then the bolt bearing capacity of the bolt group at the compression flange
needs to be checked, in accordance with NZS 3404 Clause 9.3.2.4, to ensure that it is
not less than the design shear, V*. This is achieved if:
φVb,cf ≥ φVfn (3.27)
where:
φVb, cf = bolt design bearing capacity in the column flange
φVfn = bolt design shear capacity
3.3.8 Check on column net cross-section capacity under combined actions
3.3.8.1 General requirement
The presence of bolt holes through the flanges reduces the column cross section area
resulting in the column net cross-section capacity under combined actions at the
connection needing to be checked. This check is to NZS 3404 Clause 8.3. For an I-
Section 2.2 Page 21
section column with the beam framing into the strong axis, as shown in Figure 3.3, the
check is to Clause 8.3.2. In applying these provisions:
1. The net cross section for compression is given by Clause 6.2.1.1 and for tension by
Clause 7.2.1. Note that, if the loss of area is sufficiently small that An=Ag for
compression from Clause 6.2.1.1, the tension capacity will be given by Nt=Agfy rather
than by 0.85kteAnfu
2. The net section modulus for bending is given by Clause 5.2.7.
The column design actions, M*col and N*col, are those applying at the connection.
3.3.8.2 Additional requirement for primary members
In addition to the general requirements of section 3.3.8.1, where the column is a

primary member of the seismic-resisting system in accordance with 12.2.6 and 12.2.7,
the loss of column cross-section area due to the bolt holes must be limited in
accordance with Clause 12.9.4.2.1 or 12.9.4.2.2. Where the column is a secondary
member, this restriction does not apply.
Section 2.2 Page 22

4. Determination of Column Flange and Web Tension
Capacity for Bolted Moment-Resisting Endplate Connections
4.1 Scope of Connections Covered
This section covers column flange and column web to column flange design to resist
the tension forces transmitted through the bolts from MEP-8, MEP-G8, MEP-10, MEP-
G10, MEP-12 and MEP-G12 connections given in HERA Report R4-100 [1].
Isometric views of each of these are shown in Figure 4.1. The design applies to the
column flange adjacent to the beam tension flange. The design procedure is based on
SCI P207 [9].
4.2 Preliminary Assessment of Column Flange Thickness Required

As a general recommendation for preliminary design, where the column flange is
stiffened with horizontal stiffeners adjacent to the beam tension flange and the design
section moment capacity or the overstrength moment capacity of the incoming beam
must be resisted, equation 4.1 gives a preliminary estimate of the column flange
thickness required to resist the moment-induced tension actions. This equation is
derived from Plastic Design of Low-Rise Frames [18], with the influence of the column
free end and earthquake overstrength added, based on the HERA Senior Structural
Engineer’s determination.
⎛ f ⎞
t
fc
≈ 0 .5 C φ
3 oms
(
0 .9 t + t ⎜
i
)
⎜ yi
fb ⎜ f
⎟
⎟
⎟
(4.1)
⎝ y , cf ⎠
where:
ti = thickness of endplate, from R4-100 [1] (endplate ≡ cleat from [1])
tfb = thickness of beam flange
fyi = design yield stress of endplate
fy,cf = design yield stress of column flange
C3 = 1.0 for the connection adjacent to a column free end
C3 = 0.95 for the connection not adjacent to a column free end
φoms = overstrength factor, if applicable (use 0.9 if connection is non-seismic and
1.0 if category 4 seismic-resisting system)
When less than the design section moment capacity of the beam is required to be
resisted, this equation will overestimate the column flange thickness required.
If the actual column flange thickness is less than that given in equation 4.1, it means
that additional split backing plates are likely to be required in addition to the horizontal
stiffeners.
Section 2.2 Page 23

4. Determination of Column Flange and Web Tension Capacity for Bolted Moment-Resisting Endplate
Connections
Beam tension
flange Beam tension
flange
MEP-8: two bolt row connection MEP-G8: two bolt row connection
Beam tension
Beam tension flange
flange
MEP-10: three bolt row connection MEP-G10: three bolt row connection
Beam tension Beam tension

flange flange
MEP-12: three bolt row connection MEP-G12: three bolt row connection
Figure 4.1 Connection types and definitions

Definitions for Figure 4.1
1. A two bolt row connection has one bolt row above the beam tension flange and one below. This
covers the MEP-8 and MEP-G8
2. A three bolt row connection has one bolt row above the beam tension flange and two below
3. The MEP-10 and MEP-G10 connections have three bolt rows in one direction of moment and two bolt
rows in the other
4. The MEP-12 and MEP-G12 connections have three bolt rows for both directions of moment
Section 2.2 Page 24

4. Determination of Column Flange and Web Tension Capacity for Bolted Moment-Resisting Endplate Connections
4.3 Determination of Design Tension Actions on Each Bolt Row

The design tension actions on each bolt row generated by the design moment, M*, are
given by Figure 4.2 or 4.3 and the following:
Figure 4.2 Flexural actions from joint for the design of the column flange (for three rows
of tension bolts)
Figure 4.3 Flexural actions from joint for the design of the column flange (for two rows
of tension bolts)
For a three bolt row connection:
A = d b − 0.5t fb − pf (4.1)
B = A − sp (4.2)
M* (4.3)
N r*1 =
⎡ B ⎤
2
⎢2(d b − t fb ) + ⎥
⎣ A⎦
N r*2 = N r*1 (4.4)
Section 2.2 Page 25
Connections
⎛B⎞ (4.5)
N r*3 = ⎜ ⎟N r*1
⎝ A⎠
where:
M* = design moment as given by section 3.3.2.1 herein.

db = beam depth
tfb = beam flange thickness
pf, sp = as given by R4-100 [1]
For a two bolt row connection:

M* (4.6)
N r*1 = N r*2 =
2(d b − t fb )
For both connections;

N c* = ∑ N r* (4.7)
where:
∑ N r = sum of the tension bolt row forces

*
These expressions apply for applied moment only; in the presence of axial load an
additional moment is generated and applied to either the bolts in tension or to the
flange in compression, as specified by section 3.3.2.2.
4.4 Determination of Column Flange Tension Adequacy
4.4.1 Calculation of effective tee stub length, Leff
Leff is dependent on the location of the bolt row to the column web, the presence of
column horizontal stiffeners, a free end and/or adjacent bolt rows. Leff for the various
possible options are given in Table 4.1 or Table 4.2, as appropriate, and Figure 4.7,
when required.
4.4.2 Tension resistance of the column flange
4.4.2.1 General
This is determined for each bolt row independently.
For a bolt row, the design tension capacity of the column flange, φNr, is given by the
minimum of modes 1, 2 and 3. These modes are described on pages 18 and 34 of SCI
Report P207/95 [9] and in Part 1 of HERA Report R4-100:2003 [1].
Mode 1: Complete flange yielding; no bolt extension
4φM pc + 2φM pbp

φN r ,m o d e 1 = (4.8)
m1
Section 2.2 Page 26

Mode 2: Partial flange yielding; partial bolt extension
2φM pc n ∑ (φN tf )
φN r ,m o d e 2 = + (4.9)
(m1 + n ) (m1 + n )
Mode 3: No flange yielding; full bolt extension
φN r ,m o d e 3 = ∑ (φN tf ) (4.10)
where:
φMpc = design plastic moment capacity of equivalent column flange tee stub
0.9Leff ,cf t fc2 f y ,cf η
= (4.11)
4
φMpbp = design plastic moment capacity of equivalent backing plate tee stub
(if backing plates are fitted)
2
0.9Leff ,bp t bp f y ,bp
= (4.12)
4
Leff,cf = effective length of yield line for column flange tee stub, from Table 4.1, or
Leff from Table 4.2
Leff,bp = effective length of yield line for backing plate tee stub, from Table 4.1, or
Leff from Table 4.2.
tfc = column flange thickness
tbp = backing plate thickness
fy,cf = nominal yield stress of column flange
fy,bp = nominal yield stress of backing plate
η = column axial load reduction factor, given in NZS 3404 after Equation
12.9.5.3(5)
2
= ⎛N* ⎞
1.15 − ⎜⎜ ⎟⎟ ≤ 1.0
⎝ φN s ⎠
N *
= ratio of column design compression force to design section capacity
φN s
Ns = nominal section capacity from NZS 3404, Clause 6.2.1 [2]
φ = 0.9, from NZS 3404, Table 3.3(1)
m1 = as defined in Tables 4.1 and 4.2
n = effective edge distance (see Figure 4.5)
n = ⎧b sg ⎫ (4.13)
min ⎨ fc − ;1.25 m1 ⎬
⎩ 2 2 ⎭
bfc = column flange width
sg = bolt gauge, from R4-100.2, section XVII.B.3
∑ φN tf = design tension capacity of the two bolts in the row
φNtf = bolt design tension capacity, from Clause 9.3.2.2 of NZS 3404 or as listed
in section 9.1.3(b) of the ASI Design Capacity Tables [12]
For each bolt row, φN r ≥ N r* is required.

Section 2.2 Page 27
Connections
4.4.2.2 Strength hierarchy requirement for earthquake-resisting connections.
For connections designed for earthquake load combinations (see design moment and
shear from either 3.3.2.1.2 or 3.3.2.1.3), mode 3 must not be the governing mode, in
order to generate dependable deformation capacity in the connection.
4.4.3 Design of column horizontal stiffeners to resist tension
As shown in Figures 4.2 and 4.3, the horizontal stiffener pair resisting tension is located
between rows 1 and 2.
The design requirements for the area of this stiffener pair are taken from step 6C of [9],
with the stiffeners being full depth across the column.
1. The stiffener width, bs, must satisfy equation 4.14
bs ,max ≥ bs ≥ bs ,min (4.14)
where:
(bfc − t wc ) (4.15)
bs ,max =
2
⎧ (b − t ) (0.9 bfc − t wc ) ⎫ (4.16)
bs ,min = min ⎨ fb wc ; ⎬
⎩ 2 2 ⎭
where:
2. The area of the pair of tension stiffeners between bolt rows 1 and 2 must satisfy
equations 4.17 and 4.18
Ast ,pair ≥
(N *
+ Nr*2
r1 ) − Ltw twc (4.17)
0.9fys ,min
and
m1 ⎡ Nr*2 Nr*1 ⎤
A st,pair ≥ ⎢ + ⎥ (4.18)
0.9fys,min ⎣⎢ (m1 + m22 ) (m1 + m21 ) ⎦⎥
where:
N r*1, N r*2 are determined from section 4.3

fys,min = lesser of the stiffener nominal yield stress or the column web nominal
yield stress
Ltw = effective length of web resisting tension, as given by Figure 4.4
= 1.8sg+p for connection not near a column free end
= ex+0.9sg+p for connection near a column free end
ex = distance from centre of bolt row 1 to free end (see eg. Case 2, Table 4.1)
sg, p = are given in HERA Report R4-100
m1 = as defined in Tables 4.1, 4.2 herein
m21, m22 = as defined in the notes to Figure 4.5
Section 2.2 Page 28
m21
m22
n
Figure 4.5 Geometry for force distribution to
horizontal stiffeners
Notes to Figure 4.5
If the top of the stiffener is at the same elevation as the
top of the steel beam (which is the normal case), then
m 21 = a f − 0.8t w ,s ,cf
m 22 = pf − t s − 0.8t w ,s ,cf
pf, af are given in HERA Report R4-100 Part 1
tw,s,cf = leg length of fillet weld between stiffener and
column flange = 0 (if weld is a complete or an
incomplete penetration butt weld)
Figure 4.4 Effective web lengths for

rows 1 and 2
3. The stiffener thickness must satisfy
Ast ,pair
tst ,min ≥ (4.19)
2bs
4. Select appropriate stiffener width and thickness to satisfy (1) - (3). For reversing
loading, use the greater of bs, ts for the tension resistance, from above, and bs, ts for
the compression resistance, from section 3.3.4.2
These stiffener dimensions are also used for sizing the welds between the stiffener and
column flange or between the stiffener from this section and column web. Design of
these welds for the stiffener sizes chosen is covered by section 3.3.4.3.
4.5 Calculation of Column Web Tension Adequacy
4.5.1 For bolt row 3 in a three bolt connection, with a horizontal stiffener between
bolt rows 1 and 2
φN r 3,cw = 0.9Lt ,3 t wc f y ,cw (4.20)
Section 2.2 Page 29

Connections
where:
Lt,3 = 0.5sp+0.9sg
fy,cw = nominal yield stress of column web
For the column web adjacent to the bolt row to be satisfactory without the need for an
additional tension stiffener;
N r* 3 ≤ φN r 3 ,cw is required (4.21)
where:
N r*3 is given by equation 4.5
4.5.2 For a column with no horizontal tension stiffeners
(1) The bolt rows individually must satisfy equations 4.22 and 4.23
φN ri ,cw = 0.9 x1.8s g t wc f y ,cw (4.22)

N r*,max(1.2.3 ) ≤ φN ri ,cw (4.23)
Then, either (2) or (3) apply as appropriate:
(2) For two bolt rows, equations 4.24 and 4.25 must also be satisfied:
φN r 12,cw = 0.9Lt ,12 t wc f y ,cw (4.24)
where:
Lt,12=1.8sg+p
sg,p are obtained from HERA Report R4-100
N r*1 + N r*2 ≤ φN r 12,cw is required (4.25)
(3) For three bolt rows, equations 4.26 and 4.27 must also be satisfied:
φN r 123,cw = 0.9Lt ,123 t wc f y ,cw (4.26)
where:
Lt,123=1.8sg+p+sp
sp is obtained from HERA Report R4-100 [1]
N r*1 + N r*2 + N r*3 ≤ φN r 123,cw is required (4.27)
4.5.3 Capacity of column web to column flange connection
Section 2.2 Page 30

Over the effective tension length, Lt, from the above section, the column web to column
flange connection must be able to develop the design tension capacity of the column
web.
For a hot-rolled column, this is always satisfied.
For a three plate welded column, the weld size must be checked.
Note that for reversing loading, this check is required adjacent to both beam flanges.
4.6 Flange Backing Plate Design and Detailing
Read the requirements below in conjunction with Figure 4.6.
Where pairs of backing plates are used in conjunction with a horizontal stiffener, one
plate is placed above the stiffener and one below, as shown in Figure 4.6.
The thickness of each plate above the stiffener is determined for bolt row 1, from
section 4.4.2.
Each backing plate below the stiffener covers the column flange around bolt row 2 and
bolt row 3 (if present). The thickness of this plate is the maximum of that required for
each bolt row.
The backing plates above the stiffener and below may be different thickness, as in the
design example analysed in HERA Report R4-120 [11] and presented in section 5
herein.
In all instances, t bp ≤ 1.5 t fc is recommended
Where there is no horizontal stiffener, the plate thickness is the maximum from that
required for each bolt row and the plate covers all bolt rows.
L’bp1
L’bp3
A A
Section A-A
(horizontal compression stiffener

below not shown for clarity.
Section 2.2 Page 31

Connections
Figure 4.6 Column flange backing plates
The width of the backing plate, bbp, should not be less than the distance from the edge
of the flange to the toe of the root radius, and it should fit snugly against the root radius,
as shown in Figure 4.6
The length of the backing plate L’bp above the row 1 bolt centreline and below the
bottom row bolt centreline (either row 2 or row 3 as appropriate) is given by the
maximum of the following:
(1) Where a horizontal stiffener is adjacent to the bolt row (as for bolt row 1 in Figure
4.6)
L’bp = max{0.25Leff,cf ; 0.9sg ; ex(if free end) ; 2df}
(2) Where no horizontal stiffener is adjacent to the bolt row (as for bolt row 3 in Figure
4.6)
L’bp = max{0.5Leff,cf ; 0.9sg ; 2df}
Where the connection is subject to reversing action, it may be more economical to

make the backing plate behind the bolt rows inside the beam flanges full depth between
the horizontal stiffeners, rather than two separate plates. The design example in section
5 shows an illustration of this.
Backing plates are generally supplied loose, or tack-welded in place for ease of
erection. They are not welded all round to the column flange.
Section 2.2 Page 32

Table 4.1: Leff for equivalent Tee stub, column flange with horizontal tension
stiffeners
Bolt
Diagram Leff Equations
Row
m1 e
Case 1:
For row 1, all connections,
away from a column free end:
Row1 Leff ,cf = αm1 + t bp

m2
Leff ,bp = αm1
p
Row2
Free End
Case 2:
ex
For row 1, all connections,
Row1
near a column free end:
m2
p Leff ,cf = αm1 − ( 2m1 + 0.625e ) + ex + t bp
m1 e
Leff ,bp = αm1 − ( 2m1 + 0.625e ) + ex
Row2
Row1
Case 3:
For row 2, three row connection
m2
L1 = 4m1 + 1.25e
Row2
L2 = α m1
Leff ,cf = max {0.5L1 ; L2 − 0.5L1 } + 0.5s p + t bp
sp
e m1
Row3
Leff ,bp = max {0.5 L1 ; L2 − 0.5 L1 } + 0.5 s p
Row2 Case 4:
For row 3, three row connection
sp Leff ,cf = 0.5 L1 + 0.5 s p
Leff ,bp = 0.5 L1 + 0.5 s p

Row3
e m1
Section 2.2 Page 33

Connections
Row1 Case 5:
For row 2, two row connection
p Leff ,cf = αm1 + t bp
m2
Leff ,bp = αm1
Row2
m1 e
Table 4.2: Leff for equivalent Tee stub, column flange without horizontal tension
stiffeners
Bolt
Diagram Leff Equations
Row
e m1
Case 1:
Row 1, all connections,
away from a column free end
Row1
L1 = 4 m1 + 1.25 e
p Leff = 0.5 L1 + 0.5 p
Row2
Free end
Case 2:
ex Row 1, all connections,
Row1
near a column free end
p Leff = min {ex ; 0.5 L1 } + 0.5 p
Row2
Section 2.2 Page 34

Row1
Case 3:
Row 2 in a three row connection
p
Leff = 0.5 p + 0.5 s p
Row2
sp
Row3
Case 4:
Bottom row in either
p or sp
a 2 row or a 3 row connection
Bottom Row Leff = 0.5 L1 + 0.5( p or s p )
e m1
Definitions for Tables 4.1 and 4.2

m1 = distance from bolt centre to 20% distance into the column root radius or the column web to
flange weld leg length
sg
− 0.8 (rc or t w ,cw ,cf )
= t wc
−
2 2
e = distance from bolt centre to edge of column flange
= bfc s g
−
2 2
bfc = column flange width
sg = bolt gauge, from R4-100 [1], Part 2, Section XVII.B.3
rc = root radius of rolled column
tw,cw,cf = leg length of weld between column web and column flange, three plate welded column
α = as given by Figure 2.16 of SCI P207/95 [9], or Figure 4.7 herein or using the equations given
in either [9] or [1] and which are given in these notes in section 2.3, page 4
m2 = distance from bolt centre to 20% into the horizontal stiffener to column flange weld leg length
(see Figure 4.7)
p = pitch between bolt row 1 and bolt row 2
= pf+af ; these given in R4-100 Part1.
sp = pitch between bolt row 2 and bolt row 3 in a three row connection, as given in R4-100 Part 1
tbp = thickness of split backing plate, if fitted
Notes for Table 4.1, applying to columns with horizontal stiffeners resisting tension:
1. Leff is the effective length of the equivalent tee stub for the bolt row under consideration.
2. When a backing plate is fitted and the equation for Leff includes tbp, this is applied only to the column
flange and not to Leff for backing plate. This is why, in Table 4.1, there are two values of Leff given,
Leff,cf and Leff,bp. See the explanation for this below.
3. Case 1 or case 2 applies for the top row of bolts in all connections.
4. Case 3 and case 4 apply for the second and third row of bolts respectively in a three row connection
5. Case 5 applies for the second row of bolts in a two row connection.
6. Min {expression1; expression2} means use the minimum value of these expressions.
Max {expression1; expression2} means use the maximum value of these expressions
Section 2.2 Page 35

Connections
SCI Publication P207 [9] covers the use of backing plates in lieu of horizontal stiffeners, but does not
cover their use in conjunction with horizontal stiffeners. When used in combination, the backing plate
increases the tension resistance available from the column flange through two mechanisms, namely;
(i) Its load spreading action mobilises a larger yielding region of column flange, increasing the
internal work generated and hence the tension resistance
(ii) It reduces the distance between the edge of bolt load application onto the column flange and
the adjacent column web or horizontal stiffener, which reduces the applied moment
generated by the tension force, thus increasing the tension resistance for the given column
flange moment capacity
These influences are seen in comparing the yield line actions in the backing plate and column flange,
Figures 29 and 30 and Figures 34 and 35 of [11].
When the ratio of N*r/φNr for each bolt row is determined and the governing bolt mode for the joint
analysed in [11], the closest match between observed and calculated results is obtained with the addition
of tbp to the determination of Leff,cf. These values are:
Row 1: N*r /φNr = 0.91, mode 1 governs

Notes for Table 4.2, applying to columns without horizontal stiffeners resisting tension:
1. Leff is the effective length of the equivalent tee stub for the bolt row under consideration, from [9]
2. Case 1 or case 2 applies for the top row of bolts in all connections
3. Case 3 applies to row 2 in a three row connection
4. Case 4 applies to the bottom row in either a two row or a three row connection. Pitch “p” applies for a
two row connection, pitch “sp” applies for a three row connection
5. See note 6 from Table 4.1
Section 2.2 Page 36

Bolted EndEnd
Bolted PlatePlate
Connection
Figure 4.7
Values of α for
stiffened column
flanges and
endplates (from
[9])
Note: Mathematical expressions for α are given in [1, 9]; see section 2.3, page 4
Section 2.2 Page 37

Section 2.2 Page 38
Two 610UB101 Grade 300 beams are connected to a 914UB201 Grade S275 column,
one each side, as shown in Figure 5.1. The column is a British Steel made section. The
beams and the moment-resisting steel framed (MRSF) system are category 2. The
beams are the primary elements; the columns are secondary elements. The slab is
isolated from the column so that the slab participation factor is not required to be
included.
The connection is located at a lower level of the frame, hence is not adjacent to a
column free end (i.e. the top of a column). In this design example, the beam framing
into the column from the right hand side (see Figure 5.1) is assumed to have its
negative moment end (as defined in Clause 1.3 of NZS 3404) at the column and the
beam framing into the column from the left hand side has its positive moment end at
the column.
Design actions (ignoring sign of moment) are:

M*R = 783 kNm, V*R = 350 kN
M*L = 730 kNm, V*L = 105 kN
N*col = 2000 kN
5.1 Choose Endplate, Bolts, Weld Details between Endplate and Beam
These are obtained directly from the MEP-G tables of HERA Report R4-100: Part 2 [1].
The connection chosen must have φMcon ≥ max(M*R; M*L). This connection type is
required for the category 2 seismic-resisting system. From that table, including the
dimensions of the bolt and endplate thickness:
φMcon = M*R and φVcon > V*R ∴ OK
Bolt size: M36
Bolt row 1 backing plates (90 deep x 145 wide x 10 mm thick)
Endplate thickness, ti = 25 mm
Same for
bottom 12
Gusset plate as shown
stiffener in R4-100:Part 2 [1]
R1
R2 Bolt row 2 and 3 backing plates

(520 deep x 145 wide x 25 mm thick)
R3
Same for
bottom 5
stiffener 6
plug weld doubler plate
R3 to column web, 20 mm
20
R2 diameter
R1
Tack weld in 4 910 UB 201 column

place only Changes due to
inclusion of the slab
participation factor
Figure 5.1 MEP tension/compression stiffeners
• Plug weld 25mm
Connection used in 610 UB 101 beam
dia
design example. • Doubler plate
8mm thick, 8mm
fillet weld to the
Doubler plate 824 wide x 546 deep x 6 mm thick (6mm fillet weld to
the top and bottom of the plate, butt weld down the sides) Section 2.2 Page 39
5.2 Check Column Flange Requirements
5.2.1 Column flange width required
See requirement from section 3.2.3.1 herein.
sg taken from section XVII of R4-100: Part 2 [1].
bfc ≥ sg + 3.0df = 248 mm

bfc,supplied = 303 mm (see [17])
sg = 140 mm
df = 36 mm
Condition is met.
5.2.2 Preliminary design check on column flange transverse tension capacity

when stiffened with a tension stiffener
This check uses equation 4.1 in section 4.2 herein, multiplied by:
tfc,reqd ≈ 0.5C3φoms (0.9ti + tfb)(fyi/fy,cf) = 26.91 mm

C3 = 0.95 (column not at free end)
φoms = 1.15 (from Table 12.2.8(1) of [2])
ti = 25 mm (from Part 2 of [1])
tfb = 14.8 mm (from page 18 of [17])
fyi = 350 MPa (see Part 2 of [1])
fy,cf = 265 MPa (see Part 2 of [1] for the given column flange thickness)
tfc,supplied = 20.2 mm, from [17], for a 914UB201.
Hence the stiffened column flange tension capacity, when stiffened with a tension
stiffener, is not likely to be adequate and the additional split backing plate stiffening
given in section 4 herein is likely to be needed. This is checked in section 5.3.2 below.
5.3 Design of Tension/Compression Stiffeners
5.3.1 Sizing of horizontal stiffeners resisting compression
This is covered in section 3.3.4.2 herein.
1. Calculate minimum width of each stiffener
bfc − t wc
bs,max = = 144 mm
2
⎛ (b − t ) (0.9 bfc − t wc ) ⎞
bs,min = min ⎜ fb wc ; ⎟
⎝ 2 2 ⎠
= 106 mm
bfb = 227.6 mm
bfc = 303 mm
twc = 15.1 mm
Section 2.2 Page 40
2. Calculate area of each pair of stiffeners required.
Check need for compression stiffener from NZS 3404 Clause 12.9.5.3.1 (a):
φ oms Afb ⎛ f yb ⎞
twc < ⎜ ⎟
t fb + 5 t fc + 2 t ep + 2 t wf ⎜f ⎟
⎝ yc ⎠
∴twc< 26.45 mm limit for no stiffener

twc = 15.1 mm ∴Stiffener is required
tfc = 20.2 mm
φoms = 1.15
tep = 25 mm (designated as ti in R4-100 [1])
twf = 0 mm (as butt weld is used between beam flange and endplate)
Afb = 227.6x14.8 = 3368 mm2
fyb = 300 MPa
fyc = 265 MPa
fys = 250 MPa
⎛ f yb ⎞ ⎛f ⎞
Asc≥ φ oms Afb ⎜ ⎟ − t wc (t fb + 5 t fc + 2 t ep + 2 t wf )⎜ yc ⎟
⎜f ⎟ ⎜f ⎟
⎝ ys ⎠ ⎝ ys ⎠
Asc ≥ 1995 mm2 (this is for the pair of stiffeners)
3. Calculate minimum stiffener thickness necessary to prevent local buckling in

compression.
⎛ bs ⎞⎛⎜ f ys ⎞⎟
ts,min ≥ ⎜⎜ ,min ⎟⎟ = 13.3 mm
⎝ C1 ⎠⎜⎝ 250 ⎟⎠
C1 = 8 as the incoming beam is a category 2 member
4. Select appropriate stiffener thickness, width.
Use bs = 130 mm and ts = 16 mm
As = 2x130x16 = 4160 mm2 ∴OK
5. Calculate fillet weld size required for the double-sided fillet weld between stiffener
and column flange at the ends adjacent to the incoming beams.
This is covered in section 3.3.4.3 herein.
0.9bs t s f ys
v*w,s,cf = = 1.8 kN/mm
2 bs
fys = 0.250 kN/mm2
Section 2.2 Page 41

6. Select suitable fillet weld leg length.

φvw = 1.96 kN/mm for tw = 12 mm, E48XX category SP.
Adopt tw,s,cf = 12 mm fillet weld size
7. Calculate fillet weld size for double-sided fillet weld between stiffener and column
web from equation 3.21.
0.9 bs t s f ys
v*w,s, cw = = 0.57 kN/mm
C 2 d1c
C2 = 1.0 (2 beams framing into column)

d1c = 824.4 mm (from [17])
8. Select suitable fillet weld leg length.
φvw = 0.82 kN/mm for tw = 5mm, E48XX Category SP

Adopt tw,s,cw = 5mm fillet weld size.
As described and calculated in section 5.4, a doubler plate is fitted to one side of the
beam web. The stiffener to column web weld on this side must be an incomplete
penetration butt weld; the required depth of penetration is given in section 5.4(6).
5.3.2 Check column flange adequacy

This is covered in sections 4.3 to 4.6 herein.
1. Determination of Design Tension Actions on Each Bolt Row
Calculate M* from section 4.3:
φM con
M* = φ oms = 1000.5 kNm
0 .9
A = db − 0.5tfb − pf = 505.2 mm
B = A − sp = 415.2 mm
db = 602.6 mm
pf = 90 mm
sp = 90 mm
M* = 659.3 kN
Nr*1 =
⎡ B ⎤
2
⎢2( d b − t fb ) + ⎥
⎣ A⎦
∴ N*r2 = N*r1 = 659.3 kN
⎛B⎞
Nr*3 = ⎜ ⎟Nr*1 = 541.8 kN
⎝ A⎠
N c* = ∑ N r* = 659.3 + 659.3 + 541.8 = 1860.4 kN
Section 2.2 Page 42
2. Determination of Column Flange Tension Adequacy
(a) Calculation of effective tee stub length, Leff.
NOTE: The calculations shown below are the final results of a number of iterations of
this design to get to the optimum design. This is done by assuming a different backing
plate thickness, until the closest match of S* to φRu is obtained.
Check for row 1 (R1 in Figure 5.1): Use tbp = 10 mm thick backing plate.
Use Case 1 in Table 4.1 for the calculation of Leff.
Leff ,cf = αm1 + t bp = 306 mm
Leff ,bp = αm1 = 296 mm
sg twc
m1 = − − 0.8rc = 47.17 mm
2 2
m21 = af − 0.8t w ,s,cf = 45.4 mm
rc= 19.1 mm
af = 55 mm
Calculate λ1 and λ2 to use in Figure 4.7 to find the value of α.
m1
λ1 = = 0.36
m1 + e
m21
λ2 = = 0.35
m1 + e
e= bfc s g = 82 mm
−
2 2
∴ α= 2π as stated in Figure 4.7.
Check for row 2 (R2 in Figure 5.1) Use tbp = 25 mm thick backing plate.
Leff ,cf = max (0.5L1 ; L2 − 0.5 L1 ) + 0.5 sp + tbp =220.4mm
Leff,bp = max(0.5L1 ;L2 − 0.5L1 ) + 0.5sp = 195.4 mm
L1 = 4m1 + 1.25e = 291.2 mm
Section 2.2 Page 43

L 2 = α m 1 = 296 mm
sg
twc
m1 = −
− 0.8rc = 47.2 mm
2 2
m = p − t − 0 .8 t = 64.4 mm
22 f s w , s , cf
Calculate λ1 and λ2 to use in Figure 4.7 to find the value of α.
m1
λ1 = = 0.36
m1 + e
m22
λ2 = = 0.50
m1 + e
∴ α= 2π from in Figure 4.7.
Check for row 3 (R3 in Figure 5.1): Use tbp = 25mm thick backing plate.
Leff ,cf = 0.5L1 + 0.5s p = 190.6 mm
Leff ,bp = 0.5L1 + 0.5s p = 190.6 mm
L1 = 4m1 + 1.25e = 291.2 mm
(b) Tension resistance of the column flange
Check for row 1:
4 φ M pc + 2 φ M pbp
φ N r ,m o d e 1 = = 705 kN
m1
φN r ,m o d e 2 = + = 752 kN
(m1 + n ) (m1 + n )
φN r ,m o d e 3 = ∑ (φN tf ) = 1082 kN
0.9Leff ,cf t fc2 fy ,cfη

φMpc = = 7.45 kNm
4
2
0.9Leff , bp t bp f y , bp
φMpbp = = 1.73 kNm
4
Section 2.2 Page 44
fy,cf = 265 MPa

fy,bp = 260 MPa for 10 mm plate and 250 MPa for 25 mm plate
2
η = ⎛ N*
1.15 − ⎜⎜
⎞
⎟⎟ ≤ 1.0 = 1.0
⎝ φN s ⎠
N* = 0.33
φN s
Ns = 6784 kN (see Note below)
φ = 0.9
n = ⎧b s ⎫ = 59 mm
min ⎨ fc − g ;1.25 m1 ⎬
⎩2 2 ⎭
∑ φN tf = 2x541 = 1082 kN
φNtf = 541 kN
Check φN r ≥ N r*
φNr = min( 705 ; 752 ; 1082) = 705 > Nr* = 659 ∴ OK
Mode 3 is not critical ∴ OK
Note on Ns:
Calculate the value of Ns from NZS 3404, Clause 6.2.1 [2], which states that:
Ns = kfAnfy = 6784 kN
where:
Ae
kf = = 1 (see Clause 6.2.3 in [2])
Ag
An = net area of the cross section, except that for sections with penetrations or unfilled
⎡ fy ⎤
holes that reduce the section areas by less than 100{1- ⎢ ⎥ } %, the gross area
⎣ (0.85fu )⎦
may be used. Based on this condition, when the ratio of An/Ag ≥ 0.76; the gross area Ag
may be used.
For these connections;
Ag = 25600 mm2
An = 25600 – (4x39x20.2)
= 22449 mm2
An/Ag = 22449/25600
= 0.88 ⇒ the gross cross section can be used.
Section 2.2 Page 45

Check for row 2:
4φ M pc + 2φ M pbp
φ N r ,m o d e 1 = = 746 kN
m1
φN r ,m o d e 2 = + = 702 kN
(m1 + n ) (m1 + n )
φN r , m o d e 3 = ∑ (φNtf ) = 1082 kN

φMpc = = 5.36 kNm
4
2
0.9Leff ,bp t bp f y , bp
φMpbp = = 6.87 kNm
4
φN r = min( 746;702;1082 ) = 702 > N r* = 659 ∴ OK

Check for row 3:
φN r ,m o d e 1 =
4φM pc + 2φM pbp = 669 kN
m1
φN r ,m o d e 2 = + = 686 kN
(m1 + n ) (m1 + n )
φN r , m o d e 3 = ∑ (φNtf ) = 1082 kN

φMpc = = 4.54 kNm
4
2
0.9Leff , bp t bp f y , bp
φMpbp = = 6.7 kNm
4
Section 2.2 Page 46

φN r = min( 669;686;1082 ) = 669 > N r* = 542 ∴ OK

(c) Design of column horizontal stiffeners to resist tension
(i) The stiffener width, bs, must satisfy equation 4.14
bs,max ≥ bs ≥ bs,min
(bfc
− t wc ) = 144 mm
bs ,max =
2
⎧ (b − t ) (0.9 bfc − t wc ) ⎫ = 106 mm
bs ,min = min ⎨ fb wc ; ⎬
⎩ 2 2 ⎭
(ii) The area of the pair of tension stiffeners between bolt rows 1 and 2 must satisfy
equations 4.17 and 4.18 in section 4.
Ast ,pair ≥
(N *
+ N r*2
r1 )
− Ltw t wc = 696 mm
2
0.9f ys,min
m1 ⎡ N r*1 N r*2 ⎤ = 2737 mm2
Ast ,pair ≥ ⎢ + ⎥
0.9f ys,min ⎢⎣ (m1 + m21 ) (m1 + m22 )⎥⎦
fys,min = 250 MPa

Ltw = 1.8sg+p = 342
p = 90 mm
Because the loading is reversing, these stiffeners must resist both tension and
compression. The compression requirement in section 5.3.1 above is for Asc, pair = 1995
mm2, however ts=16mm is required to suppress buckling in compression.
Therefore, use the compression stiffener dimensions for the tension stiffener.
(iii) Stiffener thickness must satisfy (for tension)
Ast ,pair 2737

tst ,min ≥ = = 10.5 mm
2 bs 2 x130
(iv) Confirm appropriate stiffener thickness, width
Use bs = 130 mm and ts = 16 mm
These dimensions are taken to satisfy the compression requirements. Also, use the
same weld details as specified for the compression stiffener.
5.3.3 Calculation of column web adequacy
(1) For bolt row 3 in a three bolt connection, with a horizontal stiffener between bolt
rows 1 and 2 (see section 4.5.1)
φN r 3,cw = 0.9Lt ,3 t wc f y ,cw = 639 kN
Section 2.2 Page 47

Lt,3 = 0.5sp+0.9sg = 171 mm

fy,cw = 275 MPa
N r*3 ≤ φN r 3,cw ⇒ 542 kN ≤ 639 kN ∴OK
(2) Capacity of column web to column flange connection
This must develop the design tension capacity of the column web. For a hot-rolled
column, this is always satisfied.
5.3.4 Flange backing plate design and detailing
The backing plate thickness for each bolt row has been determined from section
5.3.2(2) above.
In section 4.6, it is recommended that all the backing plates’ thickness should satisfy
the following condition;
t bp ≤ 1.5t fc = 30 mm
The determined thicknesses satisfy this condition;
Row 1: tbp1 = 10 mm
Row 2: tbp2 = 25 mm
Row 3: tbp3 = 25 mm ∴OK
As for the width of backing plate, bbp, it should not be less than the distance from the
edge of the flange to the toe of the root radius, and it should fit snugly against the root
radius, as shown in Figure 4.6.
bfc t wc
bbp ≥ − = 144 mm ∴Use bbp = 145mm
2 2
The length of the backing plate for bolt row 1 above the bolt centreline (see L’bp1 in
Figure 4.6) is given by (see section 4.6).
L’bp1 = max{0.25x191;0.9x140;2x36} = 126 mm
Similarly, L’bp3 is given by:
L’bp3 = max{0.5x191;0.9x140;2x36} = 126 mm
Because the joint is subject to reversing moment, the second backing plate is made
continuous between the horizontal stiffeners, as shown in Figure 5.1, thus requiring
only one plate between these stiffeners instead of two.
Section 2.2 Page 48

5.4 Column Panel Zone
1. Design estimate of doubler plate requirements.
This is covered in section 3.3.6.1. As a conservative estimate, doubler plate thickness

≈ twc is required; range of individual plate thickness to be between 5 mm and 16 mm.
twc = 15.1 mm; Consider 2 x 8 mm thick plates; Grade 250.
2. Calculation of panel zone design shear force.
ML MR
V *p = + − Vcol = 3167 KN
(d b − t fb )L (d b − t fb )R
ML = MR = C2MSX = 1.22MSX = 1059 kNm

C2 = 1.1x(1.0+0.54tef/db) = 1.22 (see Clause 12.9.5.2(b)(ii) of [2] for this category of
seismic-
resisting system)
VCOL = 440 kN (assumed for this design example and realistic for this size and
configuration of joint)
3. Calculation of panel zone design shear capacity, with the 2x8 mm doubler plates.
twc fyc + 2t p f yp
f*yp = = 268 MPa
twc + 2t p
fyc = 275 MPa
fyp = 260 MPa (from Part 2 of [1])
φVc = 0.9x0.6x268x903x31.1x1.0x1.01x10-3
=4105 kN
dc = 903 mm (from [17])
η = 1.0
φVc >> V*p ; look at reducing to 1 doubler plate
4. Calculation of panel zone design shear capacity with 1x8 mm thick doubler plate.
f*yp = 278 MPa

fyp = 280 MPa (from Table C2.2.1 of Part 2 [2])
φVc = 0.9x0.6x278x903x23.1x1.0x 1.02x10-3
= 3194 kN
φVc > V*p - Accept 1x8 mm thick doubler plate
5. Depth and width of doubler plate.
dp = 547.6 – 1 ≈ 546 mm
(dp is the vertical dimension)
bp = dc-2tfc-2r1c = 824 mm
(bp is the horizontal dimension)
Section 2.2 Page 49
6. Weld between the horizontal stiffener and column web on the side into which the
doubler plate is fitted.
⎛ v * ⎞
dwicp ≥ ⎜⎜ w + 3 ⎟⎟ = 5.5 mm
⎝ 0.48fuw ⎠
vw* = 570 N/mm (from section 5.3.1(7))

Use dwicp = 6 mm for this weld.
The stiffener thickness is 16 mm.
7. Welds between the top and bottom of the doubler plate and the column/horizontal
stiffeners.
These are sized in accordance with section 3.3.6.2(3).
v* w,dp, t and b = 0.9x0.6x280x8/103 = 1.30 kN/mm length of weld
Use a 8mm leg length weld, E48XX, φvw = 1.30 kN/mm
8. Welds between the sides of the doubler plate and the column flange/web junction.
In accordance with section 3.3.6.2(4), make these a square butt weld.
9. Doubler plate slenderness requirements
From section 3.3.6.2(5);
dp, max ⎛⎜ f yp ⎞⎟ 824 ⎛ 280 ⎞

= ⎜ ⎟ = 109 > 82
t p ⎜ 250 ⎟ 8 ⎜⎝ 250 ⎟⎠
⎝ ⎠
Hence a central plug weld is required.
10. Plug weld size
From Table 3.3, plug weld diameter = 25mm
11. Overall panel zone slenderness check
From section 3.3.6.2(7);
⎛ dc − 2t fc ⎞⎛⎜ fyp * ⎞⎟ ⎛ 863 ⎞⎛ 278 ⎞

⎜ ⎟ =⎜ ⎟⎜ ⎟ = 39 < 82
⎜ t wc + k 1 t p ⎟⎜ 250 ⎟ ⎝ 23.1 ⎠⎜ 250 ⎟
⎝ ⎠⎝ ⎠ ⎝ ⎠
Hence the overall slenderness is OK.
12. Bolt bearing capacity into the column flange
Section 2.2 Page 50

From section 3.3.7, bolt bearing capacity has been checked from [1] for a 25 mm thick
Grade 250 plate. tfc = 20.2 mm Grade S275, so the bolt bearing capacity needs
checking.
From Table 9.1.3(b) of [12], φVb = 456 kN for a 10 mm thick plate, compared with φVfn =
313 kN for the M36 bolt. The column flange is 20.2 mm thick.
φVb,cf ≥ φVfn hence bolt bearing capacity is OK.
5.5 Check on Column Net Cross-Section Capacity under the

Combined Action
As determined in section 5.3.2 (see calculations under Note on NS), the loss of column
cross section area due to the bolt holes is less than the limit at which it must be
included in design, thus An=Ag for axial load capacity and Ze does not need to be
reduced for moment capacity determination.
Therefore no specific check on column net cross-section capacity under combined

actions at the connection is required.
The MEP design example is completed. See connection details in Figure 5.1.
The performance of this connection under monotonic and cyclic loading has been
determined by advanced finite element analysis and is very satisfactory. See details in
[11].
Section 2.2 Page 51

References
1. Hyland, C. et. al.; “Structural steelwork connections guide”, HERA Report R4-100.
HERA, Manukau City, New Zealand. 2003.
Part 1: Design Procedures
Part 2: Connection Tables
(This report has been re-issued in 2008 as a Steel Construction New Zealand
publication)
2. NZS 3404:1997; “Steel structures standard and commentary”. Wellington, New

Zealand: Standards New Zealand, 1997, incorporating Amendment No 1, 2001 and
Amendment No. 2, 2007.
3. Syam, AA and Chapman, BG; “Design of Structural Steel Hollow Section

Connections; Volume 1 : Design Models”; Australian Steel Institute, Sydney,
Australia, 1996.
4. “Design and Construction Bulletin (DCB)”, HERA, Manukau City, New Zealand.
Issue number as specified in the test.
5. AS 4100: 1998, “Steel Structures”; Standards Australia, Sydney, Australia. 1998.
6. Kitipornchai, Al-Bermani and Murray; Eccentrically Connected Cleat Plates in

Compression, Journal of Structural Engineering, ASCE, Vol. 119, No 3, 1993, pp
767-781
7. Steel Construction; Australian Steel Institute; Issue as specified in test
8. Clifton, GC; “Structural Steelwork Limit State Design Guides Volume 1”, (HERA
Report R4-80). HERA, Manukau City, New Zealand. 1994
9. “Joints in steel construction: moment connections” (SCI Publication P207). Steel

Construction Institute, Ascot, U.K. 1995.
10. Mago, N. & Clifton, C. “Finite element analysis of moment end plate connections:
revision 2 “, HERA Report R4-117. HERA, Manukau City, New Zealand. 2003
11. Mago, N.; “Verification of revised MEP procedure”, HERA Report R4-120. HERA,
Manukau City, New Zealand. 2003.
12. “Design Capacity Tables for Structural Steel, Third Edition, Volume 1: Open
Sections”; Australian Institute of Steel Construction, Sydney, Australia, 2000.
13. AS/NZS 1554.1:2004; “Welding steel structures”. Standards New Zealand,

Wellington, New Zealand. 2004
14. AS 1101.3:2005; “Graphical Symbols for general Engineering Part 3: Welding and
Non-Destructive Examination”. Standards Australia, Sydney, Australia. 2005
15. NZS 1170.5:2004; “Structural Design Actions Part 5: Earthquake Actions”. New
Zealand Standards, Wellington, New Zealand. 2004
Section 2.2 Page 52

References
16. Feeney MJ and Clifton G C; “Seismic Design Procedures for Steel Structures”;
HERA Report R4-76, HERA, Manukau City, 1995; to be read with Clifton, GC; “Tips
on Seismic Design of Steel Structures; Notes from Presentations to Structural
Groups mid-2000”; HERA, Manukau City, New Zealand. 2000
17. “Structural Sections to BS 4:Part 1: 1993 and BSEN 10026:1999”, Corus, UK.
2005
18. Horne, M. R. and Morris L.J.; “Plastic Design of Low-Rise Frames”. Constrado,
London, England. 1981
19. Bird, GD; MemDes V2 – Program for Member Design to NZS 3404, Version 2;
BHP New Zealand Steel, Auckland, 2001
Section 2.2 Page 53

[This page is blank]
Section 2.3 SCNZ Notes on MEPS Connection Design
Section 2.3 Page 1

SCNZ 14.1 2007
ii. @@vemin19i criteria
Endplate connection moment capacity shall be greater than the applied design moment. End
plate rigidity requires limits on plate thickness and bolt offsets.
Shear capacity of the compression flange bolt group shall be greater than the applied ultimate
limit state shear.
End plate pull-out tension capacity shall be greater than the applied design moment divided by
the flange to flange lever arm.
Flange bolt group tension capacity shall be greater than the applied design moment over the
flange to flange lever arm.
Flange fillet welds shall develop the flange design tension yield capacity. This recognises that
load transfer from the bolts into the flange is from one side of the flange only and will induce
flexural loads across the weld, requiring redistribution.
Web welds shall develop the design tension yield capacity of the section web, reduced by the
ratio of the applied design moment to the section design moment capacity.
Bolt edge distance criteria shall satisfy requirements for manual flame cutting of end plates.
2. Design Actions
Design moment
Design shear
Flange bolt group tension
Flange tension yield
·Web tension yield force
3. Connection Design Strength Limits
Connection design moment capacity

Connection design shear capacity
4� Bolt Group Design Strength Limits
Bolt group tension.

Bolt group shear.
5. End Plate Strength Limits (Admt 7/04)
Plate shear
Pull-out tension
st
Bolt hole 1 bearing.
Gross transverse shear
6. Web Shear Strength Limits
Shear yield with maximum flexure.
7. Flange Weld Design Strength Limits
Flange tension.
Section 2.3 Page 2

SCNZ 14.1 2007
Web tension.
Web shear.
Ir>. Design Formulae
11. Governing Criteria
V* :$$Vean Shear
M* :$$Mean Moment
$N 1 � $[ N2,$N3]min ift; <20 Rigidity with minimumt. andd,I
m ::5:60 Rigidity
N; ::5:$N wr \
Flange weld
N :VW :5:$N ww Web weld
b.1-s
g� 1.75d, Bolt side edge distance
d; -2(a, +p ,)� 70 ford, ::5:20 Internal bolt pitch limits
90 ford,�24
2. Design Actions
Design moment
Design shear
• M.
Nfbt = Flange bolt group tension force
d-p,
N; =0.9b,t, fyr Flange tension yield
N:VW =0.9(d-2t,)t w fyw Web tension yield
3. Connection Design Strength Limits
$Mean =$N,1d, 1 +\JI, 2$N,2d, 2 Connection design moment capacity

$Vean = l$Vb;$V;;$V gb s ;$Vwwjmin Connection design shear capacity
$Vb = nbb$bVfn Bottom bolt group shear capacity
4. End Plate Bolt Row Design Capacities
a) General
$N rx = $[ N 1 ,$N 2 ,$N3,$N v ]min Bolt row force
$s fyJerxtl
$N =--'--- Mode 1: 4 Plastic Hinges in T-Stub
m
1
0.5$ 5 fy;l erxtl +n2$ bN tf

$N 2 =---'------- Mode2: 2 Plastic Hinges in T-Stub
m+n
$N3 =2$bN tf Mode 3: Bolt only mode
, $N v =0.6$ 50.5 fy;21 erxt ; Moment interaction pull-out shear
m=m1 Rows adjacent to web
Section 2.3 Page 3

SCNZ 14.1 2007
n=[e,1.25m1 ] min Effective edge distance adjacent to web

(b; -sg)
e= --- Edge distance
/ e1 = 2nm1 Circular yielding pattern 5

le2 = 4m1 +1.25e Side yielding pattern
/e3=o:.m1 Side yielding near flange pattern
8.13 + 4. 911.1 -3.4 11. 2 -16.711.� + 4.6611.� - 6.811.� 2 +8.7511.� -1.211.1

� :
0:.=[-1.2311.111.2 +8.3211.111.2, ]
2n
min
Stiffened end plate factor where 11.1 � 0. 75 and 11. 2 � 0.45
1.25 +39.3311. 1 -3.5811.2 -55.9411.� + 40.5411.� -55.3411. 1 11. 2 + 21.0511.� -33.0011.1
o:.=[+ 2.7911.111.� + 44.0611.�11. 2 , ]
2n
min
Stiffened end plate factor where 11.1 � 0.75 and 11.2 < 0.45
Edge distance ratios

6
A1 = _!!!j__
m1 +e
Sg tw
m 1 =----0.BtWW Bolt distance from web weld
2. 2 ,,
m 2 = Pr - t, - 0.8twr Bolt distance from flange weld
b) Effective T-Stub Length: Top Row of Bolts (Admt 7104)
One row, MEPS - F4

I er1 = l(o.5(/e2 +Ie3) Ie2 )max , Ie1 ]min for s g > 0.7b, or t, < 0.Bt;
Ier1 = l(!e2 •Ie3)max ' /e1J min otherwise
Two rows, MEPS - F6 or MEPS - F8

I er1 = l(o.5(1e2 +Ie3) /e2 )max ,I e1, (0.5/e2 ,0.5/e Jmax + 0.5s P J n for Sg > 0. 7b, or t, < 0.8t;
mi
ler1 = l(te2, fe3)max,/e 1 ,(0.S/e2,(te3 -0.5/e2))max +0.5sP1m in otherwise
c) Effective T-Stub Length: Second Row of Bolts
Second bolt row effective T-stub length
5. End Plate Design Transverse Strength Limits (Admt 7/04)
$V; = [$Vb;;$Vtti;$Vgsi ] min Plate shear
$Vb; = n bb $ 5 3.2fu;d r t; Bolt hole 1 bearing

st
$Vgsi =2$ 5 0.5 fy;d;f; Gross transverse shear yield
6. Web Design Strength Limits
Web shear yield with max. moment: HR
5 SCI, "Joints in Steel Construction: Moment Connections", P207/95, Steel Construction Institute, UK, 1995, p.19
6 ibid, pp.. 23, 139
Section 2.3 Page 4

SCNZ 14.1 2007
Web shear yield with max. moment: Welded
ii. FHange Follett WeH«li @iesfi!9Jliil Stitrellilgftllil R.filii'i'ints
Flange weld tension (Admt 7104)
SP fillet weld
8. Web Fillet Weld Design Strength Limits
�Vww = �Nww Weld Shear
�Nww = 2� w 0.6fuw 7i_ (d - 2tt) Web weld tension
� w = 0.8 SP fillet weld
9. Definitions of Terms
at Plate edge end distance to flange

dh = dt + 2 for d t � 24 Bolt hole diameter
dh = dt + 3 for dt > 24
d,1=d-Pt -D.5tt Bolt row 1 lever arm
d,2 = d,1-Sp Bolt row 2 lever arm
Pt Inside bolt to top flange distance
Bolt row force distribution triangular limit7
dt t
\JI , 2 = 1 . 0 for t; < - f-u Bolt row force plastic distribution
1.9 fy;
7 SCI, "Joints in Steel Construction: Moment Connections", P207/95, Steel Construction Institute, UK, 1995, p.25
Section 2.3 Page 5

t-- General Note: Refer to Tables & Detailing Constants in SCNZ Report 14.2 : 2007 "Structural Steelworl1 Connections Guide", for plat.e, weld and bolt specifications not otherwise shown on this drawing
0
0
....'<t
N
.-
if
N
z
0
(f)
'
- -
�=====E"===== --~~~k:t
' •~~~~~-- �
Plan
Ol
C
-�
Section 2.3 Page 6

"""
1:1
t�
or LL
I
CJ)
CL
w
2
-cl I, bto
�
+II+
�
or�
bi
i
Elevation Section
Structural Steelwork Steel Construction • STEELCON5mUCTION
NEW ZEALAND
LL 2/08 I SCNZ - 14: 2007 RE-ISSUE Connections Guide SCNZ-14: 2007 New Zealand Inc.
O I I
9/03 R4 - 1 oo : 2003
No. I Date I Details MEPS-F4 PO Box 76403, Manukau City, New Zealand
ph : +64-9-263-5635 fax : +64-9-263-5638 DRAWN JB Drawing No. Rev.
Revisions Moment End Plate Splice Flush email : info@scnz.org SCALE NTS ENG-MEPS-F4
Section 2.4 Civil 713 Volume 2 Notes on
Moment Resisting Column Endplate
Connections
17-19 Gladding Place
Manukau City
P O Box 76 134
Manukau City, New Zealand
Phone: +64-9-262 2885
Fax: +64-9-262 2856
Email: structural@hera.org.nz
No. 56 Extracts for CIVIL 713 notes JULY 2009 from JUNE 2000 Issue
The author(s) of each article in this publication are noted at the Several articles in this issue have been subjected to review
beginning of the article. and revision. The effort and input of the reviewers is greatly
appreciated.
Introduction
This covers just the notes on design concepts for In This Issue Section 2.3
Moment Resisting Column Baseplate Page
Connections in steel frames. It is applicable to both Page
seismic resisting systems and non seismic resisting
systems. A more in--depth set of these notes is given in Design Concepts for Moment- 2
Civil 714 Multi-storey Building Systems Resisting Column Baseplate
Connections in Steel Frame
Seismic-Resisting Systems
References 11
Appendix: Calculating the Tension 12
Capacity of Bolt/Plate Combinations
Section 2.4 Page 1 No. 56, June 2000

Design Concepts for Moment-
Resisting Column Baseplate
Connections in a Seismic-
Resisting System
This article has been written by G Charles Clifton,
HERA Structural Engineer.
Introduction and Scope of Article Performance Requirements on Connection

This article presents design concepts for moment-
resisting column baseplate connections of the Design objectives
type shown in Fig. 56.4. That figure shows a These are as follows:
connection with an unstiffened baseplate and
such an option should first be considered, within (1) To resist the design actions generated within
the recommended limitations on baseplate the column, including capacity design derived
thickness and grade given herein to suppress design actions when appropriate, and to
brittle fracture of the baseplate. These are given transfer these to a reinforced concrete
on pages 12-14 under the heading Limits on supporting member (which may itself then be
baseplate thickness and grade to suppress brittle connected to a steel pile).
fracture. They will restrict application of this type
of column base connection, especially where the (2) To allow for inelastic action to develop in the
column is subject to axial tension and moment. column base adjacent to the connection
where required for columns in category 1,
The design concepts given herein are based 2 or 3 seismic-resisting systems.
around the detailed guidance on column
baseplate connections given in section 6 of [9] (3) To distribute the internal actions from (1) and
and make reference to that publication for the (2) into the reinforced concrete supporting
detailed requirements. Where appropriate, member, without causing significant crushing
modifications to the requirements of [9] are made, of the supporting concrete, yielding of the
eg. for consistency with NZS 3101 Concrete hold-down bolts, weld failure or endplate
Structures Standard [10]. Guidance is given on fracture. Inelastic demand in the endplate
using R4-100 [1] where possible to facilitate rapid itself is acceptable, provided that this
design; this is especially applicable in the demand is minimal. The connection must
selection of initial components. retain most of its pre-earthquake rotational
rigidity throughout a severe event.
This article covers design of these connections for
use in seismic-resisting systems. The (4) To resist horizontal slip between steel and
performance requirements are covered first, concrete under seismic-induced shear
followed by limitations to suppress brittle fracture. forces.
This is followed by determination of design
actions. Guidance is limited to design of the steel
components, bolts and to the transfer of actions
from these components into the reinforced
concrete support. Design of the reinforced
concrete support itself is not covered.
This is followed by guidance on the initial

selection of connection components, using the
provisions of R4-100 as an initial starting point,
where practicable.
Steps involved in the design procedure are then

presented, followed by guidance on carrying out
these steps.
Finally, some design and constructability issues

are presented.

Fig. 56.4
Moment-Resisting Column Baseplate Connection With Unstiffened Baseplate (Item 2a from [5])
Reasons for these objectives Finally, the connection must be able to resist
In category 1, 2 or 3 multi-storey seismic-resisting seismic-induced shear without horizontal sliding,
systems which are subject to capacity design, the as this would significantly effect inelastic demand
columns will be secondary elements and will be on adjacent column members, concentrating this
subject to capacity design derived design actions demand into columns where shear slippage did
generated by the system response. In all not occur and increasing the potential for column
category 1, 2 or 3 seismic-resisting systems, the failure there.
columns will still be subject to potential inelastic
action at their base, when connected by moment- Limits on Baseplate Thickness and Grade to
resisting connections into the foundation system. Suppress Brittle Fracture
The connection must be able to resist the internal Background
actions generated by these conditions; hence As advised in section 6.2 of [9], the baseplates in
objectives 1 and 2. unstiffened moment-resisting column baseplate
connections are very thick, by comparison with
Unless special design and detailing end plates of steel to steel connections. When
considerations such as ring-springs under the subject to earthquake loading conditions, thick
hold-down bolts or other specialist devices are baseplates may be vulnerable to brittle fracture.
used, the deformation demand on the connection An example of this is, from the 1994 Northridge
components during a severe earthquake must be Earthquake, is shown in Fig. 56.5.
limited. Localised yieldline action can be
developed in the baseplate, although this is This earthquake and the 1995 Kobe (Great
limited by the relatively thick baseplates used. Hanshin) earthquake caused a considerable
Inelastic stretching of the hold-down bolts and/or number of failures in steel connections, including
crushing of the concrete must be suppressed to many examples of brittle fracture. The damage
the extent where the rigidity of the connection has been well documented and a large amount of
itself would be significantly reduced during the research has been undertaken from 1994 to the
earthquake, as this increases inelastic demand on present day to identify causes for these brittle
the system being supported. That is the reason fractures and develop design, detailing provisions
for suppression of the first two failure modes to suppress them. Two publications which cover
given for objective 3. The reasons behind both these areas are [11, 12]. More general
suppression of the last two failure modes is given account of steel building performance in the Kobe
in the following section. Earthquake is given in DCB Issue Nos. 8, 9 and
10.

Earthquake (paper from [11]). However, there are
no reported examples, that the HERA Structural
Engineer is aware of, of baseplate brittle fracture
(as shown in Fig. 56.5) occurring in the Kobe
(Great Hanshin) earthquake. This is despite the
earthquake generating high ground accelerations,
velocities and occurring at the end of a winter's
o
night with the air temperature just below 0 C.
The example from Northridge shown in Fig. 56.5

was one of a number of baseplate failures
observed in one building. The grade of steel is
not known, but is probably equivalent to our
Grade 250, and baseplate thicknesses ranged
Fig. 56.5 from 3 to 5 inches thick (75mm to 125mm).
Brittle Fracture of Very Thick Baseplate of Braced These failures have not been generally reported
System Column, 1994 Northridge Earthquake (the photo shown in Fig. 56.5 and the above
information was received by direct request from
The first step to suppress brittle fracture under the USA) and the HERA Structural Engineer is not
severe seismic loading lies in the design aware of failures of this type reported in other
procedure for the connection, ensuring that the buildings from the Northridge earthquake.
following aspects are adequately addressed:
Almost all brittle fractures reported, eg [11, 12],
the design actions on the connections are are from components of moment-resisting beam
based on the expected response of the to column connections, involving beam flanges
overall system adjacent to the column, diaphragm transfer plates
the internal forces generated in the attached through RHS columns or a combination of fracture
members and connection components by the involving these components and migrating into the
design actions are realistically determined column within the connection region.
connector-only failure modes are suppressed-
ie, the bolts and welds are not the weak link The only significant example reported of brittle
fracture in column material away from the
These issues are addressed in the next section of connection regions occurred in box columns of
this article, covering design actions on high-rise apartment buildings in Ashiyahama, near
connections and connection components. Kobe. This is mentioned in DCB No.8, reported
on in [11] and one example from this building is
The second step to suppress brittle fracture lies in presented as the front cover picture in [6]. The
keeping the material of each connection box columns were all external to the structure and
component above the transition temperature. comprised heavy channels welded toe-to-toe.
Extensive studies on steel moment-resisting Thickness ranged from 47mm down to 16mmm,
connections have shown that some of the steel grade was SM490 (NZ Grade 350). Of the
important factors involved in this are [11, 12]; 69 fractures reported, 3 occurred in thicknesses of
32mm and below, 66 in thickness above 32mm.
the direction and magnitude of loading, The fractures show little ductility, which is
especially biaxial tension loading consistent with metallurgical studies showing the
the rate of loading and extent of elastic and steel to be near its transition temperature at the
inelastic strain demand time of the earthquake [11]. Time-history
the grade, thickness and method of analyses have shown that the axial ductility
manufacture of steel demand in the CBF columns would have
the temperature at the time of loading exceeded 1.0 in the lower storeys of some of the
the restraint to a component such as a steel buildings where the brittle fractures were most
plate. Plates which have high biaxial or prevalent.
triaxial restraint are more prone to brittle
fracture, especially under seismic overload. There are no current limits on baseplate thickness
for moment-resisting column baseplate
Damage surveys from the Northridge and Kobe connections in a seismic-resisting system. The
earthquakes indicate that moment-resisting approach taken by the HERA Structural Engineer
column baseplate connections have performed in making the recommendations on maximum
well. The author observed one example in Kobe thickness and steel grade below has been simply
of fracture of a RHS to baseplate connection due to keep these outside the range associated with
to significantly undersized welds; further any of the occurrences of brittle fracture outlined
examples are given by Toyoda, M. in Lessons above. This will have the effect of limiting the use
Learned by Failure Examples in Great Hanshin of this type of connection detail. For that reason,

this section of the article ends with brief (4) Shear studs will resist both column tension
suggestions of an alternative column base detail and compression, thus the baseplate does
which does not require thick endplates and avoids not have to transfer high compression load
the risk of brittle fracture that may be associated into the concrete.
with these.
(5) Stud spacing should be to NZS 3404 [2]
Recommended limits on baseplate thickness Clause 13.3.2.3(f).
and grade for column baseplate connections
(6) Shear stud capacity is to NZS 3404
These are as follows: Equation 13.3.2.1 incorporating φsc = 0.80.
(1) Steel grade to be 250 or 300 or equivalent; Design Actions on Connections and
see NZS 3404 Table 2.6.4.4. Connection Components
(2) Maximum baseplate thickness to be: These are determined from the analysis for non-
seismic resisting systems. For category 3 seismic
(2.1) 50 mm for columns subject to moment plus resisting systems the connections should develop
axial compression. 100% of the design section moment capacity of
the column base. For more general seismic
(2,2) 32 mm for columns subject to moment plus resisting systems refer to pages 22-25 of DCB
axial tension. Issue No. 50. Those requirements extend
the coverage of HERA Report R4-76 [6] to
The more stringent limit for columns subject to
giving design actions at the column bases.
moment plus axial tension is for two reasons,
namely:
For MRFs, design actions for fixed bases and
for pinned bases are given. A large number of
the more severe loading condition, with
MRF time-history analyses undertaken as
regard to brittle fracture potential, of axial
part of HERA's semi-rigid connection
tension and moment
development have quantified the rotation
The more severe consequences if brittle
demand expected on a column base in both
fracture does occur
instances and the provisions in that issue for
pinned bases are designed to deliver the
The above limits are recommended for both
rotation demand required from the connection.
unstiffened and stiffened baseplates. The latter
These analyses have also shown the
are only briefly covered by this article. If stiffeners
considerable influence that the choice
are used, the load path to get actions from column
of connection rigidity (fixed or pinned) has on
to stiffener to baseplate should be carefully
the MRF response and the importance of
considered, with stiffeners positioned so as to
maintaining rigidity and moment resistance
reduce baseplate thickness for both tension and
in a fixed connection.
compression actions.
When applying the provisions of section 4.2.1,
Alternative column base connection detail
design actions for MRF fixed bases, the design
An alternative detail involves placing the column
moment is taken as φMr for design of the
in a pocket within the concrete base and using
baseplate and determination of stress block into
welded, headed shear studs to enact transfer of
the concrete (eg. as shown in Fig. 56.6) and
axial load from each element of the column into
φoms Mr for "design of the fillet welds, bolts, shear
the concrete. This concept is shown as
studs." In the case of the column onto unstiffened
Connection Item 30 in HERA Report R4-58 [5].
baseplate shown in Fig. 56.6, this means the
Some brief design and detailing points regarding following:
this connection are as follows. calculation of baseplate thickness on the
compression and tension sides uses φMr
(1) Use a BPP from [1] for the endplate and
bolts. calculation of hold-down bolt size uses φ Mr for
fixed base MRF connections
(2) Design the shear studs to resist
if using fillet welds between the column
overstrength actions from the column
flanges and baseplate, these are sized to
flanges, design actions from the column
resist the overstrength tension capacity from
web.
the column flange
(3) The shear studs should be located within if using fillet welds between the column web
the cage of transverse and longitudinal and baseplate, these are sized to resist the
reinforcement in the reinforced concrete design tension capacity from the column
section. flange.
Initial Selection of Connection Components should be if access is difficult. If tack welds
Making a realistic initial selection of connection are used, they must be applied between
components is an important step towards bolt and nut only on the unloaded face of
achieving a rapid and effective connection design. the nut). Nut and hardened washer supply
The following guidance will be of assistance: is to AS/NZS 1252 [14]; see section 5.5(2),
pp. 25, 26 of DCB Issue No. 50.
(1) Starting with the details given in HERA
Report R4-100 [1] for a MEP Cat 3 (3) Use a specified 28 day compression strength
connection detail; ( )
for concrete and mortar grout f c′ , f m′ of 30
use the general dimensions given therein MPa for initial design; only go stronger if
for connection layout necessary.
use the bolt diameter, numbers for the
hold-down bolt initial estimate (4) Take advantage of the increase in concrete
make the endplate approx. 50 mm wider bearing strength possible from confinement
each side than the cleat width, bi, given of the concrete under the baseplate given by
therein NZS 3101 Clause 8.3.5.2; typically the
start with the next thickest plate up from maximum value of 2 can be used.
the cleat thickness, ti, given therein
use the weld details between column Design Procedure
flange and web and the endplate that are
given therein. An iterative approach is required; this is
described in detail in section 6.9 of [9] and
(2) The bolts used will be fabricated from summarised in Fig. 56.6.
AISI 4140 bar with an ISO cut thread to
AS 1275 [13]. A nut will typically be provided For outer columns in a MRF, two cases will need
under an end bearing plate which is cast in consideration, namely moment in conjunction with
the concrete, as shown in Item 3d of [5]. maximum compression force and moment in
(The tack welds shown therein between the conjunction with minimum compression force.
nut and bearing plate can be omitted and
Fig. 56.6
Distribution of Forces and Equations for Equilibrium Based on Compression Stress Block Centred
Under Column Compression Flange
Note to Fig. 56.6
The notation C, T, a, b, X, h are the same as those used in step 1, section 6.9 of [9].

These are given as section 4.2.1(1) and 4.2.1(2), A2
pp. 22, 23, DCB Issue No. 50. The second case = enhancement factor due to confinement,
is called "maximum tension force". However, this A1
axial force will, in practice, be either compression 16.3.1 (now in
from NZS 3101 Clause 8.3.5.2
or tension depending on the relative magnitude of 2006 edition)
the gravity compression force from (G + Qu) and The grout packing placed under the baseplate
the seismic tension force. This second case will must have a specified 28 day compression stress,
deliver the maximum moment. fm′ ≥ f c′ .
Guidance on Carrying Out The Design When using the design procedure from [9], fcb
Procedure from equation 56.12 is used instead of 0.6fcu.
Use of partial strength reduction factors in
determining design capacity directly START
Several steps in the design procedure involve
determination of internal actions/design capacities
based on both concrete and steel contributions. Choose trial base
One such case is the calculation of baseplate dimensions and bolt
thickness, which is a function of concrete design positions
bearing stress and hold-down bolt tension (see
Fig. 56.6). In such instances, strength reduction
factors appropriate to the material are applied in Choose appropriate
calculating the internal actions/design capacities concrete bearing stress
Amend sizes
directly, rather than using nominal capacities
if necessary
(based on minimum specified material properties
STEP 1
or characteristic material properties) and a global Determine extent of
strength reduction factor. This is termed use of compression stress block
partial strength reduction factors and is consistent and bolt tension forces
with the approach used in longitudinal shear
checks on composite beams to NZS 3404 [2]
Clause 13.4.10, where concrete and steel STEP 2
contribute to both the internal actions and design Determine baseplate
resistances. thickness
In moment-resisting column baseplate connection

STEP 3
design, the recommended partial strength Determine holding down
reduction factors to use are: bolt size and anchorage
for calculation of baseplate thickness, φ =

0.90 STEP 4
for concrete bearing stress in the Check shear transfer
compression stress block, φ = 0.65 to concrete
for hold-down bolt capacity, φ = 0.80
for calculation of punching shear capacity in
STEP 5
the reinforced concrete supporting member,
Size welds
φ = 0.80.
Choice of design concrete bearing stress END

The design concrete bearing stress, fcb, for use in
determining the extent of the compression stress Fig. 56.7
block is given by: Flowchart for Moment-Resisting
Column Baseplate Design
A2
f cb = φ c 0.85f c′ (56.12) Determining extent of compression stress
A1 block and bolt tension forces
The conventional method of calculating bolt
where: tension and concrete compression actions is to
use a reinforced concrete type stress block
φc = 0.65 calculation, details of which are given in step 1,
f c′ = specified 28 day concrete compression section 6.9 of [9].
stress (designer's choice)

This involves solving the quadratic equation 6.6 X tf
from [9] to obtain the depth of compression stress e= − − 0.8t w (56.16)
2 2
block, termed X. That equation is based on the
compression stress block applying from the where:
compression side of the baseplate back for a fy,p = nominal yield stress of baseplate
distance X. The procedure specified on page 92 φ = 0.9
of [9] is straightforward to apply, but involves one e = cantilever length from edge of
substitution, which is to replace the concrete compression stress block to centreline
bearing stress 0.6fcu, specified by [9], with fcb from of "compression" flange.
equation 56.12 above.
The thickness of the baseplate on the
For moment-resisting column baseplate compression side for the more general unstiffened
connections to MRF seismic-resisting systems, baseplate case is determined from step 2(a),
the design moment is high relative to the axial page 93 of [9]. That text also mentions the
load. This typically means that the depth of requirements for stiffened baseplates.
compression stress block is relatively small Thicknesses for stiffened baseplates for a range
compared with the total baseplate depth. In this of boundary conditions can also be determined
case, the compression stress block can be using Formulas for Stress and Strain [13], by
assumed to be centred under the "compression" Roark & Young. In the 5th edition [13], table 11 on
flange side of the column, as shown in Fig. 56.6. page 399 is useful. However, the solutions given
The rotation required from the baseplate to in [13] are all for elastic plate bending;
achieve this is only 1-2 milliradians and readily thicknesses so determined need to be multiplied
achievable, especially given the baseplate
thickness limits on pages 13 and 14 herein. Also 4
by = 0.82 to account for a plastic distribution
the tension forces are relatively high and will 6
typically require four bolts (two rows of two bolts) of stress, on which the provisions of NZS 3404
placed about the column "tension" flange, also as and [9] are based.
shown in Fig. 56.6. This means that the centroid
of action of the tension forces is close to the Baseplate thickness required on the tension
centroid of the column "tension" flange. side
This is given by step 2(b), section 6.9 of [9]; see
Following the same approach as given in step 1, page 94 therein. This equation applies for either
section 6.9 of [9] allows the depth, X, to be two bolts positioned outside the "tension" flange,
calculated directly as: as shown in Fig. 6.8 of [9], or four bolts placed
symmetrically about the "tension" flange, as
⎛
X =⎜
1
( ⎞ ∗
)
⎟ M + 0.5N ∗ (d c − t fc )
⎜ f cb bp (d c − t fc ) ⎟
shown in Fig. 56.6 herein.
⎝ ⎠ In the latter case, T/2 is used to determine the
(56.13) plate thickness required in tension, tp,tens. In either
case, 0.9 fy,p is used instead of pyp from [9].
where:
fcb = concrete/grout design bearing stress
For a stiffened baseplate, the bolt/plate yieldlines
from equation 56.12
from Table 2.4 of [9] can be used, with care, to
bp = width of baseplate
determine the plate thickness required or the
dc, tfc = depth, flange thickness of column
tension capacity available from a given thickness.
M ∗,N ∗ = design moment, axial force This involves use of step 1A, section 2.8 of [9].
Conservatively, no bearing of the baseplate
Baseplate thickness required on the against the concrete can be assumed on the
compression side 2M p
When calculating the thickness of baseplate tension side, thus Pr,mode1 = Pr,mode2 = (see
required on the compression side, tp,c, for an m
unstiffened baseplate in which eqn. 56.13 is used page 18 of [9] or Fig. A56.1 herein).
to calculate X, the cantilever length beyond the
column flange governs, ie: Determine holding down bolt sizes and
anchorage
The design tension force to be resisted by the
4m c
t p,c = (56.14) hold-down bolts is given by equation 56.17.
φf y,p
m c = 0.5f cb e 2 (56.15) ⎛φ
T ∗ = ⎜⎜ oms
⎞
(
⎟⎟ C − N ∗ ) (56.17)
Use half this value for a gusseted column ⎝ φ ⎠
baseplate as two way action prevails C = f cb bp X (56.18)

where: A design friction coefficient of 0.3 is
fcb = concrete/grout design bearing stress recommended by [9], thus:
from equation 56.12
φVcon = 0.3C (56.20)
X = depth of compression stress block,
either from equation 6.6 of [9] or where:
equation 56.13 herein C = internal compression force generated
N* = design axial force on column (positive within the connection; see equation
if compression) 56.18
⎛ φ oms ⎞
⎜⎜ ⎟⎟ = allowance for overstrength moment
⎝ φ ⎠ The shear resistance available from friction is
∗ ∗
action from column, where this is adequate if φVcon ≥ Vcon , where Vcon is given by
required from DCB Issue No. 50 (eg. DCB Issue No. 50.
for MRF columns from section 4.2.1)
C = internal compression force In practice, the shear capacity available through
friction will be adequate for most MRF column
This tension force is resisted evenly by the bolts. baseplate connections; it is only for EBF or CBF
connections with high tension uplift where further
Design solutions for the hold-down bolts are given mechanical means such as a shear key will be
in step 3, sections 6.9 of [9]. It is recommended required. The maximum baseplate thickness of
that anchor plates be used, with the tension force 32mm recommended on page 14 for these will
in the threaded rod which forms the hold-down limit use of this connection for EBF or CBF
bolt transferred into the plate via a nut complying system column base connections.
with AS/NZS 1252 [14].
Check welds between baseplate and column
Recommended lengths of hold-down threaded rod shaft
(bolt) and anchor plate thickness are given in The design actions for these have already been
Table 6.3 of [9]; for M36 use 750 mm length and given and suitable weld sizes can be obtained
30 mm thick anchor plates. A combined anchor from the MEP-8 tables of [1] for the appropriate
plate incorporating the four bolt group will be column size.
required for the detail shown in Fig. 56.6, using
the concept given on page 96 of [9] to size the These weld design actions are more severe than
plate. the provisions of step 5, section 6.9 of [9],
because of the seismic actions imposed on the
The hold-down bolts/anchor plate must be connection and adjacent column.
contained within the longitudinal and transverse
reinforcement of the supporting concrete member. Detailing and Constructability Issues
In addition to the anchorage plate resisting
bearing, the concrete should be checked for General
punching shear failure. There are no appropriate This article ends with some guidance/discussion
provisions in NZS 3101 [10], so the provisions on of important detailing and constructability issues.
page 96 of [9] should be used, in accordance with
the following details: Hold-down bolt sizes, layouts
Having determined the number, size and position
(1) The equation for design concrete shear of hold-down bolts required in the connection
stress, vc, incorporates a partial strength design, the bolt holes in the baseplate must
1 comply with NZS 3404 Clause 14.3.5.2.2. This
reduction factor of
1.25 allows the hole diameter to be up to 6 mm greater
(2) The equation for vc in [9] uses concrete cube than the bolt (threaded rod) diameter. Note the
strength, fcu. New Zealand practice uses requirement for a plate washer at least 4 mm thick
which has minimum dimensions (diameter or side
cylinder strength, f c′ , which must be
length = hole diameter plus 8 mm) when the hole
converted to cube strength by: diameter is more than 3 mm and up to 6 mm
fcu = 1.25 f c′ (56.19) larger than the bolt diameter.
(3) All other terms are applied as stated in [9]. The tolerances on anchor bolt location given in
NZS 3404 Fig. 15.3.1 are written for compatibility
Check adequate shear transfer to concrete with the 6 mm diameter oversize baseplate holes
This follows step 4, section 6.9 of [9]; page 97 and should be specified in the contract
therein. documents.

Prior to the fabricator drilling the baseplate holes, In 2012 the AISC published the results of
the contractor (or whomever is responsible for seismic resisting system column MEP tests
placing the hold-down bolts into the concrete) which showed that connections constructed with
should obtain a surveyor's report that reports leveling nuts and no tensioning of the HD bolts
either the hold-down bolts are within the specified after installation, but instead with the use of an
tolerances or notes their actual locations. The expanding grout under the baseplate, delivered
fabricator should then fabricate to these positions. excellent seismic rotational stiffness with no
This is covered by Clause 4.5.1 and the initial flexibility.
specification checklist item 4.2 of the HERA
Specification, Report R4-99 [16]. This has led to the adoption of USA practice
here, where leveling nuts are used under the
Various means of building in tolerance baseplates instead of the shims. The nuts are
adjustments within the cast-in hold-down bolts are then installed on top of the levelling nuts and
available, but are not covered further herein. The tightened to snug tight, or with Belleville springs
use of templates to position the heads of the installed under them.
bolts, in conjunction with the anchor plate which
also locates the base, plus accurate positioning of A non shrink grout (this can be slightly
the bolt group by survey should mean that the expanding) is then placed under the baseplate
NZS 3404 tolerances will be sufficient to allow the which completes the clamping of the column
bolts to be cast directly into the concrete. base onto the foundation system.
Anchor bolt lengths should be made say 15mm Grouting between baseplate and concrete
longer than the minimum required to allow for the A bedding space of at least 50 mm is normal.
bolts being set too low. Similarly with the This gives reasonable access for grouting the
threaded lengths at the top and bottom of the bolt sleeves (necessary to prevent corrosion)
bolts. and for thoroughly filling the space under the
baseplate. It also makes a reasonable allowance
for levelling tolerances.
Leveling of column baseplate A non-shrink or slightly expanding, pourable grout

The original design procedure presented in DCB with specified 28 day compression strength, f′m ≥ fc
No 56 recommended that leveling nuts not be ′
, should be used.
placed under the baseplate, but instead that In baseplates of size 700 mm x 700 mm or
shims be used to level the baseplate. This was larger, 50 mm diameter holes should be
specified in conjunction with tensioning the HD provided to allow trapped air to escape and also
nuts after installation and placement of the grout for inspection [9]. A hole should be provided for
under the baseplate to clamp the baseplate onto each 0.5 m2 of base area. If it is intended to
the foundation. WIthout that there was concern place grout through these holes the diameter
that the baseplate would be very flexible under should be increased to 100 mm. The hole
small amounts of rotation, which is potentially not should be located near the centre of the
good for base fixity. This is different to USA baseplate; ie. away from regions of high design
practice where leveling nuts are always used and tension or design compression.
contractors in New Zealand have advised that
they cannot dependably get the columns plumb to Details are given in the subsequent sections of
within construction tolerance without the use of this part of the notes
leveling nuts under the baseplates.

References
1. Hyland, C; Structural Steelwork 12. Earthquake Damage to Steel Moment

Connections Guide; HERA, Manukau City, Connections and Suggested Improvement;
1999, HERA Report R4-100. Japanese Society of Steel Construction,
Tokyo, Japan, 1997, JSSC Technical Report No.
2. NZS 3404:1997, Steel Structures 39.
Standard; Standard New Zealand, Wellington.
13. AS 1275:1985, Metric Screw Threads
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State Design Guides Volume 1; HERA, Australia.
Manukau City, 1997, HERA Report R4-80.
14. AS/NZS 1252:1996, High Strength Bolts
4. Thornton, W A; Designing for Cost- With Associated Nuts and Washers for Structural
Efficient Fabrication; American Institute of Steel Engineering; Standards New Zealand,
Construction, Modern Steel Construction, Vol. 32, Wellington.
No. 2, February 1992, pp. 12-20.
15. Roark, R J and Young, W C; Formulas
5. Manual of Standard Connection Details for for Stress and Strain (Fifth Edition); McGraw-
Structural Steelwork; HERA, Manukau City, 1990, Hill International, Tokyo, Japan, (1975).
HERA Report R4-58.
16. Clifton, G C; HERA Specification for
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Design Procedures for Steel Structures; HERA, of Structural Steelwork; HERA, Manukau
Manukau City, 1995, HERA Report R4-76. City, 1998, HERA Report R4-99.
7. AS/NZS 1554.1:1995, Structural Steel 17. NZS 4203:1992, General Structural

Welding Part 1: Welding of Steel Structures; Design and Design Loadings for
Standards New Zealand, Wellington. Buildings; Standards New Zealand, Wellington.
8. Design Capacity Tables for Structural

Steel, Second Edition, Volume 1: Open Sections
(including Addendum No. 1); Australian Institute
of Steel Construction, Sydney, Australia,
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9. Joints in Steel Construction, Moment

Connections; The Steel Construction Institute,
Ascot, England, 1997 (with amendments), SCI
Publication No: 207/95.
10. NZS 3101:1995 (incorporating

Amendments 1 & 2; 1997), Concrete Structures
Standard; Standards New Zealand, Wellington.
11. Selected Articles and Document Excerpts

on Northridge and Hanshin Welded Connection
Fractures; 1996

For connections subject to design loads or effects
18. AS/NZS 3679:1996, Structural Steel; incorporating earthquake, NZS 3404 [2] Clause
Standards New Zealand, Wellington. 12.9.1.1.3 requires the load path and strength
hierarchy within the connection to be such as to
19. Design Guidelines for Control of In-Service avoid inelastic demand being concentrated into
Floor Vibration in Composite Floor Systems the connectors or connection components. This
(Appendix B13 of HERA Design Guides means that mode 3 failure must be suppressed by
Volume 2); HERA, Manukau City, 1989, means of suitably strong bolts.
Appendix B13 from HERA Report R4-49.
The calculation of plastic moment capacity, Mp,
20. Murray, T M et. al.; Floor Vibration due to uses an equivalent T-stub length, Leff. This is
Human Activity; American Institute of Steel determined from Tables 2.4 to 2.6 of [9]. These
Construction, 1997, Steel Design Guide cover:
Series 11.
(1) Table 2.4 presents Leff for eleven potential
21. Allen, D E et. al.;Minimising Floor Vibration; patterns of yieldline deformation for a bolt
Applied Technology Council, Redwood City, row acting alone. These cover almost all
USA, 1999, ATC Design Guide; 1. practical locations of a two-bolt row, with one
bolt positioned each side of a column or
22. Murray, T M; Floor Vibration and the beam web and inside or outside of a beam
Electronic Office; Modern Steel flange.
Construction, August 1998, pp. 24-28.
(2) Table 2.5 then presents effective T-stub
23. Clifton, G C; Tips on Seismic Design of Steel lengths from Table 2.4 to consider for bolt
Structures; Notes from Presentations to rows acting alone in a range of endplate and
Structural Groups mid-2000; HERA, column flange situations.
Manukau City, 2000.
(3) Table 2.6 then presents Leff to consider for
bolt rows acting in combination with adjacent
bolt rows.
Examples from these tables are shown in

Fig. A56.2; these examples comprise only the
table heading and first row in each instance.
Appendix A56: Calculating the Application of these provisions is made on a row
Tension Capacity of Bolt/Plate by row basis; with the capacity of a group of bolt
Combinations rows determined as the lesser of:
(i) The sum of the individual bolt row capacities

Introduction
(ii) The capacity of the bolt rows acting in
The design procedure presented in section 2.8 of
combination.
SCI Publication No. 207/95 [9] covers
determination of design capacity for rows of
tension bolts passing through endplates or Terminology and Notation When Applying
column flanges. The procedure is part of detailed These Provisions in Accordance with
coverage of the design of bolted, moment- NZS 3404
resisting connections presented in that When using these provisions in accordance with
publication. NZS 3404:
This appendix very briefly presents coverage of φf y is used for p y

the scope of that procedure, as applied to the φN tf is used for Pt′
articles presented in this issue of the DCB. It
concludes with brief guidance on application of
the provisions to connections subject to inelastic where:
demand. φ = 0.9 for the plate
φ = 0.8 for the bolts
Scope φN tf = design tension capacity for a bolt from
The procedure is applied on a bolt row by bolt row [2] or [8]
basis, with the tension capacity determined for
each bolt row using the minimum tension capacity Also, Grade 43 steel from [9] ≈ Grade 250 - Grade
determined from modes 1 to 3 shown in 300 steel in New Zealand and Grade 50 steel ≈
Fig. A56.1. Grade 350 steel. The high strength structural

bolts used are in [9] equivalent in strength and df = diameter of bolt (termed d in [9])
size to those from AS/NZS 1252 [14]. fuf = tensile strength of bolt (termed Uf in [9])
fy,p = yield strength of endplate or column
In New Zealand practice, the potential resistance flange, as appropriate (termed Pyp for
Pr, calculated as the minimum of modes 1, 2 or 3 endplate in [9] and Pyc for column
(mode 3 for non-seismic applications only), is flange)
expressed as φN m or φN ms , as appropriate.
The modification only needs to be made when
Modification of Bolt Row Force Distribution both endplate and column flange exceed the
The method of calculating bolt/plate tension limiting thickness.
capacity given in steps 1A - 1C, section 2.8 of [9]
is based on a plastic distribution of bolt forces, as For Grade 300 columns, Grade 250 endplates
is the method given in Appendix M of NZS 3404 and Property Class 8.8 bolts, the limiting
[2]. thicknesses are 0.88df for the column flange and
0.96 df for the endplate.
The latter method is restricted in scope to use
with four or more bolts placed as symmetrically as Application of UK Provisions to Connections
practicable about the beam tension flange in two Subject to Inelastic Demand
rows, as shown in Fig. M1 of NZS 3404. Finally, in this appendix, some brief advice on
application of the UK provisions to connections
If there are only four bolts, then the force subject to inelastic demand. These will be,
distribution on each bolt can be considered equal principally, connections in category 1, 2 or 3
for design purposes, as the deformation required seismic-resisting systems connecting primary
to achieve this is well within the capability of the members to secondary members; typically beams
bolts to deliver. If there are more than four bolts to columns.
in the group (ie. more than two bolts per row) then
the effect of differential loading needs to be This advice is as follows:
considered and this is flagged in Clause M2.4.
(1) Mode 3 failure of any bolt row (see Fig.
The method given in section 2.8 of [9] is based on A56.1 on page 31 herein) must be
two bolts per row, with as many bolt rows as can suppressed.
be practicably included in the connection. When
multiple bolt rows above or below the tension (2) The tension capacity of the column
flange are used, then bolt rows adjacent to a flange/bolts must be sufficient to resist the
beam flange or tension stiffener will attract a overstrength action generated by the
higher load than bolt rows further away. For incoming beam. For a beam subject to
combinations of thick endplates and smaller bolts, moment alone, this action is given by
the bolt rows in the stiffest parts of the connection ∗ φ M
N fbt = oms s
may not have the deformation capacity to allow (db − t fb )
them to plastically deform to the extent necessary
to develop the full plastic action in the bolt rows (3) The same situation applies for the
located in the more flexible parts of the endplate/bolts when any row of bolts is
connection. In such instances, the resistance governed by mode 2 capacity (see Fig.
available from the bolt rows in the more flexible A56.1). If mode 1 governs for all bolt rows,
parts of the connection may need to be reduced. then the action can be based on the design
section capacity generated by the incoming
Full details on how to do this are given in step 1C
beam. There is no simple, general criterion
of [9].
that can be used in advance to predict which
This modification needs to be considered for mode will govern, so, especially for design by
combinations of thick endplates/column flanges hand, the overstrength action could be used
and relatively small bolts, when more than one in all instances.
row of bolts either above, or below or one row
each side of the beam tension flange are used. (4) The beam flange to endplate weld, if it is a
The limiting equations for endplate or column double-sided fillet weld, must develop the
flange thickness are given as equations 2.5 and overstrength tension capacity from the beam
2.6 from [9]. The basic equation is as follows: flange.
(5) The beam web to endplate weld, when it is a

df f uf
t limit < (A56.1) double-sided fillet weld, must develop the
1.9 f y,p design section capacity in tension from the
where: beam web.
tlimit = limiting thickness of endplate or column
flange, as appropriate

Fig. A56.1
Extracts from [9] Showing Modes of Failure and Associated Equations
Seciton 2.34 Page 14 No. 56, June 2000

CIVIL 713

SEMESTER 1 – 2020
SECTION 3 – Portal Frame Design Overview and Details
SCNZ Steel Advisor Paper GEN7001 expanded for this course,

covering:
• Introduction
• Types of portal frames
• Preliminary sizing of portal frames
• Purlins
• Frame analysis
• Rafters
• Central props
• Frame deflections
• Column bases
• Roof and wall bracing
• Durability
• Structural fire severity and fire resistance
• References
This page is blank
CIVIL 713 SECTION 3:
Portal Frame Design Overview and Details
Paper originally written by Clark Hyland and Kevin Cowie of Steel Construction New Zealand Inc as Steel Advisor
Paper GEN7001, published 2010. The University of Auckland has been given permission from SCNZ to use this in
the Civil 713 Structures and Design 4 course.
In 2019, and 2020, Charles Clifton has reviewed the paper and made a number of modifications and elaborations
to enable it to be used as the principal written notes for the section of the course on portal frame design.
Key Words
Portal frame, design tips
Introduction
In October, 2009, Steel Construction New Zealand Inc., (SCNZ) ran technical seminars throughout New Zealand,
including on “Portal frame design tips.” These proceedings outline the main messages delivered on this topic at
the seminar series and were edited by Kevin Cowie. A summary of that material was presented in the SCNZ
paper GEN7001. The material came principally from two Australian Steel Institute (Woolcock et al, 1999; Hogan
et al, 1997), and one Steel Construction Institute (Salter, 2004) publications, contextualised for New Zealand
practice in accordance with the New Zealand Steel Structures Standard NZS 3404 (SNZ, 2007). The use of
these referenced documents in particular are gratefully acknowledged.
SCNZ have then given permission to the University of Auckland to use this in their Civil 713 course notes. The
details given below comprise the SCNZ GEN7001 paper with elaborations by Charles Clifton to include material
targeted at final year undergraduate students, who will be undertaking components of a portal frame design
example as part of their course and also including guidance given to SESOC who are developing a portal frame
design guide for consulting engineers in 2020.
The notes for Civil 713 also cover aspects of seismic design, support of precast concrete wall panels and
development of load paths in more detail that is in the SCNZ paper. During the course these concepts will also
be illustrated through drawings of actual portal frame projects that will be on CANVAS.
Portal Frame Types
Pitched Roof Portal (Fabricated from UBs)
A single-span symmetrical pitched roof portal frame (Figure 1) will typically have:
• A span between 15 m and 50 m
• An eaves height between 5 and 10 m
• A roof pitch between 3o and 5° is commonly adopted. Below 3o potentially leads to issues with ponding
in heavy rain; above 10o increases the axial load in the rafter which can lead to rafter instability under
combined actions needing to be addressed.
• A frame spacing between 8 m and 12 m (the greater spacings being associated with the longer span
portal frames)
• Sometimes haunches in the rafters at the eaves and apex, although more commonly a tapered welded
three plate section would be used instead of a haunched hot rolled I section.
Disclaimer: SCNZ shall not be held liable 1:

Figure or Single-span
responsible forSymmetrical
any loss or damage
Portalresulting from the2004)
Frame (Salter, use of this document
Civil 713 lecture notes portal frame design

The original GEN7001 paper is copyright to SCNZ and this version is copyright to the University of Auckland
Section 3 Page 1
Propped Portal Fram e
Where the span of a portal frame is large (greater than approx 30 m), and there is no need to provide a clear
span, a propped portal frame (Figure 2) can reduce the rafter size and also the horizontal thrust at the base,
giving economies in both steelwork and foundation costs.
This type of frame is sometimes referred to as a "single span propped portal", but it is a two-span portal frame
with a central gravity column in terms of structural behaviour. The central column typically does not participate
in the strength and stiffness for lateral loading but needs to be designed for a bending moment from the
support into the rafter, based on application of the vertical load at an eccentricity equal to half the column
depth.
Figure 2: Propped Portal Frame (Salter, 2004)
Tapered Section or Cellular Beam Portal Fram e

In recent years portal frames have been constructed using tapered welded sections and cellular beams. Cellular
beam frames commonly have curved rafters (Figure 3), which are easily achieved using cellular beams or
welded sections. Where splices are required in the rafter (for transport), they should be carefully detailed, to
preserve the architectural features for this form of construction.
Figure 3: Tapered Section or Cellular Beam Portal Frame (Salter, 2004)
Many cellular beam portal frames in the span range of 40 m to 55 m have been constructed in the United
Kingdom; greater spans are possible. Elastic design is used because the sections used cannot develop plastic
hinges at a cross-section, which is an essential criterion for elastic design with moment redistribution.
Preliminary Design of Portal Frames.

The following information is taken from the HERA Steel Design and Construction Bulletin Issue No 33. It relates
to portal frames built from Grade 300 hot rolled or welded doubly symmetric I sections. It also relates to a
“standard” portal frame which has long run steel roofing, building insulation paper and supporting chicken wire
in the roof plane between the purlins and with minimal services in the roof. The portal frames are free spanning
between the two sides and the walls are either light weight cladding or precast concrete panels.
For these portal frames, a preliminary estimate of the I section depth for a hot rolled or welded I section, d P,
can be made from;
d P = 13L+50
where:
L = span of the portal frame, in metres
Section 3 Page 2
d P = preliminary estimate of the I section depth, in mm.
For example, for a 24m span portal frame, d P = 362mm will result from this equation.
Having obtained this preliminary member depth, then the lightest available section size from the nearest weight
designation (either above or below) to eg the ASI Design Capacity Tables should be selected. In making this
selection, cognisance needs to be taken of the structural system category for the portal frame that might be
needed in design of the frame under seismic considerations. For this stage of the design process, an estimate of
the seismic category that is likely to apply to a single bay portal frame is as follows:
1. For a portal frame with light-weight clad walls, use category 4 from NZS 3404 for locations where the zone
factor, Z, from NZS 1170.5, equals or exceeds 0.2 and don’t consider a seismic category when Z < 0.2.
2. For a portal frame with one long wall comprising full height blockwork or precast concrete panels supported
laterally by the frame, using category 2 where Z ≥ 0.4, use category 3 where 0.2 ≤ Z < 0.4 and use
category 4 where Z < 0.2.
3. For a portal frame with both long walls comprising full height blockwork or precast concrete panels
supported laterally by the frame, using category 2 where Z ≥ 0.2 and use category 3 where Z < 0.2.
For example, in the case of the 24m span portal frame mentioned above, with light weight steel walls and
located in Auckland, where Z = 0.13, no seismic category of frame and hence seismic category of member
needs to be considered. Hence the preliminary member size selected would be a 360UB45.
As stated above, this preliminary sizing guidance is applicable to “standard” portal frames which have the
following characteristics:
(i) They do not support crane loads or items of heavy plant or machinery
(ii) The have a standard gable shape, with rafter slope of at least 3° to avoid ponding in heavy rain, but with a
slope of not more than 15°
(iii) They are subject to combinations of permanent, imposed, wind and snow loads as appropriate that lie
within the common range experienced in New Zealand
(iv) The frame spacing varies between 6 to 10 metres and the height to the knee varies between 5 to 8 metres
(v) The column bases are designed as nominally pinned.
Purlins
Purlin Deflections
The following deflection limits are recommended for purlins and girts.
• Under dead load alone: Span/360
• Under live load alone: Span/180
• Under serviceability wind load alone: Span/150
Purlin Bolts
The standard bolt used to be a Property Class (PC) M12-4.6/S which comes with loose washers. However, the
current standard for high strength bolt use (AS/NZS1252 2016) now recognises the PC M12 – 8.8 bolt which can
be used in both the snug tight (8.8/S) and fully tensioned (8/8/TB) modes. It should be remembered that
washers under both the head and nut are essential. For the PC 8.8 bolt/nut assemblies, the washer under the
component being turned has to be a hardened washer in accordance with AS/NZS1252. This is because the
standard punched holes in purlins are too big for M12 bolt heads and nuts, even though the height of the hole
through lapped purlins is less because of the lapping. Such flexibility is suppressed through the full tensioning in
practice even if the connections are designed as Tension Bearing (/TB) mode.
Frame Analysis
General
NZS 3404 permits a number of types of structural analysis, consisting of first and second order elastic analysis.
First order elastic analysis assumes the frame remains elastic and that its deflections are so small that secondary
effects resulting from the deflections (second order effects) are negligible. First order analysis is generally
carried out using plane frame analysis computer programs. Despite the basic assumption of first order analysis,
second order effects are not negligible. Second order effects are essentially P-∆ effects, which arise from the
sway ∆ of the frame under lateral loading or spread of the frame under vertical loading, or P-δ effects which
Section 3 Page 3
arise from the deflections δ of individual members from the straight lines joining the members' ends. NZS3404
requires that the bending moments calculated by first order analysis be modified for second order effects using
moment amplification factors. First order elastic analysis of portal frames in accordance with NZS 3404 utilises a
simple procedure that does not account for P-δ and P-∆ effects.
The use of moment amplification factors can be avoided by using second order elastic analysis. Second-order
elastic analysis essentially involves a number of iterations of first order elastic analysis with the deflected shape
of the previous iteration being used for the second and subsequent iterations until convergence is obtained.
Second order elastic analysis programs are now widely available, and as the moments obtained do not require
amplification and are generally less conservative than amplified first order elastic moments, second order elastic
moments is recommended ahead of first order amplified elastic analysis.
It should be noted that second order analysis should only be performed for load combinations and not for
individual load cases. Second order elastic analysis is performed on load combinations and not on individual
load cases, since the second order analyses using the individual load cases cannot be superimposed. Therefore,
it is necessary to have two separate sets of output for second order elastic analysis: the first for load cases and
load case deflections (as obtained by first order elastic analysis) and the second for member forces and
reactions for load combinations (as obtained by second order elastic analysis).
Elastic Analysis
Although the use of elastic analysis with moment redistribution of portal frames at the ultimate limit state is well
established in New Zealand, it is not widely used internationally. Furthermore, there are situations where elastic
analysis is more appropriate e.g. where:
• Tapered or cellular members are used.
• Instability of the frame is a controlling factor.
• Deflections are critical to the design of the structure
Elastic Analysis w ith M om ent Redistribution

Plastic hinges may form in the members within the structure as their plastic moment capacity is reached as the
structure redistributes moments. It assumes that the members behave elastically up to the full value of the
plastic moment capacity, then plastically (without strain hardening) to allow redistribution of moments around
the frame. Members required to redistribute ultimate limit state moments are required to have sufficient flexural
torsional restraint to ensure development of the plastic section capacity of the section. Examples of this will be
shown in the course.
This method has several advantages including that it optimises the use of a single hot rolled section in a frame
leading to fabrication simplification.
Fram e Design w ith Haunches (included for background inform ation as haunches are not used in
Civil 713)
For preliminary computer analysis, selection of the rafter and column sizes is from experience or by guesswork.
The computer model should have at least two nodes near each knee joint to allow for modelling of the rafter
haunches in the final design phase. Nodes at the mid-height of each column and at quarter points of the rafter
can give useful bending moment diagrams in some cases, although this is generally unnecessary when using
modern computer packages.
Haunches don’t need to be included in the initial computer run as they do not have much effect on the frame
bending moments. However, significant reductions in deflection can be achieved later in the analysis. The portal
frame example in the project will involve use of straight lengths of I section rafter and column, without
haunches.
Once the first computer analysis is run, the limit state bending moments in the column and in the rafters should
be checked against the section capacities to check the assumed sizes.
For preliminary design, reducing the column bending moment to the underside of the haunch or reducing the
section capacity to allow for coincident axial forces can be disregarded. The calculated moment at the knee
should be checked against the column section capacity φMsx. Implicit in this check is that sufficient fly braces
can be provided to ensure that the full section capacity is achieved.
The calculated bending moments in the rafter should be similarly checked against the section moment capacity
φMsx. except in the vicinity of the knee joints if haunches are provided to cater for the peak rafter moments in
Section 3 Page 4
these areas. Some small margin in flexural capacity should be retained in order to cater for axial forces. Note
that when doubly symmetric I sections are used which are compact, they can use the alternative provisions for
combined actions member design meaning that the axial load capacity is available even when M * = φM sx . The
member sizes assumed should then be adjusted accordingly and the frame analysis re-run.
Haunch Properties (included for background information as haunches are not used in Civil 713)
Once the member sizes have been established with more confidence, it is appropriate to model the haunches.
The standard AISC haunch (AISC, 1985) is formed from the same section as the rafter. It is common to model
the haunch with two or three uniform segments of equal length although reference (Hogan et al, 1997)
indicates that there is no benefit in using more than two segments.
The depth of the haunch is calculated at the mid-point of each segment and the section properties can he
calculated accordingly. Some frame analysis programmes can calculate haunch properties automatically.
Alternatively, the properties of standard UB's which are contained in standard software libraries can be used to
model the haunch segments approximately.
A comprehensive AISC publication (Hogan et al, 1997) in 1997 investigated the design of tapered portal frame
haunches fabricated from universal section members. The publication deals with detailing the cost of fabrication,
the calculation of elastic and plastic section properties, computer modelling (including the effect of varying the
number of segments), and section and member design to AS 4100, which is the source document for much of
NZS 3404. It also reviews the testing of haunches in other literature.
Modelling Base Fixity

General
Column bases are usually considered as being nominally pinned at the ultimate limit state. This simplifies the
design. NZS 3404 Clause 4.8.3.4.1 (a) specifies that a nominally pinned base should have a rotational stiffness
of 0.1(EIx/L)column. However, recently completed full scale bare portal frame testing at UofA by Amir
Shahmohammadi has shown that the pinned bases in his tapered column frame had a rotational stiffness of
0.38(EIx/L)column. Thus we should be considering larger base rotational stiffnesses than are specified by NZS3404.
Rotational stiffnesses for uniform section column frames are lower but are typically double the nominal pinned
base NZS 3404 value for ULS design. Stiffness at the base can reduce the deflections and increase the stability
of the frame considerably but also imposes more demands into the foundations. Foundations that are designed
to resist large moments are considerably larger than those designed for axial load and shear forces only and
consequently, are much more costly (Salter et al, 2004). It is therefore important for a nominally pinned base
that it be modelled for some rotational stiffness but not too much and the recommendations below are based on
the best assessment of the current state of knowledge. Furthermore, the cladding makes a significant
contribution to the stiffness of the portal frame building.
Taking all these factors into account, the following rotational base stiffnesses are recommended for design of an
individual portal frame, where this base rotational stiffness is included in the model by the use of spring stiffness
or dummy members at the column base. The rotational stiffnesses for a pinned base are based on the baseplate
with 4 bolts for erection stability which is the recommended practice.
Ultimate Limit State
At the ultimate limit state:
• A nominally rigid base should be modeled with a spring stiffness equal to 4EIcolumn/Lcolumn for working out the
lateral deflections under earthquake loading
• A nominally rigid base should be modeled with a spring stiffness equal to 1.67EIcolumn/Lcolumn for working out
the moments in the frame column and rafter under any loading condition
• A nominally pinned base of a tapered column should be modeled with a spring stiffness equal to 0.4Icolumn/
Lcolumn for frame stability checks and for lateral deflection calculations under earthquake loading
• A nominally pinned base of a uniform cross section column should be modeled with a spring stiffness equal to
0.2EIcolumn/Lcolumn for working out the moments in the column and rafter under any loading condition and the
moment demand on the foundations
Serviceability Limit State for deflection determination, including the effects of whole building action:
• A nominally rigid base can be modeled with full fixity.
• A nominally pinned base can be modeled with a spring stiffness equal to 1.67EIcolumn/Lcolumn. for lateral load
induced deflections and 0.4EIcolumn /Lcolumn for vertical load induced deflections
•The original GEN7001 paper is copyright to SCNZ and this version is copyright to the University of Auckland
Section 3 Page 5
This takes partially into account the stiffening effect of the cladding, but will still overestimate the lateral
deflections in a fully clad building.
Rafters
N om inal Bending Capacity M bx in Rafters
Simplified Procedure
NZS 3404 uses a semi-empirical equation to relate the nominal bending capacity Mbx to the elastic buckling
moment Mo and the section strength Msx, which for Universal and Welded Beams and Columns can be taken as
Zexfy. This philosophy uses a set of semi-empirical equations to relate the member strength to the plastic
moment and the elastic flexural torsional buckling moment.
Equation 5.6.1.1(1) of NZS3404 expresses the nominal member bending capacity Mbx as
Mbx = αmα sMsx ≤ Msx
where αm is a moment modification factor to account for the non-uniform distribution of major axis bending
moment, and αs is a slenderness reduction factor which depends on Msx and the elastic buckling moment of a
simply supported beam under uniform moment Mo. The standard gives comprehensive values of αm which would
be met in practice. The conservative option of taking αm equal to unity is also permitted.
For category 2 and 3 members in seismic resisting frames, α m α s ≥ 1.0 . For category 1 members α m α s ≥ φ oms ,
reflecting the need to maintain stability under over-strength actions.
The slenderness reduction factor is expressed in Clause 5.6.1.1(3) of the standard as
  2   M 
 M  
α s = 0.6   sx  + 3 −  sx 
 M
  oa   M
   oa  
Where Moa may be taken as either (i) Mo which is the elastic buckling moment for a beam with a uniform
bending distribution and with ends fully restrained against lateral translation and twist rotation but unrestrained
against minor axis rotation; or (ii) a value determined from an accurate elastic buckling analysis.
The elastic buckling moment Mo may be determined from the accurate expression given in equation 5.6.1.1(4)
as
 π 2EI   2  

Mo = 
y  GJ +  π EI w  
 L
 e
2   L 2  
  e  
Where Le is the effective length, and Ely, GJ and EIw are the flexural bending rigidity, the torsional rigidity and
the warping rigidity respectively. Values of the section properties Iy , J and Iw are given in the ASI Design
Capacity Tables for Structural Steel (AISC, 1997). The use of Equation 5.6.1.1(4) requires the effective length
Le, and determination of this is made using clause 5.6.3.
Alternative Procedure
Clause 5.6.4 of NZS3404 allows the designer to use the results of an elastic buckling analysis. This is not used in
Civil 713 or in routine design but can be very useful for unusual loading conditions or non-uniform members
used in the design of the frame.
Effective Length and Moment Modification Factors for Bending Capacity

General
If the simplified design procedure in Clause 5.6.1 of NZS 3404 is used, then the effective length Le of the rafter
must be determined in accordance with Clause 5.6.3. The effective length depends on the spacing and stiffness
Section 3 Page 6
of the purlins and fly braces, and the degree of twist and lateral rotational restraint for a chosen segment as
follows:
• Whether the connection between the purlins and rafter is rigid, semi-rigid or pinned.
• The flexural rigidity of the purlins, in that regard NZS 3404 classifies purlins qualitatively as flexible or stiff.
No numerical yardstick is given.
• The load height in that regard NZS 3404 allows, for example, for the destabilising effect of loads applied at
or above the shear centre in a beam subjected to downward loads.
• Whether the top or bottom flange is the critical flange. For a portal frame, the compression flange is the
critical flange.
• The degree of lateral rotational restraint provided at the ends of a segment by adjoining segments.
Top Flange in Compression

Under gravity loads, the top flange is mostly in compression, except near the knees. Purlins provide lateral
restraint to the top flange, but full twist restraint to the rafter from the purlins relies on the use of either
standard sized holes in the purlins or fully tensioned Property Class 8.8 bolts. The connection using a standard
purlin cleat and two bolts, slotted or nominal, can be regarded as a partial twist restraint connection in terms
of Figure 5.4.2.2 in NZS 3404. Fortunately, the standard permits partial twist restraint at the critical flange (in
association with lateral restraint) to be classified as full restraint of the cross-section as shown in Figure
5.4.2.1(b). Therefore, for each segment between purlins when the top flange is in compression, both ends are
fully restrained (FF) and the twist restraint factor kt is 1.0.
Although gravity loads are applied through the purlins at the top flange, the load height factor kt of 1.4 in Table
5.6.3(2) in NZS 3404 does not apply because the load is not free to move sideways as the member buckles. In
other words, the load is applied at a point of lateral restraint and kt should be taken as 1.0.
The degree of lateral rotational restraint provided at each end of the segment by adjoining segments depends
on whether the adjoining segments are fully restrained laterally or not, as described in Clause 5.4.3.4 of NZS
3404. (A fully restrained segment in accordance with Clause 5.3.2 is essentially one with M b not less than Ms
which means its αmαs value is greater than unity.) The standard permits full lateral rotational end restraint or
none. No intermediate option is provided. While segments between purlins under downward loading are short
and are likely to be fully restrained laterally, full restraint in accordance with Clause 5.3.2 cannot be guaranteed.
It follows that lateral rotational restraint should strictly speaking be disregarded. There is, however, a high
degree of lateral rotational restraining which would allow kr to be taken safely as 0.85.
In summary, the effective length Le is given by ktklkrL as L e = 1.0 x1.0 x 0.85S p = 0.85S p
However, not all the purlins in the positive moment length of the rafter may be effective points of restraint, for
the reasons given in the section Using Purlins to Provide Restraint to the Critical Flange given on pages 11 and
12. Because of this, the segment length, Lzs, should be taken as the length between purlins which are
considered as effective points of restraint, so Lzs may be more than 1 multiple of Sp. The use of k ez = 0.85 is still
valid in this case.
The spacing between purlins is short in comparison with the length of the rafter (Figure 4), so the moment
modification factor αm will be close to or equal to 1.0 in the positive moment regions when every rafter is an
effective point of restraint but may be a bit larger when Lzs > Sp.
Bottom Flange in Compression

With Fly Bracing Under Uplift
Under uplift, most of the bottom flange of a portal frame rafter is in compression. In such cases, the rafter is
attached to the purlins at the tension flange level, and the compression flange of the rafter is unrestrained. In
order to achieve increased member capacity, it is customary to restrain the bottom flange of the rafter laterally
by providing fly bracing using small angle section members joining the bottom flange to the purlins.
With the bottom flange in compression, NZS 3404 classifies a fly brace restraint as a full or partial cross-sectional
restraint depending on whether the purlins are flexible or stiff. Which applies depends on the ratio of purlin depth
to rafter depth and the purlin span/depth ratio. Details are in Commentary Clause C5.4.2.2, second paragraph.
Therefore if partial cross-sectional restraint applies at one or both ends of the segment, the twist restraint factor
kt will be greater than 1.0 in accordance with Table 5.6.3(1) of NZS 3404. However, unless fly braces are closely
spaced or the rafter has an unusually high flange to web thickness ratio, kt will normally be close to 1.0.
Section 3 Page 7
A value of 1.05 can be used for one end P and the other end F or L; 1.1 for both ends P.
It may appear that there should be a useful reduction in effective length because the wind loads act at the more
favourable tension flange level. However, the benefit of this is not significant as most of the bending moment
within a segment is due to end moments, and the segment should not be likened to a simply supported beam
under uniformly distributed load applied at the tension flange level. Moreover, the method of calculating the
benefit of this bottom flange loading is in NZS 3404 Appendix H which is complex to apply. For this reason, kl
should be taken as 1.0.
For a segment between fly braces and with the bottom flange in compression, the lateral rotational restraint
provided at the ends of the segment by adjoining segments should strictly speaking be disregarded because it is
unlikely that the adjoining segments are fully restrained laterally in accordance with Clause 5.4.3.4 of NZS 3404.
There is, however, a degree of lateral rotational restraint which would allow kr to be taken as 0.85.
In summary, the effective length Le for segments between fly braces for uplift conditions is given by ktkekrL as
Le = 1.0 x 1.0 x 0.85Sf = 0.85Sf .
The moment modification factor αm for segments between fly braces will usually be greater than 1.0 and often
much greater. For segments which have a reversal of moment, the critical flange changes sign along the
segment length and the benefit of this is captured in the member moment modification factor, αm. Note that
when the moment changes sign along a segment length the way in which the critical flange is restrained will also
vary from one segment end to the other.
Without Fly Bracing under Uplift
Although some fly bracing is recommended, it is interesting to consider the rafter behaviour under uplift where
there is no fly bracing at all. In this case, the segment length is from the last purlin in the positive moment region
adjacent to one knee out along the rafter to the first purlin into the positive moment region adjacent to the
opposite knee. This is a long segment length, but a rotational restraint in plan factor of 0.85 can be used to
further reduce it. The resulting long effective length and moment modification factor αm = 1.13 (for a UDL
along this length which is appropriate) will mean that the member moment capacity is low. There are ways to
determine an increased value from the top (tension) flange restraint but this is beyond the scope of the course
and requires a rotationally stiff connection between the purlins and the top flange at each purlin to rafter
connection.
With Fly Bracing under Downward Load

The effect of the bottom flange near the columns being in compression due to gravity loads or other loading
should be considered even though most of the bottom flange of the rafter is in tension. A fly brace is
recommended near each knee and near the ridge to restrain the inside corners of the frame at kinks. A stiffener
between column flanges as indicated in Figure 4 effectively extends the bottom flange of the haunch to the
outside column flange which is restrained by girts. This effectively provides some restraint to the inside of the
knee. However, a fly brace near the knee is still recommended. With fly braces at least at the knees and the
ridge, the effective length will be 0.85 times the spacing between fly braces.
An alternative approach is to consider the rafter segment between the column and point of contraflexure if
accurately known, or nearest purlin beyond the inflection point. The inflection point is considered to be
unrestrained in determining the effective length. This approach is described in an example by Clifton,
Goodfellow and Carson (1989)
The value of the moment modification factor αm for the segment should be determined using one of the three
methods in NZS 3404, but using a specifically calculated αm in Clause 5.6.1.I(a)(iii) is likely to be most
appropriate if there is no intermediate fly brace between the knee and ridge. It is recommended that any haunch
should be ignored in determining the design bending capacity φMbx of the segment, but the applied bending
moments should be reduced by factoring the moment at any haunch section by the ratio of the elastic section
modulus of the unhaunched section to the corresponding elastic modulus of the haunched section. Alternatively if
each end of the haunch happens to be fly braced as in the design example, the haunch may be treated as a
tapered segment in accordance with clause 5.6.1.1.1 of NZS 3404.
Section 3 Page 8
Figure 4: Effective Length Factors for Bending in Rafters and Columns (Woolcock et al, 1999)
Major Axis Compression Capacity Nc

In NZS3404, the nominal member capacity Nc for buckling in plane about the major axis is required in the
combined actions rules for determining the in-plane member capacity in Clause 8.4.2.2. It is obtained from
Clause 6.3.3 as
Nc = α c k f A n f y
where An is the net rafter cross-sectional area, which is generally the gross area for portal frame members (see
Clause 6.2.2 of NZS3404). The member slenderness reduction factor αc is given in tabular form in the standard
for values of the modified slenderness ratio λnx = (L e / rx ) k f fy 250 where Le is the effective length equal to
keL based on the actual rafter length L from the centre of the column to the apex.
The second order effect check for the sway frame is undertaken using the lambda_c calculation from NZS
4.9.2.4 (or directly from the SAP analysis). Once this is done then the effective length for x axis buckling of the
rafter, ie Lex, is from knee to knee. The effective lengths for y axis buckling in compression, Ley, and for lateral
buckling in bending, Lez, are covered in these notes and also in the additional notes on Civil 713 2020 Lec 9.5
200512 Determining Restraint Actions on the Braced Bay step 7.5 (d) and (e).
The form factor kf which accounts for local plate buckling are given in the steel producers’ section handbooks.
Minor Axis Compression Capacity Ncy
The nominal member capacity Ncy for buckling about the y axis is required in the combined action rules of NZS
3404 for determining the out-of-plane capacity in Clause 8.4.4.1. It is obtained by taking the effective length Le
as the distance between purlins, since the purlins are restrained longitudinally by roof sheeting acting as a rigid
diaphragm spanning between the roof bracing nodes. The theoretical effective length of an axially loaded
member (rafter or column) with discrete lateral but not twist-rotational restraints attached to one of the flanges
may be greater than the distance between the restraints. Unfortunately, there is no simple method of
determining the effective length of such a member. In the case of a rafter restrained by purlins, some degree of
twist-rotational restraint would also exist. The combined full lateral and partial twist-rotational restraint provided
by the purlins to the outside flange should be effective in enforcing the rafter to buckle in flexure between the
purlins. The capacity Ncy is obtained using the minor axis modified slenderness ratio in clause 6.3.3 of NZS 3404.
Section 3 Page 9
(
λny = L e ry ) k f fy 250
Combined Actions for Rafters

The effect of axial tensile or compressive forces in rafters combined with bending should be included in the
design as described in clause 8.4.4.1 and 8.4.4.2 of NZS 3404. Note that in most instances the axial force in the
rafter can be classified as not significant in accordance with NZS 3404 Clause 8.1.4 so combined actions need
not be determined in the rafter.
Flexural Torsional Buckling Restraints

Fly Braces
As discussed previously, fly braces are diagonal members bracing the bottom flange of rafters back to purlins, or
the inside flange of columns back to girts to stabilise the inside flange when in compression. Fly braces can take
many forms, with the most common being a single angle each side of the bottom flange, as shown in Figure 5.
The design bracing force is determined from Clause 5.4.3 of NZS 3404, which gives criteria for the strength of
braces to prevent lateral displacement of the braced compression flange. For each intermediate brace, the
design force is 2.5% of the maximum compression force in the braced flange of the segments on each side of
the brace. In this case, a segment is the length of the member between fly braces. Sharing between multiple
intermediate braces is not permitted but each bracing force is related to the local maximum flange compression
force rather than to the maximum flange compression force in the whole rafter or column. It should be noted
that NZS 3404 permits restraints to be grouped when they are more closely spaced than is required for full
lateral support, the actual arrangement of restraints being equivalent to a set of restraints which will ensure full
lateral support.
Figure 5: Double Fly-brace (HERA Report R4-92)
Under these conditions, the capacity of single bolted fly brace angles will be close to their concentric capacity
based on minor axis (y-y) buckling. For this case, even the smallest angle, a 25x25x3, has the capacity in
compression to sustain the force calculated. However, it is not really practical to use a bolt smaller than an M12,
and a 25x25 angle is too small for an M12 bolt whose washer diameter is 24 mm. The smallest angle which can
accommodate an M12 bolt is a 40x40x3 angle. While it seems unnecessary to use fly braces on both sides of the
rafter when a small angle on one side is adequate, there is a very good reason to do so, which is explained in
section 2.5.3 of HERA Report R4-92 and will be covered in the course. The restraint forces are shown in Figure 5
and put an additional bending moment into the purlin. If either brace in compression has sufficient capacity to
restrain the bottom flange from lateral movement, then on one side the actions on the purlin are opposite to
those generated by the direct loading on the purlin meaning that the restraint force will reduce the moment
demand on that purlin. It means that the purlins don’t need to be checked for additional moment from the
restraint action.
The brace will deliver concentrated load into the purlin web and care must be taken to avoid local crippling of
the purlin web under this loading. The connection should be always as close as practicable to the TOP flange of
the purlin, which is connected to the roofing, and at least within the top quarter of the purlin depth, as shown in
Section 3 Page 10
Figure 5 and the bearing stress on the purlin web may need checking to the following equation (from
AS/NZS4600 (AS/NZS4600 2018). The relevant equation is:
φVb = 1.8fupdftp,
where:
φVb = design bearing capacity of the purlin web with washers on both sides
fup = tensile strength of the purlin steel (480 MPa, typically)
df = diameter of fastener
tp = thickness of purlin web ( = 2tp for lapped purlins where the flybrace is going through both webs)
In some cases, there may be practical or aesthetic objections to fly braces because of the presence of a ceiling
above the bottom flange of the rafter. This could occur in a supermarket for example. In this case, a wider
purlin cleat and four high strength bolts, and a web stiffener on one or both sides to prevent cross-sectional
distortion, as shown in Figure 6 could be used to brace the bottom flange. The bolt shear forces in the friction
type joint can be calculated for the combined case of purlin uplift and moment due to the lateral bracing force at
the bottom flange level. The disadvantage of this approach lies in the non-standard purlin cleats and non-
standard holing of purlins.
Figure 6: Alternative Torsional Restraint (Woolcock et al, 1999)
Purlins as Braces
Where the top flange is in compression, it was assumed previously in the rafter design section that the purlins
provided adequate restraint to the top flange. NZS 3404 permits restraints to be grouped when they are more
closely spaced than is required for full lateral support, the actual arrangement of restraints being equivalent to a
set of restraints which will ensure full lateral support.
In summary, where the top flange is in compression, it is recommended that the restraint spacing necessary to
provide the required member capacity be determined. If the required restraint spacing is much greater than the
purlin spacing, then some of the purlins can be ignored as restraints, and two or three purlins near the notional
brace point could be considered as sharing the required bracing force at that point.
Using Purlins to Provide Restraint to the Rafter Critical Flange
Refer to Figure 7 in conjunction with the text below for how the purlin and roof system provides restraint to the
rafter critical flange. In this example the braced bay has two panels on each side of the roof, thus 4 braced bay
panels in each braced bay.
This means the following:

1. For negative moment, the flybraces should be effective in tension or compression and placed on each side
of the internal portals so that the purlins don’t need to be designed for the additional moment or axial load
from their restraining role as per HERA Report R4-92 and the details in the section above and Figure 5.
2. For positive moment, the collector purlins providing axial restraint are going to ideally be on grids A, A+, B,
B+,C and the purlins which are providing this restraint are going to be those for which the rafter moment is
positive at the location of the purlin. The rafter segments for checking are between these gridlines.
3. If the purlins are fully utilised in bending moment and cannot take additional design compression load in
their restraining role, then the purlin between grids 1 and 2 for example must carry the accumulated
restraint force from the 4 spans between gridlines 2 and 6. If the gable frame is made the same as the
internal frames the gable portal doesn’t need to be included in this BUT it may need restraint back to the
rafter at points where the gable vertical wall columns are placed to resist face loading from eg wind on the
wall back into the roof plane. The gable wall column will come into the bottom of the rafter through a
downwards facing cleat which should have slotted holes to allow for the gable rafter to move vertically
relative to the top of the gable wall column but not to move horizontally.
Section 3 Page 11
4. If the purlins can take additional compression, then the restraining actions can go either way and this
effectively halves the magnitude of them going into the braced bays. This iis recommended.
5. Note that for the purlin line to be effective in providing axial restraint to the top flange, the holes must not be
slotted in the purlins and the bolts should be fully tensioned. Pulin suppliers don’t provide slotted holes as
standard details. When nominal sized holes are used the bolts and purlins can be designed for /TB mode
behaviour. The bolts should be fully tensioned in the /TB mode meaning PC 8.8 bolts between rafter and
purlin should be used
6. Once the restraint forces are dependably delivered into the braced bays, that is the “point of effective
anchorage or restraint” required by NZS 3404. The braced bay members must be include the effects of these
restraining system forces for the critical load case. See the specific additional published notes for this. If
purlins other than those aligning with the braced bay panel collector beams are used, they will induce bending
into the braced bay rafter top flange. This needs to be considered in the design check on the braced bay
inner rafter for the load case which generates the highest restraining forces.
7. To minimise twist of the braced bay rafters due to the eccentricity of line of action between the roof plane
and the plane of the braced bay, the braced bay components (collector beams, tension bracing elements,
gusset plates) should be placed as closely as practicable to the underside of the rafter top flange, MAKING
SURE that a sufficient gap exists between the top side of the braced by gusset plate and the underside of the
rafter top flange on the braced bay side. This gap should be sufficient to get a welding rod into the join
between the gusset plate and the rafter web at an angle of 45 Deg preferred and at a MAXIMUM angle of 60
deg from the vertical to the rafter web (ie 30 Deg minimum off the horizontal to the gusset).
Figure 7 Portal Frame Roof and Brace Bay Layout for Purlin Restraint (courtesy of SESOC)
Central Props
General
In large span industrial buildings, a central prop is often used to reduce the rafter span and to limit rafter and
external column sizes. An efficient central prop is a square hollow section (SHS) as central props are typically
long and can buckle about both axes. Other sections such as UB’s, UC’s, WB’s or WC’s can also be used
effectively, particularly if the lateral stiffness requirements of the portal frame are a problem. The columns can
be detailed with flexible or rigid connections to the rafter. In both cases, there is a need to determine the
effective lengths both in-plane and out-of-place in order to calculate the compression capacity under axial load
alone. In the case of a rigid top connection, there will be in-plane bending moments generated in the column,
and these moments will need to be amplified if a first order elastic analysis has been carried out. If a flexible
connection between the column and rafter is detailed, it would be prudent to check the central column for both
Section 3 Page 12
pinned and rigid top connections as there will be some in-plane moments generated with most practical flexible
connections.
Figure 8: Effective Length of Central Prop (Woolcock et al, 1999)
There can be some uncertainty about how to calculate the effective length for determining the nominal capacity
Ncx in the plane of the portal frame (see Figure 8). The uncertainty arises partly because the top of the rafter is
attached to the apex of a portal frame which can sway sideways. This is dealt with in the following sections.
Effective Lengths of Props for Axial Com pression

Top Connection Pinned
If the top of the central column is connected to the portal frame by a flexible connection such as a cleat
perpendicular to the plane of the frame, it would be reasonable to regard this connection as pinned. In this
case, the central column does not interact in flexure with the frame, but the frame must have a certain
minimum stiffness to effectively brace the top of the columns as shown in Figure 8. For a pinned base column
the minimum spring stiffness to ensure that its effective length Le is equal to and not greater than the length L
of the column is π 2EI c / L3 .
In practical frames, the side-sway stiffness of the rigid frame with its relatively stiff side columns and rafter is
usually quite sufficient to brace the top of a slender central column. Designers can readily determine the
sideways stiffness by analysing a special load case with a single horizontal load at the apex of the frame.
Top Connection Rigid

If the top connection is rigid, then there should logically be some reduction in effective length of the central
column. However, in accordance with NZS 3404, it is not possible to determine directly the effective length of
individual members in non-rectangular frames. The standard in Clause 4.9 requires a rational buckling analysis
of the whole frame to determine the frame elastic bulking load factor λc. The only practical way of determining
λc is by means of a frame analysis. These programs also convert the λc value for each load combination into
effective lengths for each member by use of Equation 4.5.
Combined Actions with First Order Elastic Analysis
If the top connection is rigid, the frame elastic buckling load factor λc for each load combination is used in
Clause 4.4.3.3.2(b) to determine the amplification factor δc which is applied to any bending moments from a
first order elastic analysis. The capacity of the central column is then checked under Clause 8.4.2.2 of NZS 3404
using an effective length factor ke of 1.0 for combined actions, and also an effective length factor calculated
from λc for axial load alone.
Section 3 Page 13
If the top and bottom connections are assumed to be pinned, there will be no moments from the frame analysis
but a nominal eccentricity in each direction is recommended. The effective length factor ke will then be 1.0 for
both combined actions and for axial load alone if the minimum spring stiffness in Section 4.6.2.1 is provided.
Combined Actions with Second Order Elastic Analysis
Ironically, if a designer has access to programs such to determine λc for amplifying first order moments, then it
is likely that the designer also has access to the second order elastic analysis option of these programs. In this
case, a designer would ideally use the second order elastic analysis as this obviates the need to amplify the
moments. The capacity of the central column is then checked as described in the previous section.
Frame Deflections
General
Portal frames are generally designed on the basis of strength first, and are checked for the serviceability
(deflection) limit state according to some arbitrary criteria. Deflection limits can govern the design of portal
frames, and it is therefore important that any deflection limits be realistic.
The selection of deflection criteria for industrial steel frames is a subjective matter. In general, standards are
not prepared to give specific recommendations, probably because deflection limits have not been adequately
researched. The Australian steel code AS4100 states that the responsibility for selecting deflection limits rests
with the designer, but still gives some recommendations. For a metal clad building without gantry cranes and
without internal partitions against external walls, the standard suggests a limit on the horizontal deflection of the
eave as column height/150 under serviceability wind loads. This limit reduces to column height/240 when the
building has masonry walls. The limits suggested in Appendix B of AS 4100 are based on the work in (Woolcock
et al, 1986).
Problem s of Excessive Deflection

The potential problems of excessive deflections in industrial buildings include:
• Damage to cladding and fixings thereby affecting the hold down capacity of fixings and water
tightness.
• Ponding of water on low pitched roofs and possible leakage because of ponding or insufficient pitch.
• Visually objectionable sag in rafters or suspended ceilings whose ceiling hangers are difficult to adjust
for sag, e.g. heavy acoustic ceilings.
• Visually objectionable sag in the ridgeline because of the deflection of the apexes of internal rafters
relative to the end wall apexes. The end wall rafters do not sag because they are supported by end
wall columns.
The results of the survey were reported in (Woolcock et al, 1986). It is interesting to note that in many answers,
there was no clear consensus of opinion among engineers. What is regarded as acceptable to one engineer is
not necessarily acceptable to another. The results of the survey were rationalised, and deflection limits were
proposed. These are summarised in Figure 9, Figure 10 and Figure 11. It is emphasised that these limits should
be used for guidance rather than as mandatory limits. Further research is required to establish deflection limits
with more confidence.
Figure 9 Notation for Deflection Limits (Woolcock et al, 1999)

Section 3 Page 14
Figure 10: Recommended Rafter Deflection Limits
(Woolcock et al, 1999)
Figure 11: Recommended Lateral Deflection Limits
(Woolcock et al, 1999)
Notes:
The wind load deflection limits apply to serviceability wind loads.
• L is the rafter span measured between column centrelines.
• Precamber or pre-set may be used to ensure that the deflected position of the rafter under dead load corresponds to
the undeflected design profile, or is within the above limits of the undeflected design proflle. Even so, pre-set may be
advisable for internal rafters to avoid visual sag in the ridge line.
• For low roof pitches, the check for ponding is really a check to ensure that the slope of the roof sheeting is nowhere
less than the minimum slope reconnnended by the manufacturer. The slope of the rafter in its deflected state can be
determined from the joint rotations output from a plane frame analysis program. The slope of the roofing should also
be checked mid-way between rafters near the eaves where purlins are more closely spaced and where the fascia
purlin may be significantly stiffer than the other purlins.
• Where ceilings are present, more stringent limits will probably be necessary.
Suppressing Deflection in Rafter Splices

When the knee connection between rafter and column is a (shop) welded connection, this must be combined
with a rafter splice. If the rafter length is very long (more than about 18 metres for each length of rafter
between knee and apes, at least one splice per length of rafter must be included to enable transport.
These rafter splices are typically bolted beam splices (BBS) or bolted welded beam splices (BWBS). Details are
given in SCNZ Steel Connect.
It is very important that these are designed for at least the minimum moment actions required for a beam splice
carrying moment from NZS 3404, this being best represented by NZS 3404 Clause 12.9.2.1.1 (b) being 30% of
the design section capacity of the member at the splice. This is required when the portal frame is designed for
earthquake to any category and can also be used as a simpler application of the general Clause 9.1.4.1 (b).
Furthermore, the bolts used in the flanges must be fully tensioned and this must be done during erection while
the rafter is straight across each side of the splice location. If this is not done then, with the accuracy of modern
fabrication layout, the splice will rotate as the flange plates and bolts go into bearing, leading to visible local
rotation at the splice location and a significant loss of functionality.
Differential Deflections
Generally, where a rafter and post frame has been used, it will be braced and will therefore be much stiffer than
the adjacent portal frames. In practice this is also true with a portal frame gable wall because it will be stiffened
by the cladding. Differential deflection between the gable frame and penultimate frame can therefore be
relatively large, and may be of particular concern if there are cranes, masonry construction, or sensitive cladding
attached to the frame.
Section 3 Page 15
Ways of reducing differential deflections include:
• Bracing in the roof between the gable frame and the adjacent frame will reduce the deflection of the
adjacent portal frame to some extent, but this is normally not quantifiable without a 3-D analysis of the
whole structure.
• A penultimate frame can be provided of greater stiffness than the other frames to reduce the differential
deflection due to eaves spread and wind loading. This is not usually a sensible option in terms of fabrication
efficiency.
• The portal frames should be pre-set carefully to ensure that all dead load deflections result in frames that
line up with the gable frame under dead load only, thus reducing to some extent the differential deflection
due to eaves spread.
W all Claddings and How they are Supported off the Fram es
There are typically two types of wall claddings; long run metal cladding and concrete panels.
Long run metal cladding is what is used on the roof and often the same profile and thickness is used. Wall
claddings don’t need to be sufficiently strong or shaped to enable roofing installers to walk on them without
damage, but they may be subject to impact loading in practice and the requirements for carrying concentrated
load introduced through an installers foot is also useful for wall impact resistance. The self weight of lightweight
wall cladding is not more than 0.2 kPa and the cladding elements span vertically between girts which are
supported back to the portal frame columns.
Concrete panels are typically 150 to 175 mm thick, weigh around 3.6 to 4.2 kPa and are supported for vertical
load off the foundation of the building, typically sitting on a ring beam either in a pocket cast into this or
anchored with pin in ducts set into the wall panels. The panels typically span full height of the wall onto a PFC
supporting beams which spans between the column knees. These PFCs are typically installed toes down for
durability reasons and connected by their webs to horizontally placed cleats, effectively being a Web Plate
connection but oriented to carry horizontal load between the supporting beam and the columns. The panels are
supported off connections at each panel corner or approx 2m centres, whichever is the least. The panels are not
connected structurally to the columns.
These panels may form the lateral load resisting system in the longitudinal direction of the building or may be
isolated laterally from the building structure. They are designed to NZS 3101 as nominally ductile systems.
For determining the seismic weight of the walls onto the frames for frame analysis, the weight of the top half of
the wall is lumped at the column knee and the weight of the bottom half of the wall taken into the foundations.
The self weight of the wall is used for determining the seismic mass of the frame. However, the forces that the
walls generate on the supporting beam at the column knee, for design of the frame to the wall and for design of
the knee beam and attachment to the columns are based on the Parts and Portions section of NZS1170.5. This
reflects the fact that the horizontal accelerations at the top of the panel will be greater than at the bottom due
to the lightly damped nature of the building system.
More details on this will be covered in the class.

Column Bases
Base Plates
Mild steel Grade 4.6 bolts are preferred because they can be adjusted by bending on site particularly if there is a
sleeve or pocket around the holding down bolt for this purpose. Mild steel bolts can also be tack welded into a
cage, whereas Grade 8.8 bolts should not be tack welded because welding can have an adverse effect on steel
properties in the vicinity of the weld. Regardless of the steel grade, it is recommended that holding down bolts
be hot dip galvanised.
Holding Dow n Bolt Design Criteria

There are many considerations in the design of holding down bolts (Trahair et al, 1998), the most important
being as follows:
• The bolts themselves should have sufficient capacity in combined tension and shear.
• The grouting or bedding under the base plate should have sufficient capacity in compression to cater
for applied compression and bending moment at the base of the column.
Section 3 Page 16
• The concrete or the grout filling the space around the bolts and sleeves should have sufficient strength
in bearing to transmit the shear force in the bolt.
• If the bolts do not have a suitable head or other anchor at the head to prevent pullout or bearing
failure under the head, the bolts must be sufficiently long or must be suitably cogged or hooked to
satisfy the anchorage requirements for plain deformed bars (as appropriate) in the concrete standard
NZS 3101 (SNZ, 2006).
• If the bolts have a suitable head or anchor, the embedment must be sufficient to prevent the bolts
pulling out a cone of concrete (cone failure).
• If there is insufficient edge distance, the bolts must be lapped or anchored with reinforcing bars in
accordance with the concrete standard.
• Account should be taken of fabrication and erection tolerances when detailing and installing holding
down bolts.
• The likelihood of corrosion must be considered carefully. Hot dip galvanizing is recommended.
• A minimum of four bolts rather than two bolts is favoured by riggers to assist in supporting columns
during erection.
Base M om ents for Foundation Design

It should be noted that, as far as the base moment (and associated forces) for foundation design is concerned,
the following applies:
• Where partial base fixity is used to reduce the moments for which frame members have to be designed
(compared to those obtained assuming pinned bases) the base moments should be taken into account in
designing the foundations. This applies for both elastic analysis and elastic analysis with redistribution of
moments.
• Where a nominal 10% base stiffness is used only in assessing effective lengths (or elastic critical load
factors) or in determining whether an unbraced frame is 'sway-sensitive' or 'non-sway' it is not necessary to
take account of the base fixity moment in foundation design.
See the extracts from HERA DCB No 56 included in these notes for design of the column baseplate connection.
Roof & Wall Bracing
General
Roof and wall bracing often consist of panels of double diagonals which are so slender as to have negligible
capacity in compression. Such members include pre-tensioned rods, slender tubes and angles. In the design of
double diagonal tension bracing, one of each pair of diagonals is assumed to act in tension as shown in Figure
12, depending on the direction of wind loading, and the other diagonal is usually ignored. In addition to tension
forces, roof bracing diagonals have to carry their own weight whether by cable action in the case of rods, or by
beam action in the case of tubes and angles.
Figure 12: Roof and Wall Bracing (Woolcock et al, 1999)
As common as tension bracing is, there is not a widely accepted method of design which accounts for tension
and self weight. This problem was investigated in References (Kitipornchai et al, 1985; Woolcock et al, 1985).
Section 3 Page 17
Tem porary Bracing
Portal frames can collapse during construction if adequate care is not taken to use permanent or temporary
bracing to withstand wind gusts. The procedure to be used varies from building to building depending on the
type and location of the permanent roof and wall bracing bays and whether the end wall frame is a braced
frame or a portal frame.
Roof Plane Bracing

Roof plane bracing is placed in the plane of the roof. The primary functions of the roof plane bracing are:
• To transmit horizontal wind forces from the gable posts to the vertical bracing in the walls.
• To provide stability during erection.
• To provide a stiff anchorage for the purlins that are used to restrain the rafters.
Rafter Bracing Forces
In addition to the longitudinal wind forces, the bracing system could also be considered as resisting
accumulated, coincidental purlin or fly brace forces. When the top flange is in compression, the purlins act as
Figure 13: Bracing Layout Options (Woolcock et al, 1999)
Section 3 Page 18
braces, whereas fly braces restrain the bottom flange when it is in compression. However, it is unclear whether
the bracing forces should be accumulated. Purlins and fly braces together could be considered as providing a
rotational restraint system in accordance with Clause 5.4.3.2 of NZS 3404. In this case, it would not be
necessary to treat the compression flanges of rafters as parallel restrained members in accordance with Clause
5.4.3.3, and therefore it would not be necessary to accumulate the forces. On the other hand, purlins and fly
braces could be considered as providing restraint against lateral deflection of the compression flange (Clause
5.4.3.1) and in this case the bracing forces would be accumulated.
It is interesting to compare roof trusses as far as accumulation of bracing forces is concerned. The bottom
compression chord of a series of large span roof trusses under net uplift is usually braced back to the end
bracing bays by a system of struts or ties. In this case, the bracing forces should be accumulated and then
combined with forces due to longitudinal wind. When the top chord is in compression, it is usually regarded as
being braced by purlins back to the end bracing bays. Logically, the top chord bracing forces should also be
accumulated, but as the compression in the top chord is generally due to gravity loads, there are no other
longitudinal forces in combination and so the loads on the end bracing bays are not likely to be critical.
It could be similarly argued that the top or bottom flange bracing forces of UB or WB rafters, whichever flange is
in compression, should also be accumulated. However, even if the lateral restraint argument (as opposed to the
rotational restraint argument) is accepted, the accumulated bracing forces are usually a small part of the total
longitudinal force for portal frame buildings. It is therefore considered reasonable for UB or WB rafters, to ignore
accumulated bracing actions in the design of the roof and wall bracing bays.
A summary of the advantages and disadvantages of various options for bracing layouts is as follows as shown in
Figure 13.
Option I: Two End Bays Braced
This is the simplest and most direct option. Intermediate eaves and ridge struts are sometimes used as shown
dashed. However, purlins are usually sufficient to brace internal rafters so that no intermediate struts are
required.
Longitudinal wind loads, as a combination of pressure on the windward wall, suction on the leeward wall and
friction, could be shared between braced bays if purlins have the capacity to transfer some compression load
from one end to the other. However, it is recommended that the bracing at each end be designed to resist loads
from external pressure and internal suction on the adjacent end wall (plus half of the frictional drag forces if
applicable). This keeps the purlin design simple as purlins can then be designed without considering combined
actions. Diagonals are crossed which means that CHS sections, which are efficient as long ties under self weight,
cannot easily be used. This option also excludes the use of the top flange as a bracing plane with angle
diagonals crossed back to back unless higher purlin cleats are used. End bay bracing can have detailing
difficulties at the end wall rafter.
Option II: Double Diagonal Bracing Over Two Bays at Each End
• Diagonals intersect at rafters and therefore tubes can be used as diagonals without difficulty if they are
not crossed.
• The number of diagonals is the same as for Option I but more struts are required.
Option III: Second Bay from Each End Braced
• This option can overcome any detailing difficulties associated with end bay bracing but extra struts are
required to transfer the end wall wind loads to the braced bays unless the purlins can act as struts.
Option IV: One Bay Braced
• Struts in the unbraced bays are required to transfer end wall wind loads to the braced bay which is
expensive unless the purlins can act as struts.
Option V: Single Diagonal Tension Bracing at Each End
• Unstable during erection.
• The windward braced bay takes all of the longitudinal wind loads.
• Purlins are usually sufficient to brace internal rafters as in Option I. Leeward end wall forces are
transmitted to the active braced bay at the windward end by purlins in tension.
• Tubes can be used for diagonals without difficulty as they are not crossed.
• Single diagonal rods with turnbuckles should not be used as there is nothing to tension against.
Section 3 Page 19
• Temporary diagonals may be necessary to create a double diagonal bracing system for erection
purposes in which case there is little advantage in a single diagonal system.
Bracing using Circular Hollow Sections

In the United Kingdom circular hollow section bracing members are generally used in the roof and are designed
to resist both tension and compression. Many arrangements are possible, depending on the spacing of the
frames and the positions of the gable posts (Salter et al, 2004).
Tension Rod Bracing
Rods cater for the lower end of the range of tensile forces, and are very common in light industrial buildings.
Rods differ from tubes and angles in that they must be pre-tensioned to reduce their self weight sag. However,
there are certain aspects of rod pre-tensioning which are not widely understood. The aspects which need to be
considered are as follows:
• The minimum level of pretension force needed to reduce the sag sufficiently to avoid undue axial slack
in the rod.
• The level of pretension used in practice. The effect of pretension on the tensile capacity.
• The effect of pretension on the end connections, and on the adjacent struts in the roof bracing system,
when wind loads are applied.
Pre-tension actions should be 10% to 15% of the yield capacity. While these levels of pretension may be
adequate, it is not practical to measure or control the pre-stress level in practice. To answer the questions above
properly, it is necessary to examine the behaviour of pre-tensioned rods in some detail.
Long rods behave like cables whose self-weight is carried by tension alone; the tension being inversely
proportional to the sag. For small sags in roof bracing situations, the tensile stress fat versus sag yc relationship
has been shown (Clifton 1994) to be independent of the rod diameter and is given by
 L2 
fat = 9.62 x 10 −6   MPa
y 
 c 
in which L is the length of the rod and both y, and L are in mm. This relationship is presented graphically in
Figure 14. Using this equation, it can be demonstrated that as a rod is tensioned, very little force is required to
reduce the sag until the sag gets to about span/100. The rod then begins to stiffen suddenly and behave as a
straight tension member. This is shown graphically in Figure 15. Therefore, the maximum sag of a rod to avoid
undue axial slack should be about span/100. Surprisingly, a stress of only 20 MPa is required to reduce the sag
of a 20 metre cable to the L/100 deflection. However, typical stress levels in practice could be much higher.
In experiments at the University of Queensland (Woolcock et al, 1985), six different laboratory technicians were
asked to tighten rods ranging in diameter from 12 mm to 24 mm with spans up to 13 metres long. They were
told to tighten the nuts as if they were working on site. Once tightened at one end, the force in the rod was
measured with a calibrated proving ring connected to the other end. The experiments revealed that the average
level of pretension force was well in excess of the value of 10% to 15% suggested in (Gorenc et al, 1996). In
fact, it was found that 16 mm diameter rods were tensioned close to their design capacity, while 20 mm rods
were tensioned to between 40% and 55% of their design tensile capacity. Because of these unexpectedly high
pretension forces, excessive sag is not a problem, even for a 20 metre span.
The presence of pre-tension does not affect the ultimate tensile capacity of the rod itself. However, there are a
few other factors that need to be considered in the design of roof bracing rods. In some cases of over-
tensioning, the active tension diagonal may yield under the serviceability wind load, although yielding will relieve
the pretension in the system to some extent. Fortunately, the fracture capacity of the threaded section typically
exceeds the yield capacity of the rod itself. This means that the main body of the rod will generally yield before
failure of the turnbuckle section. Because of the pretension, the rod connections should be designed so that
their ultimate or fracture capacity is equal to or greater than the ultimate or fracture capacity of the rods. This is
particularly important because oversized rods are often used. For example, a 20 mm diameter rod may be used
because of its robustness where only a 16 mm diameter rod is required. This philosophy for the end connection
design of rods is covered in Clause 9.1.4(b)(iii) of NZS3404.
Section 3 Page 20
Figure 14: Effect of Axial Stress on Cable and Rod Deflections (Woolcock et al, 1999)
Pre-tensioning could also result in overloading of the struts in the roof bracing system, especially if rods larger
than that required are used. A check should therefore be made in the design of the struts to cater for forces in
the diagonals due to combined pretension and wind load as shown in the design example.
Figure 15: Effective Axial Stiffness of Cables and Rods (Woolcock et al, 1999)
Tubes and Angles in Tension

In contrast to rods, tubes and angles are not easily pre-tensioned and must be sized as beams to limit self
weight sag. The uncertainties for designers, as far as tube and angle section members are concerned, are firstly
the effect of self weight bending on tensile capacity, and secondly deflection limits. Some engineers combine
self weight bending actions with axial tensile actions, while many engineers intuitively ignore the bending
actions.
It can be shown theoretically (Woolcock, 1985) that self weight bending has a marginal effect on the ultimate
fracture capacity of a tube or angle. This is because the sag and self-weight bending moments reduce as the
tension increases. It can therefore be concluded that self-weight bending actions need not be considered in
combination with axial tension.
As proposed for rods, a maximum sag of span/100 is suggested to avoid undue slack. However, it is advisable to
limit deflections to span/150 to avoid lack of fit without propping during erection, and for aesthetic reasons.
Note that even with a span/150 deflection, there is occasionally concern expressed during construction as the
sag can be quite evident if one sights along the member. The sag is not generally obvious from floor level.
Of course, the designer has the option of suspending the diagonals from the purlins, but very flexible diagonals
(other than rods) can be difficult to erect before the purlins are in place because of lack of fit. If the purlins are
erected first, the stability of the portal frames without bracing may be inadequate and lifting the diagonals into
Section 3 Page 21
place will be more difficult because of obstruction from the purlins. Furthermore, the extra labour necessary to
drill and suspend may cost more than the material saved. The effect of purlin uplift loads on the capacity of
diagonals should also be taken into account. With all these factors considered, suspending very flexible
diagonals from purlins is not recommended.
Side-w all Bracing

General
The primary functions of vertical bracing in the side walls of buildings are:
• To transmit the horizontal loads, acting on the end of the building, to the ground.
• To provide a rigid framework to which side rails may be attached so that they can in turn provide stability to
the columns.
To provide temporary stability during erection, the bracing system will usually take the form of:
• Circular hollow sections in a V pattern.
• Tension only cross-braced rods.
• Circular hollow sections in a K pattern.
• Crossed flats (within a cavity wall).
• Crossed hot rolled angles.
The bracing may be located at:

• One or both ends of the building, depending on the length of the structure.
• At the centre of the building (but this is rarely done due to the need to begin erection from one braced bay
at, or close to, the end of the building).
• In each portion between expansion joints (where these occur).
Where the side wall bracing is not in the same bay as the plan bracing in the roof, an eaves strut is required to
transmit the forces from the plan bracing into the wall bracing.
Side-wall Bracing Using Circular Hollow Sections
Circular hollow sections are very efficient in compression, which eliminates the need for cross bracing. Where
the height to eaves is approximately equal to the spacing of the frames, a single bracing member at each
location is economic. Where the eaves height is large in relation to the frame spacing, a K brace may be used.
An eaves strut may be required in the end bays, depending on the configuration of the plan bracing. In all
cases, it is good practice to provide an eaves tie along the length of the building.
Durability.
Durability for the steelwork in portal frames is covered by SNZ TS 3404 (SNZ_TS3404 2018) which is a simplified
procedure developed from HERA Report R4-133 (Clifton and El Sarraf 2011). The inside environment is typically
less corrosive than the outside environment unless the internal operating processes used in the building cause
dampness or other high levels of internal corrosion. See Table 2 from SNZ TS 3404 for determining the
appropriate surface specific corrosivity category to use.
For concrete panels, use the provisions of NZS 3101 Chapter 3 to determine the value for the wall which will
typically be governed by the external environmental conditions.
Structural Fire Severity and Fire Resistance.

The structural fire severity generated within portal frames is linked to the fire load energy density and the
expected response of the structure to severe fires, in the manner described in section 4.3.2.2 of HERA Report
R4-91 (Clifton and Forrest 1996). In summary this gives the following:
1. For low fire loads (Fire Load Energy Density (FLED) up to 500 MJ/m2 floor area) the structural fire severity
is up to 30 minutes. Structural fire severity is a measure of the time of exposure to the ISO Standard Fire
and ≡ FRR as specified by Fire Engineers; ie use FRR = 30.
2. For moderate fire loads, 500 < FLED ≤ 1000, FRR is typically around 45 minutes but can get to 60 minutes
and so is typically taken as 60 minutes.
3. For high fire loads, 1000 < FLED ≤ 1500, FRR = 60 can also be used.
4. For fire loads 1500 < FLED, use FRR 90 or FRR 120. The latter would be for very high fire load exposures.
Section 3 Page 22
The two most applicable values are FRR 30 for eg gymnasia, sports halls, manufacturing with low fire loads,
storage non combustible material , steel fabrication plants and FRR60 for most other applications.
If the FLED ≥ around 500 MJ/m2 floor area, the rafter will sag and this will increase as the fire load increases. It
will open up roof venting which significantly reduces the structural fire severity. This raises the question of the
stability of the external walls.
For stability during and after the fire the walls are not allowed to be cantilevered without design support from
the steel frame. There is explicit requirements on this in NZS 3101 Clause 4.8. A cantilevered wall without top
support from the structure will collapse outwards in a severe enough fire, whereas with top support or support
off the columns it would collapse inwards if it were allowed to collapse at all. Wall collapse is avoided by
determining whether the collapsed wall condition can be allowed, based on the separation from the boundary
(to the wall + wall height) being sufficient to allow 100% openings. If it is, then the wall can be allowed to be
pulled inwards by the collapsing rafter and the column need not be fire rated. If it isn’t then the column and the
wall top support must be able to carry a load of 0.5kPa on the wall at a reduced strength associated with a
temperature of 500 Deg C for the column base and 680 Deg C for all other components.
The time equivalent ( = FRR) for determining the FRR of the fire resisting external walls (and any internal walls
required to have an FRR) can be determined using Paragraph 2.4.4 of C/VM2 with Ah/Af = 0.2, kb = 0.04,
km = 1.0 and Av as calculated for the vertical openings. Roller shutter doors are taken as shut for this
calculation.
References
AISC, Design Capacity Tables for Structural Sections-Volume 1: Open Sections, 2nd edition & Addendum No.1,
Australian Institute of Steel Construction, Sydney, 1997
AISC, Standardized Structural Connections, 3rd Edition, Australian Institute of Steel Construction, Sydney, 1985
Bradford, M.A., Lateral Stability of Tapered Beam-Columns with Elastic Restraints. The Structural Engineer,
66(22), 376-384, 1988
Clifton, G. C., Goodfellow, B., Carson, W., Notes Prepared for a Seminar on Economical Single Storey Design
and Construction, HERA Report R4-52, New Zealand Heavy Engineering Research Association, Manukau City,
1989
Dux, P.F., Kitipornchai, S., Buckling of Braced Beams, Steel Construction, Journal of the Australian Institute of
Steel Construction, AISC, 20(1), 1-20, Sydney, 1986
Gorenc, B.E., Tinyou, R., Syam, A.A., Steel Designers Handbook. NSW University Press, Sydney, 1996
Hogan, T.J., Syam, A.A., Design of Tapered Haunched Universal Section Members in Portal Frame Rafters, Steel
Construction, Journal of the Australian Institute of Steel Construction, AISC, 31(3), 1-28, Sydney, 1997
Kitipornchai, S., Woolcock, S.T., Design of Diagonal Roof Bracing Rods and Tubes. Journal of Structural
Engineering, ASCE, 115(5), 1068-1094, 1985
Salter, P.R., Malik, A.S., King, C.M. Design of Single-span Steel Portal Frames to BS 5950-1:2000, Steel
Construction Institute, Silwood Park, 2004
Section 3 Page 23
SNZ, Concrete Structures Standard, NZS 3101:2006, Standards New Zealand, Wellington, 2006
SNZ, Steel Structures Standard (Incorporating Amendments 1 and 2), NZS 3404:1997, Standards New Zealand,
Wellington, 2007
Trahair, N.S., Bradford, M.A., The Behaviour and Design of Steel Structures to AS4100, 3rd Edition, E&FN Spon,
London, 1998
Wong-Chung, A.D., Theoretical and Experimental Studies of the Geometric and Material Nonlinear Behaviour of
Partially Braced and Unbraced beams, PhD Thesis, The University of Queensland, 1987
Woolcock, S.T., Kitipornchai, S., Bradford, M.A., Design of Portal Frame Buildings, AISC, Sydney, 1999
Woolcock, S.T., Kitipornchai, S., Deflection Limits for Portal Frames. Steel Construction, Journal of the
Australian Institute of Steel Construction, AISC, 20(3), 2-10, 1986
Woolcock, S.T., Kitipornchai, S., Tension Members and Self Weight. Steel Construction, Journal of the Australian
Institute of Steel Construction, AISC, 1(1), 2-16, 1985
AS/NZS1252. 2016. "High-strength steel fastener assemblies for structural engineering - Bolts, nuts and
washers." In. Sydney, Australia and Wellington, New Zealand: Standards Australia and Standards New
Zealand.
AS/NZS4600. 2018. "Cold-formed steel structures." In. Wellington, New Zealand: Standards New
Zealand/Standards Australia.
Clifton, G. C. 1994. "New Zealand Structural Steelwork Limit State Design Guides Volume 1, HERA Report R4-
80." In. Manukau City, New Zealand: New Zealand HERA.
Clifton, G. C., and R. Z. El Sarraf. 2011. "New Zealand Steelwork Corrosion Coatings Guide, Second Edition." In,
1-90. New Zealand Heavy Engineering Research Association.
Clifton, G. C., and E. Forrest. 1996. "Notes prepared for a seminar on design of steel buildings for fire
emergency conditions." In.: HERA Report ; R4-91, Manukau City, New Zealand : HERA.
SNZ_TS3404. 2018. "Durability requirements for steel structures and components." In. Wellington, New
Zealand: Standards New Zealand.
Section 3 Page 24
CIVIL 713

SEMESTER 1 – 2020
SECTION 4 – COMPOSITE BEAM DESIGN and DETAILING
Section 4.1: Notes on Composite Beam Behaviour and Design

Covers:
• design principles of composite construction: notes from
a seminar on this topic in 2002 with updates in 2016 to
2019
• notes on deflection of composite beams to be read in
conjunction with the notes and design example in
section 4.2 and the Student Standard for Design of Steel
Structures
• mesh reinforcement details
Section 4.2: Composite Beam Design.

Comprehensive set of notes covering:
• sequencing of design prior to composite action
occurring
• concept of composite action
• composite beam theory
• composite beam design example for strength
• serviceability limit state checks
This page is blank
Civil 713: Volume 2 Section 4.1
Design Principles of Composite Construction by G Charles Clifton
These notes have been written by Charles Clifton, previously HERA Structural
Engineer and modified for CIVIL 714 September 2008 with minor revisions 2012.
They provide brief coverage of the principles for design of composite floor systems.
They make reference to a number of publications which designers are expected to
have ready access to. These are:
1. NZS 3404 [1]
2. NZS 4203 [2]
3. HERA DCB [3] with issue number and pages as given in the text.
4. HERA Design Guides Volume 2, Section 13 [4]. This section of the design
guides volume 2 provides a general background to composite design.
Terminology, notation and references are out-of-date, however specific
material from there is still current and is referenced from these notes.
5. Manufacturers’ design manuals, especially for steel decking.
6. HERA Report R4-82 [5] for design of composite slabs in fire.
Section 4.1 Page 1

Section 4.1 Page 2

For secondary beams, construction stage 1:

• beams carry decking dead load and decking construction live load. See Table
13.3 of DGV2 [4] for the magnitude of construction live load
• beams are unrestrained between end supports hence member moment capacity
will govern
• Mbx = m s Msx
m = 1.18 (for UDL over length)
s calculated for kl = kr = 1.0
kt determined for the twist restraint offered by end support conditions.
See R4-92 [7] for details.
This is typically FF, giving kt = 1.0
For secondary beams, construction stage 2:
• beams support wet concrete load, including ponding allowance and concrete
placement construction live load, Qcon. These are applied as UDLS. See Table
13.3 from [4] for the magnitude of Qcon.
• beam has full lateral restraint from decking at this construction stage
• Msx is the relevant moment capacity to use in the bending moment adequacy
check.
For primary beams, construction stages 1 and 2:
• lateral support conditions are the same under stage 1 and stage 2, therefore
• concrete placement (stage 2) governs, as the loads are much higher
• segment length is the distance between incoming secondary beams
• moment across segment is near uniform
• Mbx calculated from NZS 3404 Clause 5.6
m  1.0 (for uniform moment)
s based on 0.85 Lsegment ; kl = 1.0
For deflection
• based on bare steel properties, ie. Is
• see section 1.4 of R4-107 [6] for method of calculation; covered in later sessions
• allowance for support restraint: see section 1.4.2 of [6]
- simple connections: no support moment
- continuous beams: use 0.38 x simply supported deflection.
- for rigid or semi-rigid connections: calculate support moment and use in
deflection calculations as described in section 1.4.2.2 of [6].
Section 4.1 Page 3

Refer to R4-92 [7] for the restraint classifications for all commonly used details.
With secondary beams and ribbed decks:
• ribs perpendicular to span of beam means ribs oriented at 90o  45o to the beam
span
With primary beams:
• these are divided into segments by incoming secondary beams

• the critical segment incorporates the position of maximum design moment
Section 4.1 Page 4

Determination of deck strength and deflection characteristics requires extensive

experimental testing. The appropriate standard is BS5950.4 [8]. The decking
manufacturers will do this and publish the results in design manuals. The designer then
uses these manuals for decking design.
Design aspects to consider and where to obtain the material from:
1. From manufacturer’s design manual for
- design considerations
- section properties
- wet concrete spanning capabilities, unpropped and propped
- reinforcement for normal crack control
- fire emergency additional reinforcement in accordance with R4-82 [5]
- load carrying capacity
- support requirements
- end closure details
2. For concentrated load application
- BS 5950.4 [8]
3. For diaphragm action (eg. as seismic floor diaphragm)
- see advice in CIVIL 714 Lecture Notes on seismic design
- may need to reinforce irregular shaped diaphragms as in-plane deep

beams to develop the required in-plane resistance. This involves deep
beam or strut and tie design to NZS 3101 [14]. The steel beams
supporting a concrete slab on steel deck provide tension members for
diaphragm resistance
Section 4.1 Page 5

Slab effective thickness and width are covered in the next slide
Full or partial composite action?
• partial is preferred
• down to 50% of full requirements is allowed for strength
• down to 25% of full requirements is allowed for deflection
• see NZS 3404 Clauses 13.1.2.4 and 13.4.6
• partial composite action is only permitted in positive moment regions, but
• not permitted in positive moment yielding regions due to earthquake
Negative or positive moment action required?
• both covered by NZS 3404

• composite action is typically only used in positive moment regions
Shear stud number and spacing covered in later transparency and in CIVIL 714
lecture notes
Longitudinal shear capacity covered in later transparency and with regard to

diaphragms in CIVIL 714 lecture notes
Suppression of bottom flange yielding under serviceability conditions.
If the beams are unpropped, the bare steel section supports the wet concrete
loading, which introduces tension stresses in the bottom flange. These stresses are
“locked in” when the concrete sets.
Once the composite section is fully effective, the applied load causes increased
tension in the bottom flange. Under maximum serviceability loading, the combined
tensile stress in the bottom flange must not exceed 0.9fy, in order to keep the beam
behaviour in the elastic range.
Section 4.1 Page 6

&LYLO9ROXPH6HFWLRQ
This requirement is prescribed in NZS 3404 Clause 13.1.2.6(a). It requires the following
to be satisfied:
where:
= moment generated on the bare steel beam during construction stage 2
by the wet concrete loading (including ponding) and the beam, deck
self-weight.
Zx = elastic section modulus for the steel beam
= moment generated on the composite section by the superimposed

dead load and the short-term live load (\sQ)
Ztcb = elastic section modulus of the composite section taken with respect to
the bottom flange.
I = 1.0, from NZS 3404 Table 13.1.2(2).
Suppression of yielding in the reinforcement over the support regions.
This check is only required for continuous composite beams and is specified by NZS
3404 Clause 13.1.2.6(b). It follows the same concepts as above, but using
appropriate values for , , Zx and using Z tct, where this is calculated at the level
of the reinforcement.
6HFWLRQ Page 7
Slab effective thickness provisions are

shown in the opposite figure.
Slab effective width provisions are shown

below for midspan positive moment
regions.
For support negative moment regions,
use 0.6x the positive moment region
value.
For support positive moment regions, see
Clause 13.4.2.3.2. of [1].
The last two are unusual applications in
buildings.
To ensure that the positive moment
composite action can develop before the
slab suffers concrete compression failure
[4]:
2.9 As fy
bec,min  '
f c (d  t o )
This will not be critical in typical building
applications. The background to this
equation is given in section 13.3.5.2 of [4].
Section 4.1 Page 8

7
C.G. of concrete resistance

Effective width bec
Concrete compression 0.85 fc'
area t
Rcc = 0.85fy' a b
P.N.A. ac
hrc
e'
Tension d Rtc = As fy
area
d/2 C.G. of steel resistance
fy
ac = (Asfy)/(0.85fc’b)
Mrc = Afy e' (see NZS 3404 where As is called A)
e' is not given by NZS 3404; it is given by

d a
e'   to - c
2 2
This case is not typical, as full (100%) composite action is not

commonly used.
Section 4.1 Page 9

Effective width bec
Concrete compression 0.85 fc' C.G. of steel resistance
area t t/2
Rcc = 0.85fy' t bec
to hrc t1/2
P.N.A. Rsc = (Asfy - Rcc)/2
Steel compression t1 fy
e'
area, Asc e
d
Steel Tension Rtc = Rcc + Rsc
area
tf
C.G. of steel in tension
bf fy
(a) Plastic Neutral Axis in Steel Flange

Effective width bec
area t t/2
Rcc = 0.85fy' t bec
to tf hrc d3
Rsc = (Asfy - Rcc)/2
d1
P.N.A. fy
Steel compression d e e'
area, Asc tw
Steel Tension tf Rtc = Rcc + Rsc

d2
area
bf fy C.G. of steel in tension
(b) Plastic Neutral Axis in Steel Web
Force Equilibrium of Composite Section

with Full Shear Connection (Case 2)
Section 4.1 Page 10

&LYLO9ROXPH6HFWLRQ
This applies when the steel beam is large relative to the effective area of concrete in the
slab and 100% shear connection is used.
(a) For PNA in steel flange:
(b) For PNA in steel web:
6HFWLRQ Page 11

Effective width bec ac = Rss / (0.85fc' bec)
area t
ac Rcc = 0.85fy' ac bec = Rss
to hrc t1/2
P.N.A. Rsc = (Asfy - Rss)/2
Steel compression t1 fy
e'
area, Asc e
d
Steel Tension Rtc = Rcc + Rsc
area
tf
C.G. of steel in tension
bf fy
(a) Plastic Neutral Axis in Steel Flange

Effective width bec ac = Rss / (0.85fc' bec)
area t
ac Rcc = 0.85fy' ac bec = Rss
to tf hrc d3
Rsc = (Asfy - Rss)/2
d1
P.N.A. fy
d e e'
Steel compression
area, Acr tw
Steel Tension tf Rtc = Rcc + Rsc

d2
area
bf fy C.G. of steel in tension
(b) Plastic Neutral Axis in Steel Web
Force Equilibrium of Composite Section

with Partial Shear Connection (Case 3)
Case 3 applies for partial composite action and is the normal case.
The equations as used in case 2 are applied, replacing the effective slab thickness, t, with
the depth of concrete compression stress block, ac.
Section 4.1 Page 12

Symmetrical about
CL
W/2
Rh* = R CC Rcc
e' Case 1
Rtc
W/2
W/2
Rh* = R CC
Rcc Case 2
Rsc
e e' or
Rtc Case 3
W/2
The horizontal shear flow is given by NZS 3404 Clause 13.4.7. It is dependent on the amount
of composite action required.
Rh *
No. of shear studs over half length of beam in positive moment shown above, n 
scqr
Over the whole length of beam in positive moment, 2n studs are required.
The above is shown for positive moment, which is the typical case. If composite action is
being applied over the negative moment regions, then the same concept applies. In this
instance, n is the number of studs required from the point of maximum negative moment to the
adjacent point of zero moment.
Section 4.1 Page 13

These requirements are illustrated for a simply supported beam by the figure below:
* * * *
M m1 Mm M m1 Mm
A B C D A B C
n' n''=n-n' n' n''=n-n'

n n
( ) ( )
* *
Rh M m1 - Ms Rh M m1 - Ms
n= ; n' = n *
n= ; n' = n *
sc
qr Mm - Ms
sc
qr Mm - Ms
(a) Point Loads (b) Heavy U. D. L.

Number of connectors between M = maximum design moment
AC = n
M = design moment at B
AB = n'
BC = n'' M = design capacity of steel section alone
Rh* = Rcc or
Rh* = total factored connector forces
q = design shear capacity of a connector
Distribution of Connectors as Prescribed

by NZS 3404 Clause 13.4.9.2
In addition, note the detailing requirements in regard to shear stud position within the rib and spacing
along the beam given by NZS 3404 Clauses 13.3.2.2 and 13.3.2.3.
The requirements for stud position within the rib and for rib width are illustrated in Fig. C13.3.2 of [1].
Where a rib profile is such that the stud cannot be placed in the centre of the rib, the studs are placed
on alternate sides of the centre of the rib.
Section 4.1 Page 14

Composite beams that are part of a seismic-resisting system may require shear studs for two
reasons, namely:
(1) To mobilise composite action to resist the full factored gravity loading
(2) To transfer diaphragm shear forces between the floor and the seismic-
resisting system.
The number of shear studs required is covered in section 9.3 of the CIVIL 714 lecture notes on
the seismic-resisting system design.
Section 4.1 Page 15

Differential shear force across the concrete slab can introduce splitting of the concrete.
To suppress this splitting requires reinforcement to pass across the potential failure
plane, with this reinforcement developing a clamping force to resist the splitting action.
There are two cases to consider:
1. Vertical shear splitting from shear transfer at the beam into the surrounding
slab. This is well researched and NZS 3404 Clause 13.4.10 provides design
requirements. The additional reinforcement runs across the beam and the
potential shear failure plane. See the figure C13.4.10 in NZS 3404 [1] Part 2.
2. Horizontal shear splitting from shear transfer through the shear studs into the
overall slab at supports. In this instance a horizontal shear failure can occur,
with the concrete above the shear studs over the beam seperating from that
below. This has been seen in some Australian composite beam tests and
some New Zealand push-off tests for shear stud capacity determination.
This is known as a type 4 shear failure. It is sudden and causes a loss in

load-carrying capacity. However it only occurs in certain circumstances.
The approach in NZS 3404 Clause 13.4.1.3 is to identify the circumstances

that may lead to a type 4 shear failure in the deck and either to:
(i) Make the beam non-composite for strength
(ii) Design and supply suitable reinforcement to suppress a type 4

shear failure. This reinforcement is available from BHP in
Australia.
More details on type 4 failure are covered in session 4.3.
Section 4.1 Page 16

&LYLO9ROXPH6HFWLRQ
6HFWLRQ Page 17
These items are all covered in lecture notes for preliminary design.
Herein is covered determination of the moment of inertia of the transformed section for
composite section deflection calculations.
Details are in the next transparency.
Section 4.1 Page 18

bec/n
t t/2
y
yc
E.N.A.
n = E/Ec (short term)

to +d/2
n = E/2.5Ec (long term)
E = E for steel = 205 GPa
Ec = E for concrete
from NZS 3101 d/2
Cross-Section of Composite Beam For Deflection

Calculations(Transformed Elastic Steel Properties)
The calculation is undertaken as follows:
Section Transformed Distance Ay Ay2 Ilocal

Steel From
Area Top of Slab
mm2 mm mm3
Concrete
Steel
2
Total A _ Ay Ay Ilocal
Section 4.1 Page 19

Calculate
Check that - if it is the details below are correct.
For the case of , calculate Itc as:
and once It is determined, then:
If then you need to account for the area of the equivalent steel beam above
and below the elastic neutral axis, ENA, to calculate Itc. This changes the calculation of
Itc from that given above and requires it to be determined from first principles, taking the
second moment of area of each element of the section about the ENA and summing
these to get Itc.
6HFWLRQ Page 20
Edge supports must be capable of supporting the wet concrete load on them with
minimal deformation (say 5 mm max).
This may require explicit design, especially for cantilevers, or the use of standard pre-
designed details, such as shown above, which is from DGV2 [4].
Good sources of information are:
 SCI Publication P300 [11] - for details

 the decking manufacturer’s design manuals
Section 4.1 Page 21

Good sources of information for suitable details are:
 HERA Report R4-58 [12]

 DCB’s for HERA semi-rigid connection details
 SCI Publication P300 [11]
Decking edge supports to allow the decking to be square cut around obstructions
should always be provided for:
 around columns
 around bolted beam to column connections
 around bolted beam splices
Section 4.1 Page 22

References to notes:
1. NZS 3404: 1997, plus Amendment No. 1: 2001 and Amendment No 2, 2007,
Steel Structures Standard; Standards New Zealand, Wellington
2. NZS 4203:1992, General Structural Design and Design Loadings for Buildings;
Standards New Zealand, Wellington. Now replaced by the AS/NZS 1170:2004
set
3. HERA Steel Design and Construction Bulletin; Bi-Monthly Technical

Publication; Issue Nos. 1-66 published to February 2002
4. HERA; New Zealand Structural Steelwork Design Guides Volume 2; HERA,

Manukau City, 1989/1991, HERA Report R4-49.
5. Barber, DJ; Calculation of the Fire Resistance of Composite Concrete Slabs

With Profiled Steel Sheet Under Fire Emergency Conditions; HERA, Manukau
City, 1994, HERA Report R4-82.
6. Clifton, GC; Draft for Comment: Guide to the Practical Aspects of Composite
Floor System Design and Construction, Including Concrete Placement - Parts
1 to 3; HERA, Manukau City, May 2005, HERA Report R4-107.
7. Clifton, GC; Restraint Classifications for Beam Member Moment Capacity

Determination to NZS 3404:1997; HERA, Manukau City, 1997, HERA Report
R4-92.
8. BS 5950.4:1994, Code of Practice for Design of Composite Slabs With Profiled

Steel Sheeting; BSI, London, England.
9. NZS 3109:1997; Concrete Construction; Standards New Zealand, Wellington.
10. Feeney MJ and Clifton G C; Seismic Design Procedures for Steel Structures;
HERA, Manukau City, 1995, HERA Report R4-76.
11. Couchman, GH et.al; Composite Slabs and Beams Using Steel Decking: Best
Practice for Design and Construction; The Steel Construction Institute; Ascot,
England, SCI Publication No. P300.
12. Manual of Standard Connection Details for Structural Steelwork, Second

Edition; HERA, Manukau City, 1990, HERA Report R4-58.
13. Clifton, GC; Structural Steelwork Limit State Design Guides Volume 1; HERA,
Manukau City, 1994, HERA Report R4-80.
14. NZS 3101:1995 (Including Amendment No. 1, 1999), Concrete Structures

Standard; Standards New Zealand, Wellington.
Section 4.1 Page 23

[This page is blank]
Section 4.1 Page 24

HERA REPORT R4-107:2005
NZ HEAVY ENGINEERING RESEARCH ASSOCIATION Extracts from the

Composite Floor Construction
Handbook on deflection of
composite beams for CIVIL 714
DQGFDOFXODWLRQVIRUVKULQNDJHGHIOHFWLRQ
RQWKHODVWWZRSDJHVRIWKLVVHFWLRQ
Section 4.1 Page 25

HERA REPORT R4-107:2005
Extracts from the Composite Floor

Construction Handbook covering
deflection of composite floor
systems

&KDUOHV&OLIWon, HERA Senior Structural Engineer

Raed Zaki, HERA Structural Engineer
Section 4.1 Page 26

February 2005
Published by:
New Zealand Heavy Engineering Research Association
17-19 Gladding Place
P O Box 76-134
Manukau City
New Zealand
Phone : +64-9-262 2885

Facsimile : +64-9-262 2856
Email : structural @hera.org.nz
February 2005
ISSN 0112-1758
The New Zealand Heavy Engineering Research Association (HERA) is a non-profit

research organisation dedicated to serving the needs of the metal-based industries in New
Zealand. It is the national centre for steel construction, welding, metal fabrication and
machining, and promotes the effective and efficient use of steel in structures.
HERA’s work includes the sponsorship of research and development, the provision of
educational, advisory and information services, the dissemination of technical knowledge
to specifiers, fabricators and suppliers, participation in the activities of relevant national
and international bodies and in the writing of standards and codes of practice.
The results of HERA’s research are published as reports and in the HERA Steel Design
and Construction Bulletin. Wherever possible, this material is formulated to present the
information in a form available for immediate use in design and construction.
Disclaimer
Every effort has been made and all reasonable care taken to ensure the accuracy and
reliability of the material contained herein. However, HERA and the authors of this report
make no warrantee, guarantee or representation in connection with this report and shall
not be held liable or responsible in any way and hereby disclaim any liability or
responsibility for any loss or damage resulting from the use of this report.
Section 4.1 Page 27

Table of Contents
Part 1. Practical Aspects for Designers
1.4 Calculation of Wet Concrete Deflection in Unpropped Steel Beams

1.4.1 Calculation of Mid-Span Deflection 8
1.4.2 Allowance for Support Restraint
1.4.2.1 For simple connections 9
1.4.2.2 For semi-rigid and rigid connections 9
1.4.2.3 For continuous beams 9
1.4.3 Determinations of Deflections at Points other than Mid-Span 9
1.4.4 Special Considerations for Spandrel Beams 10
1.4.5 Special Considerations for Floors with Slopes 10
1.7 Typical Magnitude of Long-Term Floor System Deflections 16
1.8 Factors Influencing the Magnitude of Long-Term System Deflection 17
1.9 Locations Requiring Particular Care in Designing for Long-Term Deflection 18
1.10 Composite Slab Serviceability Limits 18
1.11 Summary and Key Points to Consider Regarding Serviceability Deflection Issues 20
References
Section 4.1 Page 28

Part 1: Practical Design Aspects for Designers
1.4 Calculation of Wet Concrete Deflection in Unpropped Steel Beams
1.4.1 Calculation of Mid-Span Deflection
The mid-span deflection of an unpropped steel beam under wet concrete deflection is determined from
elastic bending theory. A simple and general equation is given in section 2.4.1.2 of HERA Design Guides
Vol. 1 (Clifton, 1994).
For the range of situations required in this application, the mid-span deflection, Δc, is determined from:
L2
Δc = [kMc + 0.312(M A - MB )] (1.2)
I
where:
ΔC = mid-span deflection (mm)
L = span of beam (mm)
I = second moment of area of steel beam (mm4)
MA, MB = support moments (refer to Table 1.2) (kNm)
MC = simply supported mid-span moment (kNm)
k = deflection coefficient (from Table 1.2)
In equation 1.2, the units for the variables must be as defined therein. Expressions for the simple beam
moments MC and values of coefficients k are given in Table 1.2 for a number of loading distributions.
The wet concrete load applied to a secondary beam is uniformly distributed. The wet concrete load applied to
a primary beam is principally input as concentrated loads from the incoming secondary beams. Table 1.2
covers incoming secondary beams at mid-span, third points and quarter points. Other positions may be
obtained from combinations of these. When assigning wet concrete load to the supporting beams, the load is
initially carried by the decking in one-way action to the secondary beams, which transfer it to the primary
beams.
Table 1.2 Mid-span Deflections of Steel Beams (Clifton, 1994)
The load to be considered is made up of two components, namely:
Gc = dead load of steelwork, decking and wet concrete, with the latter determined for the effective
thickness of slab (he from NZS 3404 [15]) based on the specified slab thickness
Gpond = dead load due to wet concrete ponding, calculated using section 1.3.2.
There is no construction live loading used in this calculation, as no permanent set is caused by that loading.
Section 4.1 Page 29

1.4.2 Allowance for Support Restraint
1.4.2.1 For simple connections
Simple connections are those assumed not to develop bending moments under the design actions
(NZS3404:1997) Clause 4.2.2.3. Examples are the Web Side Plate (WP), Flexible End Plate (FE) and Angle
Cleat (AC) connections from HERA Report R4-100, (Hyland, 2003).
In practise these simple end connectors may develop very small bending moments under beam and rotation
due to the wet concrete loads. However, these moments will reduce the mid-span deflection by not more
than 5% and are neglected; hence the deflections are calculated for no support moment restraint.
1.4.2.2 For semi-rigid and rigid connections
Semi-rigid connections are used in seismic-resisting system design, eg. The Flange Bolted Joint (FBJ)
described in (Clifton, 2004).
Rigid connections also are typically only used in seismic-resisting system design, although they may be
occasionally used on long-span primary beams to reduce the beam deflections. Examples include the
Welded Moment (WM) and Moment End Plate (MEP) from (Hyland, 2003).
Both types of connection behave as rigid under the wet concrete loading condition and the support negative
moments generated by this condition can be determined from elastic analysis and then input into equation
1.2 to calculate the net mid-span deflection.
1.4.2.3 For continuous beams
For continuous beams, such as continuous secondary beams over a stub girder (see Figure 13.51 of DG Vol
2 (Clifton, 1989/1991)) or continuous primary beams running past the supporting column (see Figures 15. 28
and 15.29 of DG Vol 2 (Clifton, 1989/1991)) use (0.15/0.4 = 0.38) x the simply supported deflection. This
assumes sequential loading of beam spans with concrete, taking the appropriate factors from Table 1.1.
1.4.3 Determination of Deflection at Points other than Mid-Span
The deflected shape of a member uniformly loaded under wet concrete theoretically approximates a
parabola. However the ratio of mid-span deflection to beam length is so small that the deflected shape can
be considered to be a segment of a circular curve. This is also the case for pre-camber calculations (Ricker,
1989).
To find the deflection at other locations along the span of a segment of a circular curve, the “factor fraction”
method proposed by (Ricker, 1989) is easiest to use. Divide the span into an even number of equal
segments, with the number of segments as desired. Figure 1.4(a) shows 8, Figure 1.4(b) shows 6. Number
the points as shown, starting with zero at the support. Multiply the points as shown to form a factor fraction,
which gives the fraction of mid-span deflection that will occur at that point.
From Figure 1.4 it can be seen that:
• Δ1/3 point = 8/9 of Δmid-span

• Δ1/4 point = 12/16 of Δmid-span = ¾ of Δmid-span
This method is used in pre-camber determination and in setting ponding allowances for concrete placement,
both of which are covered in Part 2 of this document.
Section 4.1 Page 30

(a) Factor fractions for 8 equal segments
(b) Factor fractions for 6 equal segments
Figure 1.4 Use of Factor Fractions to Calculate Deflections along Span
1.4.4 Special Considerations for Spandrel Beams
Spandrel beams provide an edge to the slab. Typically, this edge is formed through the use of a slab edge
form, which is made from the same galvanized steel coil as the decking, and the height is made equal to the
specified slab thickness. An example is shown in Figure 1.5.
Where the spandrel beam is either propped during concrete placement or precambered to compensate for
the wet concrete deflection, this edge form is used as a screed line for concrete placement and will be
effectively level following concrete placement.
Where the spandrel beam is unpropped and not precambered, it will sag during concrete placement,
requiring the finished concrete surface to conform to the deformed shape. A maximum limit of 10 mm is set
for spandrel beams, as given in Table 1.4, to keep the top of slab within acceptable limits.
Figure 1.5 Slab Edge Form at Spandrel Beam

Note to Figure 1.5:
In this photo, only one deck support chair has been placed; these must be at around 750mm centres each way to adequately support
the mesh. The trimmer bar must be hard against the edge bar hooks, as shown in the right hand figure.
1.4.5 Special Consideration for Floors with Slopes
When the top of the concrete is required to be sloped so as to provide falls for drainage, the extent of slope
provided is typically 1 in 50, or 20 mm per metre. An example is an area of floor exposed to rain in a car
parking building, details of which are given in DCB Issue No. 49, (Clifton, 1999).
One method of providing for this slope is to keep the beams level and adjust the concrete thickness. This is
practical for increases in thickness of up to around 80 mm, which can accommodate a 1:50 slope over 4 m.
However, the varying concrete weight will need to be taken into account in floor system design, including
ponding calculations and the maximum unsupported length of slab.
Section 4.1 Page 31

The other option, which is more generally applicable, is to slope the supporting steel beams. In this instance,
the beams should be sloped to the specified fall, in which case the concrete placement follows basically the
same procedure as for placing on supporting steel beams that are level. How to account for the slope in
concrete placement and in testing for flatness and levelness are covered in Parts 2 and 3 of this document,
respectively.
1.7 Typical Magnitude of Long-Term Floor System Deflections
Those involved in composite floor system design, construction and fit-out need to have an appreciation of the
typical magnitude of long-term deflection to be expected from a composite floor system. In the case of fit-
out, this deflection needs to be considered by contractors installing fixed walls and partitions, especially walls
required to meet specified sound performance and/or fire resistance ratings.
The sources of long-term deflection are:
• Shrinkage
• Creep
• Magnitude of long-term applied load on the composite system.
The principal factors influencing the magnitude of long-term deflection are briefly described qualitatively in
the next section. Some guidance on the expected magnitude and time-frame of occurrence of the first two
sources of long-term deflection are now given:
(1) Shrinkage in normal weight concrete slabs of composite beams, where the concrete has been properly
placed and cured in accordance with Section 7 of (NZS 3109:1997), will generate around 1 mm vertical
downwards deflection (at mid-span) for each metre length of composite beam, with a minimum value of 5
mm. This is independent of applied load. Around 45% of the shrinkage-induced deflection will occur in the
first month following concrete placement, with 90% occurring in the first year.
Shrinkage deflection is a significant component of composite floor deflection following construction. For
example, in the system shown in Figure 1.3, if the secondary beams are 10 m long and the primary beams 8
m long, the expected downwards deflection due to shrinkage at the midpoint of the slab panel, relative to the
supports, will be around 10 + 8 = 18 mm. Shrinkage deflection will typically comprise over 50% of the total
long-term deflection, especially in lightly loaded floors that were unpropped during construction.
The time-dependant and one-off nature of shrinkage induced deflection can be taken account of, to
advantage, when an absolute deflection limit from Table 1.4 applies to deflection sensitive non-structural
elements (e.g. floor movement after fit-out) and fit-out does not occur immediately on completion. Based on
the time-dependent recommendations from (Park and Paulay, 1974), if fit-out does not take place until 6
months following concrete placement in a dry internal environment, the value of Δsh to still take place may be
reduced (conservatively) to 33% of the (NZS 3404:1997) calculated value.
If fit-out is taking place 12 months or more following concrete placement, no further shrinkage is likely to
occur.
(2) Creep is an enhancement of the long-term deflection due to permanent, irreversible changes in the
concrete under load. It typically adds around 15% to the calculated long-term deflection based on the elastic
properties of the concrete specified by (NZS 3404:1997).
As a rule of thumb, the creep component of deflection will vary from 20% to 45% of the shrinkage deflection.
The lower proportion applies to unpropped beam systems, the higher proportion to a slab supported on
propped beams during construction and with relatively high long-term loading.
1.8 Factors Influencing the Magnitude of Long-Term Floor System Deflection
The following factors have a considerable influence on the magnitude of long-term deflection:
(a) Beam Propped or Unpropped During Construction
If the beam is propped during construction, then the full dead load of the concrete slab plus finishes is carried
by the composite section, increasing the creep loading. If it is unpropped, only the superimposed dead load
Section 4.1 Page 32

of applied finishes and the long-term component of the live load contribute to creep deflection of the
composite section.
(b) Concrete Age at Removal of Props
If any supporting props are removed too early, a significant additional creep component of deflection can
occur, especially if the beam in question carries propped loads from the floor above. (Follow the guidance of
section 1.6.4 to minimise this effect).
(c) The Magnitude of Live Load Applied Long-Term to the Floor System
This component contributes directly to creep-induced deflection. So live loads with a significant long-term
component (e.g. storage, with (NZS 4203:1992) assessing 60% as long-term) will generate more creep-
induced deflection than will typical office live loads (NZS 4203:1992 assesses 40% of these as long-term).
(d) Steel-Concrete Interfacial Slip and the Use of Partial Shear Connection
These will decrease the short-term and long-term elastic stiffness, and so increase vertical deflection. This is
simply catered for in the design procedure, by following the requirements of (NZS 3404:1997) Equation
C13.1.2(1).
(e) Splitting Capacity of Concrete Around Base of Stud
The longitudinal shear resisted by the shear studs in preventing horizontal slip between the concrete slab
and supporting steel beam develops very high localised compression bearing stresses at the base of the
stud. As these stresses increase towards the ultimate limit state level, in most stud configurations the
concrete at the base of the stud will split, reducing the build-up of bearing compression and allowing
additional longitudinal slip to occur. This in effect decreases the elastic stiffness of the composite beam,
increasing the vertical deflection.
The factors affecting the splitting strength of the concrete around a stud are explained in depth in DCB No.
55 pages 18 to 28 (Clifton, et. al., 2000).
The allowances for interfacial slip and partial shear connection described in (d) do not include the effect of
concrete splitting under serviceability load levels; hence it is important that this is suppressed. Guidance on
how to achieve this is given in the above referenced DCB article; this must be followed for shear stud
applications not covered by (NZS 3404:1997), as described in section 1.2.
(f) The Sectional Profile of the Concrete Slab
The sectional profile of the slab, whether it is ribbed or solid, has an effect on vertical deflection. Once again,
this is simply catered for in the design procedure through (NZS 3404:1997) Clause 13.1.2.5. See also
section 1.2 herein.
(g) The Amount of Shrinkage Strain that Occurs in the Concrete Slab
Guidance is given in (NZS 3404:1997) Commentary Clause C13.1.2.6(c)(iii). The shrinkage strain developed
will be influenced by the profile of the composite decking, being greater at right angles to the decking than it
is along the line of the ribs. The value of shrinkage strain specified by (NZS 3404:1997) and its method of
application in calculating shrinkage-induced deflection correspond to a free shrinkage strain of approx 950
microstrain. A 20% reduction in this value is made to account for the restraint offered by the decking and
reinforcement. See references from Part 2 of (NZS 3404:1997) for more details.
(h) The Strength of Concrete Used in the Composite Slab
This has a minor influence on the magnitude of long-term deflection for normal weight, as it affects the
concrete elastic modulus, Ec.
It has a more significant influence for lightweight concrete.
(i) The Amount and Position of Reinforcement Within the Composite Slab
Section 4.1 Page 33

The reinforcement quantity and location has only a minor influence on creep-induced deflection, however its
influence on crack width and spacing is greater.
1.9 Locations Requiring Particular Care in Designing for Long-Term Deflection
Deflection-sensitive finishes have the potential to be damaged both by the magnitude of deflection occurring
over the lifetime of the finish and by the deflection gradient of the floor on which the finished element is
supported.
Points of maximum deflection gradient can be seen from Figure I.3 on page v and include:
(1) Beams framing into supporting columns, where the gradient is opposite in sense on each side of the
column
(2) Secondary beams running parallel to very stiff elements such as shear walls or perimeter frames. With
reference to Figure I.3, for example, if point B lies on a perimeter frame, its deflection under vertical load
will be negligible. In contrast, point D will deflect by an amount (Δ1+Δ2) and may cause distress to a non-
structural element spanning from B to D. In such locations, the secondary beam size at D may need to be
increased, to keep differential deflections between B and D (which equal Δ1+Δ2) within the acceptable
limits for protection of non-structural elements given in Table 1.4 below (With reference to Figure 1.3, this
would involve limiting the differential deflection between B and D to say L/180 or 20 mm, with L being the
span between B and D.)
1.10 Composite Slab Serviceability Limits
Recommended deflection limits for composite floor systems and components (decking and beams) are given
in Table 1.4. These limits combine the recommendations of DCB No. 33 with the update of DCB No. 52
(Hyland and Kaupp, 1999) along with further updating of the wet concrete deflection limits to those given in
section 1.3 herein.
Design procedures addressing serviceability deflection considerations for composite floor systems are given
in Clause 13.1.2.6 and Clause 13.1.2.6 of (NZS 3404:1997) and in section 13.3.6.4 of HERA Design Guides
Volume 2 (Ruddy, 1986). These requirements are also incorporated into the computer program COBENZ 97;
see details in DCB No. 35, pages 4 and 5 (Clifton, 1997) and in the Hi-Bond Design Wizard (Bird and
Klemick, 2002).
Some brief guidance on using these procedures is given below:
(1) Immediately measured elastic deflection from applied long-term loads is often slightly less than
calculated, whereas the measured total deflection over time (ie. elastic component plus shrinkage and
creep) is usually similar to that calculated.
(2) The recommended shrinkage strain of 300 microstrain from (NZS 3404:1997 Part 2) may be
considered for shrinkage occurring after initial fit-outs, for purposes of determining limits for deflection-
sensitive finishes. Reductions in shrinkage deflection for fit-out occurring later on can be made as
detailed in section 1.7 (1) above.
(3) The creep-induced deflection should also be considered as occurring after initial fit-out for setting
deflection limits. For subsequent changes to long-term loading, a change in creep deflection would
also be expected, but the effect is usually too small to warrant calculating reduced creep deflections
for fit-out occurring well after construction is completed.
(4) The elastic second moments of area (short-or long-term as appropriate) should include an allowance
for loss of stiffness due to interfacial slip, effect of partial composite action and use of ribbed decking,
as specified in (NZS 3404:1997) Part 2, Equation Clause 13.1.2(1).
Section 4.1 Page 34

Table 1.4 Recommended Deflection Limits for Composite Slab on Decking and for Composite Beams
Control Occupancy Comment Deflection Deflection Action Being

Phenomenon Limit Components Considered
From:
Constructability Decking Unpropped L/180 < 20 mm Gc + Gpond Sag (Fisher and West,
Ponding: Soffit 1990)
Beams: Gc + Gpond - Δpc Sag (HERA R4-49,
Internal Secondary Unpropped L/333 ≤ 20 mm 1989/1991)
Internal Primary Unpropped L/500 ≤ 10 mm
Beams: Gc + Gpond - Δpc Sag (HERA R4-49,
Spandrel Secondary Unpropped L/333 ≤ 10 mm 1989/1991)
Spandrel Primary Unpropped L/333 ≤ 10mm
Visual Carpark Unpropped L/300 Gs +ψlQ+Δcr+Δshr Ripple (NZS 4203:1992)
Reflective slab top
surface finish Propped L/300 G+ψlQ+Δcr+Δshr Ripple
Line-of-sight across Carpark Unpropped L/250 Gs+ψsQ+Δcr+Δshr Sag (NZS 4203:1992,
beam soffit Office L/360 -Δpc Fisher and West, 1990)
(underside of
beam) Residence L/360
Carpark Propped L/250 G+ψsQ+Δcr+Δshr Sag
Office L/360
Residence L/360
Composite slab Span/thickne L/to ≤ 35 Sag
ss ratio
Carpark Unpropped L/250 < 35 mm Gs+ψsQ+Δcr+Δshr Sag (NZS 4203:1992)
Functionality
Office L/360 < 25 mm
Surface Slopes
Residence L/360 < 25 mm
(Top of slab)
Carpark Propped L/250 < 35 mm G+ψsQ+Δcr+Δshr Sag
Office L/360 < 25 mm
Residence L/360 < 25 mm
Protection of Non- Ceiling/Partition Both L/360 ψsQ+Δcr+Δshr Damage at architraves
Structural Elements Junctions (Fisher and West, 1990)
Floor movement Joint cracks (NZS
after fit out Framed Gypsum 4203:1992, Fisher and
Both L/300 ψsQ+Δcr+Δshr West, 1990)
Walls
Masonry Walls Both L/500 Gwall+ψsQ+Δcr+Δshr Wall cracks (NZS
4203:1992, Fisher and
West, 1990)
Check across the Carpark Propped and Ldiagonal/250 Gs+ψsQ+Δcr+Δshr Sag across the
diagonals of a slab Office unpropped diagonals of a slab panel
panel bounded by supporting
Residence beams
See note 2
Notation for Table 1.4: Note; the limits of 25mm

Gc = Dead load of steelwork, decking and wet concrete, excluding ponding effects and 35mm for
Gs = Dead load superimposed upon composite section
Gpond = Dead load due to wet concrete ponding functionality of the
G = Gc + Gs + Gpond (for calculation of Gpond, see Table 1.1)
Gwall = Dead load due to masonry wall
composite beam has
Δpc = Precamber (note precamber is upwards!) been increased to 40mm
ψlQ = Long term portion of live load, from (NZS 4203:1992)
ψsQ = Short term portion of live load, from (NZS 4203:1992)
as of 2014
Δshr = Concrete drying shrinkage induced deflection, calculated to (NZS 3404:1997) Commentary Clause C13.1.2.6(c)(iii)
Δcr = Concrete creep deflection under the long term load, calculated to (NZS 3404:1997) Commentary Clause C13.1.2.6(c)(ii).
For an unpropped beam, the long term load consists only of the long term superimposed load on the composite section.
For a propped beam, the long term load includes the full weight of the slab plus beam
to = Overall thickness of composite slab
Notes to Table 1.4
1. The concrete surface is assumed to be screeded to level in all instances covered by this table
2. The check across the diagonals of a slab panel bounded by supporting beams in two directions can be critical even when both sets
of supporting beams meet their appropriate limits. This check is made for the length on the diagonal
Section 4.1 Page 35

3. When calculating Δcr as used in the combinations given in this table, Δcr must be just the creep component. If the first option in (NZS
3404:1997) C13.1.2.6 (c)(ii) of increasing the calculated deflection under long-term loading (calculated using the short-term modular
ratio) by 15% is used, then Δcr as used in this table represents just the additional creep component, ie. 0.15ΔG + ψlQ
Designing the floor system for satisfactory in-service vibration characteristics is another critical serviceability
criterion. Excellent guidance is available (Murray, et. al., 1997) and (Allen, et. al., 1999), with an overview of
this given on pages 25 to 28 of DCB No. 56 (Clifton, 2000). The computer programs HiBond Design Wizard
and NZFl_Vib 1 (Khwaounjoo, 2002) do vibration design, with the latter covering all the floor systems
described and shown in Figure I.2, including long span decking.
1.11 Summary and Key Points to Consider Regarding Serviceability Deflection Issues
The key points to be kept in mind by designers / specifiers / constructors of composite floor systems, with
regard to serviceability deflection issues, are:
(1) Floor systems utilising composite steel concrete beams must be designed for both short- and long-
term deflections.
(2) The principal components of deflection can be reliably identified. Creep and shrinkage must be
accounted for in any composite beam deflection assessment.
(3) The magnitude of these components of deflections can be dependably estimated by the existing
design provisions.
(4) Matching the magnitude of short- and long-term deflections against appropriate deflection limits
ensures acceptable serviceability performance. A table of reasonable deflection limits, which are
based on local and international experience for typical occupancy uses, is included as Table 1.4.
(5) In particularly susceptible locations, either deflection-sensitive finishes should be avoided or detailing
provided to accommodate the expected movement.
(6) Decisions made by the designer on choice of decking, supporting beam system and layout, propping
and precambering have significant influences on concrete placement and constructability. The design
engineer must clearly communicate to the contractor, through the contract documents, the following
information:
• allowances for wet concrete deflection in setting concrete heights for screeding (Part 2)
• propping requirements (section 1.6 and Part 2)
• precambering requirements (section 1.5 and Part 2)
• surface finish tolerances (Part 3)
• the expected mid-span beam deflection under wet concrete loading (section 1.3.3 and Part 2)
If these key points are followed, composite floor systems will behave as expected and predicted during
construction and will deliver acceptable in-service performance with regard to concrete finish, deflections and
in-service vibration.
Section 4.1 Page 36

References
ACI; Standard Specifications for Tolerances for Concrete Construction and Materials; American Concrete
Institute, Detroit, USA; 1990, ACI 117-90.
Allen, DE et.al.; Minimising Floor Vibration; Applied Technology Council, Redwood City, USA, 1999, ATC
Design Guide 1.
ASTM E1155M-01, Standard Test Method for Determining FF Floor Flatness and FL Floor Levelness
Numbers [Metric]; American National Standards, Washington, USA. 2001.
AS/NZS 4671:2001; “Steel Reinforcing Material”; Standards New Zealand, Wellington, 2001.
Bird, GD and Klemick, MP; HiBond Design Wizard for Composite Design of the Hi-Bond Flooring System,
Version 1.2; Dimond, Auckland, 2003.
Clifton, GC; Structural Steelwork Limit State Design Guides Volume 1; HERA, Manukau City, 1994, HERA
Report R4-80.
Clifton, GC; Structural Steelwork Design Guides Volume 2; HERA, Manukau City, 1989/1991, HERA Report
R4-49.
Clifton, GC; Notes Prepared for “Composite Steel Design and Construction Seminar”, HERA Report R4-113,
2002.
Clifton, GC and Robinson, J; Notes Prepared for a Seminar on The Behaviour and Design of Multi-Storey
Steel Framed Buildings for Severe Fires, Revised June 2001; HERA, Manukau City, HERA Report R4-105.
Clifton, GC; “Semi-Rigid Joints for Moment-Resisting Steel Framed Seismic-Resisting Systems” ; PhD
Thesis Report, University of Auckland, Auckland, 2004.
Cook, D; Floor Tolerance Measurements from Around the World; New Zealand Concrete, March, April 2001,
pp 30,31.
Couchman, GH et. al.; Composite Slabs and Beams using Steel Decking: Best Practice for Design and
Construction; The Steel Construction Institute, Ascot, England, 2000, SCI Publication P300.
Fisher, JN and West, MA; Serviceability Design Considerations for Low-Rise Buildings; American Institute of
Steel Construction, Chicago, USA, 1990.
Guide to Concrete Floor and Slab Construction; American Concrete and Slab Construction; American
Concrete Institute, Detroit USA; 1996, ACI 302.1R-04.
HERA Steel Design and Construction Bulletin; Periodical technical Bulletin, Issue No. and Page Nos. as
specified in text.
Hi-Bond Design Manual; Dimond Structural, Auckland, 1997.
Hyland, C; COBENZ 97, Hyland Consultants Ltd, P.O. Box 23-508, Papatoetoe, Manukau City
Hyland C; Structural Steelwork Estimating Guide v1.0; HERA, Manukau City, 2004, HERA Report R4-96.
http://www.hera.org.nz/steelest/index.asp.
Hyland C; Structural Steelwork Connections Guide; HERA, Manukau City, 2003, HERA Report R4-100.
Hyland, C and Kaupp, T; Optimising the Cost of Steel Buildings in New Zealand; HERA Steel Structures
Analysis Service, Manukau City, 1999.
Section 4.1 Page 37

Appendix A
Khwaounjoo, YR; Report and Users’ Manual for NZFl_Vib 1 Program; HERA, Manukau City, 2002, HERA
Report R4-112.
Larson, JW and Huzzard, RK; Economical Use of Cambered Steel Beams; Proceedings, 1990 National Steel
Construction Conference; American Institute of Steel Construction, Chicago, USA, 1990.
Mullet DL and Lawson RM; Design of Slimflor Fabricated Beams Using Deep Composite Decking; The Steel
Construction Institute, Ascot, England, 1999, SCI Publication P248.
Murray, TM, Allen, D. E., and Ungar, E. E.; Floor Vibrations Due to Human Activity; American Institute of
Steel Construction, 1997, Steel Design Guide Series 11.
NZS 3109:1997, Concrete Construction; Standards New Zealand, Wellington.
NZS 3114:1987, Specification for Concrete Surface Finishes; Standards New Zealand, Wellington,
NZS 3404:1997 (Incorporating Amendment No. 1: 2001), Steel Structures Standard; Standards New
Zealand, Wellington
NZS 4203:1992, General Structural Design and Design Loadings for Buildings, Standards New Zealand,
Wellington.
Oehlers and Bradford; Composite Steel and Concrete Structural Members, Elsevier Science Ltd, Oxford,
U.K. 1995.
Park, R and Paulay, T; Reinforced Concrete Structures; John Wiley and Sons, 1974.
Ricker, DT; Cambering Steel Beams; Engineering Journal / American Institute of Steel Construction, Fourth
Quarter, 1989, pp. 136-142.
Ruddy, JL; Ponding of Concrete Deck Floors; Engineering Journal / American Institute of Steel Construction,
Third Quarter, 1986, pp. 107-115.
Speedfloor Design Manual; Speedfloor Holdings Ltd; Auckland, 2000.
Svelte-floor Design Manual; Formsteel Industries; Auckland, 2003.
Tray-dec 300 Design Manual; Forgan Jones Structural Ltd; Auckland, 1994.
Uni-floor Design Manual; Formsteel Industries; Auckland, 2003.
Zaki, R, Charles Clifton and John Butterworth; Shear Stud Capacity in Profiled Steel Decks, HERA, Manukau
City, 2003, HERA Report R4-122.
Section 4.1 Page 38

Civil 714 Deflection in Composite Beams
written by G Charles Clifton 03 October 2008, updated July 2009.
Calculation of shrinkage deflection:

• this is the most significant component of long term deflection
• pulls the beams downwards
• magnitude in simply supported beam ≈ 1mm per metre length but not less than
5mm
• See extracts from NZS 3404 C13.1.2.6 next page. Read in conjunction with diagram
below:
Concrete cover t yc = y-t/2
=sh Ec Ac =sh Ec Ac
y E.N.A.
Steel beam
Assumed uniform moment diagram
Msh = yc =sh Ec Ac
2 2
sh = Msh L /8EItc = yc =sh Ec Ac L /8EItc
Shrinkage Deflection of Composite Beams

(by analysing the structure as an eccentrically loaded column)
Deflection of composite beams.
Section 4.1 Page 39

Extracts from NZS 3404 Commentary Clause C13.1.2.6 on
shrinkage determination, allowance for interfacial slip from
partial composite action and creep deflection
Ec = short term modulus of elasticity, concrete
To account for increased flexibility from interfacial slip and partial shear connection,
calculate the deflections using an effective moment of inertia given by Equation C13.1.2.(1)
To account for creep, increase the elastic deflections due to dead loads that act on the
composite section and long term live loads calculated from the above equation by 15%.
This is appropriate for floor systems so loaded after the concrete has attained its specified
28 day strength (which is typically after 21 days). If it is loaded earlier (such as in propped
floors if the props are removed earlier than 21 days after the concrete is poured) the creep
deflection will be higher.
Deflection of composite beams.
Section 4.1 Page 40

Mesh reinforcement details from HERA Report R4-131 SPM Method
Table 1: Mesh Relationship between Mesh Type and Bar Diameter, Spacing
Mesh Type Reinforcement Details
(1)
Bar Diameter Bar Spacing (1) Cross Sectional Strength Grade(3)
Ductility Class(4)
(mm) (mm) Area (mm2/m)
668 4.0 150 84(2) 500 L
666 5.0 150 131(2) 500 L
665 5.3 150 147 500 L
664 6.0 150 188 500 L
663 6.3 150 208 500 L
662 7.1 150 264 500 L
661 7.5 150 295 500 L
HDM 430-150 (5) 7.0 256 (6) 150 430 E
HDM 430-200 (5) 8.0 251 (6) 200 430 E
HDM 430-250 (5) 8.0 251 (6) 250 430 E
HDM 430-300 (5) 10.0 260 (6) 300 430 E
Notes to Table 1:
1. The bar diameter and spacing apply in both directions
2. These areas are not sufficient for preservation of integrity (see section A4.3, A5.3) for use of the SPM method in fire
Section 4.1 Page 41

3. This is the nominal yield stress, in MPa
4. This is the minimum uniform elongation, as specified by Clause 5.2(c) and Table 2 of AS/NZS 4671 [11]. The three Ductility Classes are L, N and E; minimum elongation requirements are given
in Table 2 of [11]
5. Details of this mesh are given in [12]
6. These bar spacing are the average for the sheet of mesh; the spacing for the edge two bars in each sheet are different to the others.
713 – Structures and Design 4 – 2020– Composite Beam Design
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SECTION 4.2 - Composite Beam Design

Introduction
This section comprises a summary set of notes taken from Civil 714 and modified for
Civil 713, followed by the notes below which are from Colin Nicholas’s original course on
composite system design and provide a more general explanation of the principles and
application of composite beam design than section 4.1 provides. The recommendation is
to read through the section 4.1 notes, then to use the notes in this section for more
detailed coverage of the key aspects of composite beam design.
Also note that a new composite standard, AS/NZS 2327 has been published in 2017 but
contains quite a number of errors and an errata is being published in 2019. Until then we
will continue to teach from the NZS3404 provisions for composite steel/concrete design,
which are in Section 13. The notation herein aligns with NZS3404.
Sequencing of design prior to composite action occurring

1. More often than not during construction the steel structure (whether primary or
secondary) is subjected to construction loads such as self weight, formwork
(permanent or temporary), wet concrete loads, and imposed construction loads
2. At this point the steel beams act on their own under the propped or unpropped
spans, end and lateral support regime pertaining at the time
3. This condition needs careful consideration and often requires the designer to discuss
or at least think through a given construction sequence of which the building
contractor should be aware through the drawings or specification for the project.
4. Any deviations from this “understood” sequence of construction can create problems
if not unsafe (and possibly catastrophic) conditions on site. This is the specific
reason for NZS 3404 Clause 13.2.4 Bases for design and construction. It is very
important that the bases on which the design has been undertaken are clearly
communicated to the fabrication, erection and construction teams.
5. In designs of long span, no propping and wide tributary areas, a “ponding” allowance
should be made if significant deflection occurs during concrete placement while steel
only is operating. This can be significant but depends largely on the span and the
initial size of steel member chosen. An initial allowance of 10% additional thickness
to the wet concrete slab thickness can be tried and back checked. This is also
covered in the following notes. Note that continuity of the steel beams can
substantially reduce deflections but take care when concrete deck placement can
mean wet concrete is present in one span and not the other
6. Lateral restraint afforded by steel decking can be checked by reference to the
following notes. The extent of lateral restraint depends on whether the decking is
spanning parallel to the beam or perpendicular to the beam and it differs from the
construction stages to the finished composite floor stage. Refer to the slides at the
beginning of this section for more details and to HERA Report R4-107 for in-depth
coverage.
7. In bridge beams especially, often the steel beams are designed to be simply
supported during construction then continuous in service. It means that the
construction loads are taken by the simply supported steel beams while the
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operating loads are taken by the continuous beams. This offers efficiency in design
for carrying the operating loads which for a bridge are high, but means careful
analysis of the continuous beam where the supports are resisting negative moments
by composite action and the beams away from the supports are resisting positive
moments by composite action. Accurate determination of the stiffness of the positive
and negative moment regions of the beams in this type of construction is critical.
Concept of composite action (covered in this course only in positive moment

regions which is typically the case for buildings) –
Connection between Slab and Steel
In order for the structural system to act compositely and transfer horizontal shear forces
(ie parallel to the flange of the beam) at the interface between slab and steel flange,
shear connectors are provided – usually
• Proprietary arc welded studs (buildings and bridges)
• Rolled Steel Channels or Joists welded to the flange (bridges not buildings)
This course will only consider 19 mm proprietary studs in solid insitu concrete slabs –
these comply with the geometry required by the NZS 3404 Section 13
Design Rules have been adopted after extensive testing in Europe and Canada and
three probable modes of failure are considered
• Pull out of the shear stud through the concrete where

qr = 0.13 . αdc . Asc . fu . √f’c …………….Eqn 13.3.2.1
• Shearing of the Connector
qr = 0.8 . fu . Asc ………………Eqn 13.3.2.1
• where Asc = shank area of the stud
fu = 415 MPa Ult Tensile Strength of Shear Connector
• The standard requires a reduction factor of αdc to be applied if steel deck ribs,
parallel or perpendicular to the beam, are used. In this course assume insitu
concrete slab – ie αdc = 1.0 – Refer Cl 13.3.2.1
• A third failure can occur where the welded stud pulls out of a very thin flange – in
this respect the Standard requires the stud diameter be no greater than 2.5 times
the flange thickness – eg for a 19 mm stud the minimum thickness of flange = 7.6
mm Ref Cl 13.3.2.3 (c)
Note also that the shear force (for an applied UDL) which can develop between the
beam and the slab increases from zero at midspan to a maximum at the support. In
these design rules we calculate the shear force which can be transferred by providing
sufficient shear connectors over half the span uniformly distributed over the span. While
this course only deals with UDL load distributions, for the effect of point loads and
different shaped shear force and bending moment diagrams the recommendations of
NZS 3404 Clause 13.4.9 and HERA Report R4-107 need to be referred to
1. Shear Stud requirements (19 mm proprietary item) Cl 13.3.2.3

a. Edge distance from stud centerline > 35 mm
b. Minimum longitudinal spacing 100 mm
c. Maximum longitudinal spacing 800 mm or 8 times slab thickness
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2. Effective Slab Widths and Thicknesses
References – HERA Notes of Seminar on Composite Construction Design

AISC Composite Construction Design Guide
CIVIL714 Charles Clifton Notes on Composite Design
In order to act as a composite section of concrete slab and steel beam it is important to
determine how much of the slab can be considered as part of the structural cross section
– ie a very short span for example would not act as well as a very long span – the
designer could take a substantially greater proportion of the slab into account for the
longer span. In addition when considering the stress block within the concrete the
concrete thickness comes into play which is dependent on whether or not the slab is full
insitu depth or ribbed (due to profiles of steel decking) perpendicular or indeed parallel to
the steel beam
• The effective slab thickness is dependent on the type of slab – refer attached
sheet for various steel deck profiles. This course only deals with full depth insitu
slabs but profiled steel decking incorporating a bottom profile which is not flat
would need specific attention (refer attached sheets). “Span” = full span for
simply supported members or 0.7 span for continuous members.
• For this course with insitu full depth slab tc = to
• The effective width of slab to be taken into account in design is the lesser of
o Span / 4 (building floors) or Span/5 (bridges)
o (S1 + S2) / 2 (Average of the distances between adjacent support points)
o 12 times the least thickness of the slab (bridges only)
• Note special case of an edge beam refer attached sketches
• Note also various slab depth calculations depending on the type and orientation
of rib within a ribbed slab
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Composite Beam Design – Theory and Example

The concrete slab carries compression forces generated by the number of shear
connectors provided between the slab and the steel and this is a function of the –
- concrete strength f’c
- slab effective depth to
- slab effective width, b1
- the area of the steel beam As
- the yield stress of the structural steel
The steel beam carries both tension and compression as the plastic neutral axis should
lie within the steel section (either within the flange of the steel beam or in the web and
requires analysis)
Strength Reduction factors (φ) are given in Table 13.1.2 (1)
R =
Qr ss
= either
0.85f’c.a.bor
0.85f'c.a.b1 ec
OR Asfy
Asfy
Plastic NA
Rsc
Cr = (Asf-y-R
= (Asfy Qr)cc/)/2
2
Steel in Tension
Tr = Cr + Qr
Rst = Rsc +Rcc
The design process is to select a beam size and analyse conditions for long term
strength requirements and the number of shear connectors and then check construction
and final serviceability and strength requirements
The number of shear connectors to be provided is at the discretion of the designer.
She/He can provide “full” shear connection which achieves in one half the span (or from
the point of zero moment to the point of max moment) sufficient shear connectors to put
the beam into yield (Asfy) or the concrete into crushing (0.85f’c.a.bec). However, the
provision of shear connectors in composite construction is a costly exercise (high labour
content) and often congestion of shear connectors can be a problem on site. For this
reason designers attempt to minimize the number of shear connectors required. Too
fewer shear connectors mean you are not getting the full benefit of loads being
transferred to the concrete. Moreover, using say a 50% “partial shear connection” will
achieve over 80% of the full composite flexural capacity of the composite section so any
additional shear connectors above say 50% only provide some 20% extra strength – not
particularly cost effective
Accordingly minimum code requirement for partial shear connection is 50% full
requirement. In general this is found to give adequate strengths for normal load
requirements. In addition (and often) the choice of steel beam is created by construction
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requirements (say, deflection or strength under wet concrete conditions), the slab
thickness is governed by between-beam strength and so the “composite action” analysis
is to ensure that sufficient strength is there to cater for the ultimate design actions and
therefore only serves to check this and achieve the number of shear connectors needed
Note the expression “one half of the span” is badly worded and actually means between
the position of zero moment to the point of maximum moment. This means if
calculations show you require 12 shear connectors to provide adequate shear
connection then the beam should have 12 shear connectors either side of the point of
maximum moment, that is 24 connectors in all if uniformly loaded. This can get messy
for composite beams under point loads (refer 13.4.9.2) which happens with primary
beams which are loaded from the incoming secondary beams.
Spacing and Geometry of shear connectors – for a uniformly distributed load the
value of horizontal shear at any point in a cross section along a beam varies and follows
the same shape as the vertical shear force, that is, triangular, zero at midspan and a
maximum at the ends. At first glance, it would appear appropriate that shear connectors
transferring this shear from the steel beam flange into the concrete slab should be
spaced closer at the ends of the beam with wide spacing at the centre satisfying the “so-
called” triangular distribution. In practice this doesn’t occur. Uniform distribution of load
is often not the norm with industrial beams having to carry point loads or in bridge beams
where moving loads create a plethora of shear force diagram shapes. In addition the
high concentration of shear connectors as implied by the “triangular distribution” can
create congestion on site which is far from favourable. Moreover research has shown
that shear connectors prior to ultimate failure have a propensity to achieve redistribution
of loads between them. Accordingly the recommendation is to provide uniform spacing
to the connector groups. Refer Student Standard Cl 13.3.2.3 for Shear Stud geometry,
minimum spacing and edge distances
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Derivation of the formula

𝐴𝐴𝑠𝑠 𝑓𝑓𝑦𝑦 − 𝑅𝑅𝑐𝑐𝑐𝑐
𝑅𝑅𝑠𝑠𝑠𝑠 =
2
Cr'R=ccQr= Rss
astic NA Cr
fy Rsc
As
Steel in Tension
Tr = Cr + Qr
Rst = Rsc +Rcc
fy
At Ultimate Conditions it is considered that the Steel beam is subject to compression

stresses above the plastic neutral axis of fy and tension stresses below the plastic
neutral axis of fy
For ease of the Calculation, extend the stress block so that the entire section of the steel
beam (Area As) is under tension stresses of fy and increase the compression block by an
identical amount of fy
𝑇𝑇 = 𝐴𝐴𝑠𝑠 𝑓𝑓𝑦𝑦
2𝑅𝑅𝑠𝑠𝑠𝑠 + 𝑅𝑅𝑐𝑐𝑐𝑐 = 𝐴𝐴𝑠𝑠 𝑓𝑓𝑦𝑦
𝐴𝐴𝑠𝑠 𝑓𝑓𝑦𝑦 −𝑅𝑅𝑐𝑐𝑐𝑐

Therefore - 𝑅𝑅𝑠𝑠𝑠𝑠 = 2
Values of Modulus of Elasticity of Concrete used in the calculation of the modular

ratio and in the calculation of deflections due to concrete shrinkage
Ec = 3320 f c' + 6900 = 25084 MPa for say, 30MPa concrete (Clause 3.8.1.2 of the
Student Standard for Reinforced Concrete)
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EXAMPLE – COMPOSITE BEAM ANALYSIS
Single span, simply supported, single lane bridge beam, UDL only (no point loads)
Steel beam spans 10 m, carries 2.5 metres of deck slab (Say, Distributed LL = 5.5
kN/m2). No propping or vertical support inspan during construction. Formwork does not
provide any lateral support to the beam. Ultimate design actions given by…
M* = 424 kNm (say) V* = 170 kN (say) Note these are given figures
and not calculated
Assume solid full depth insitu 150 slab, f’c = 30 MPa

Use 19 mm (fu = 415 MPa) Shear Studs
Check 410UB54 Steel beam acting compositely with the concrete deck
Capacity Shear Connectors –
Asc = π * 192 / 4 = 284 mm2 αdc = 1.0 fu = 415 MPa
Crushing of concrete qr = 0.13 * 1.0 * 284 * 415 * √ 30
= 83.9 kN ← Governs
Shearing of Stud qr = 415 * 284 * 0.8
= 94 kN
Flange thickness ~ 10.9 mm - Max Stud dia of 2.5 * 10.9 = 27 mm
Actual Stud dia = 19 mm OK So ADOPT – qr = 84 kN/stud
The maximum force which can be transferred from the steel beam into the concrete slab
is governed by the number of shear studs provided
NZS 3404 recommends lower limit of 50% of shear connection required for FULL shear
connection
Full Shear connection φRss is the lesser of
As fy - Steel Yielding
OR 0.85 . f’c . bec . tc Concrete crushing
In this instance –
As = 6890 mm2 fy = 300 MPa
tc = to = 150 mm f’c = 30 MPa Ec = 25084 MPa (P 12)
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Slab effective width b1 is the least of three criteria: -
Span / 5 = 2000 mm or Average spacing between beams = 2500 mm
or 12 * 150 = 1800 mm ← Governs
ADOPT bec = 1800 mm
Then As fy = 2067 kN ← Governs for full shear connection
OR 0.85 . f’c . bec . tc = 6885 kN
Try 13 Studs ~ 13 * 84 = 1091 kN (For 50% shear connection, total capacity of the
shear studs should be at least ~ 50% of 2067 = 1034 kN OK)
This means for a uniformly distributed load (with maximum moment at midspan) that the
beam requires 13 studs either side of midspan - assume clearance to first stud at each
end is, say, 190 mm and that 26 studs provide 25 spacings,
190 + 190 + 25 * s = 10000 Therefore s ~ 385 mm cs
Recommended maximum stud spacing ~ 8 . to or 800 mm = 800 mm OK
Then ADOPT 26 studs @ 385 cs + 190 end distance (each end) – layout and spacings
complies with Cl 13.3.2.3? OK
Then φRss = φΣqr = 13 * 84 = 1091 kN ← check > 1034 kN OK
Concrete Compression –
Rss = Rcc = 0.85 . f’c . a . bec = 1091 kN
Then a = 1,091,000 / (0.85 * 30 * 1800) = 23.8 mm
Rsc (Compression in the Steel Beam) = (As . fy – Rcc) /2
= (2067 – 1091) / 2 = 488 kN
Therefore Rsc = 488 kN
Locate Position of the neutral axis
Maximum possible Rsc available in the compression flange of the 410UB54 is given by
(taken to full yield stress)
Flange Area * fy = 178 * 10.9 * 300 = 582 kN > 488 kN
So PNA is within the flange of the 410 UB
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𝑅𝑅𝑠𝑠𝑠𝑠 48,000
𝑡𝑡 = = = 9.1𝑚𝑚𝑚𝑚
178 𝑥𝑥 300 53,400
Crsc acts at the centroid of

R
the area of steel in t1 = depth of compression
compression block in steel flange
Plastic NA
e
R
Trst acts at centroid of
area of steel in tension
As = total steel area and e is the distance between Rst and Rsc
Centroid of the area in tension – when NA is at the top of the top flange the moment of
the tension force in the steel is fy * As * D/2. As the NA moves into the flange a
compression force is induced and reduces the moment -
D t
As . − B.t1 . 1
t 2 2
From top of the top flange e+ 1 =
2 As − B.t1
As .D − B.t1
2
t
e= − 1
Then 2.( As − B.t1 ) 2
As = 6890 mm2 D = 403 mm
B= 178 mm t1 = 9.1 mm
Then e = 2776700 - 14740 - 4.6

2 (6890 - 1620)
= 267.3 – 4.6 = 262.7 mm
t1 a
AND e' = e + + t0 − where a is the depth of the concrete
2 2
compression block
Therefore e’ = 262.7 + 4.6 + 150 – 11.9 = 405 mm
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Moment of Resistance of the combined section is given by
Mrc = Rsc . e + Rcc * e’ (Taking moments about Rst)
= 488 * 0.257 + 1091 * 0.405
Therefore Mrc (Ideal) = 567 kNm
φ Mrc = 0.85 * 567 = 482 kNm > 424 kNm OK
Could recycle with lower weight steel beam such as 360UB51 (not that much saving) or
360UB45 – if depth of beam is critical could try 360UB of any weight
ACCEPT as OK in Flexural Longterm Composite Action NOTE: in practice, you should

iterate the design to achieve a slab/beam combination which meets the requirement cost
effectively – being more than about 20% is not good engineering
Check Shear – Usually carried on the Web only of the steel beam. This means that the
“in-service” shear design action is the worst case for the beam at all times – that is there
is no need to check shear, web crushing and web buckling during construction as these
loads are substantially less than the in-service design actions
φ Vs = 0.9 * 0.6 * 403 * 7.6 * 300
= 496 kN > V* of 170 kN OK ACCEPT as OK in Shear
Check Web Bearing and Buckling - Should also at this point check local web effects
such as Web buckling and crushing by knowing or estimating the length of bearing at the
beam support and application of sections 5.13 and 5.14 of the Student Standard. This is
required to be done in Assignment 2 – Refer also Section 2 Notes. –
Notes : Check BOTH web crushing and web buckling
• If required stiffen the web using flat bar stiffeners fillet welded (full profile)
between the flanges of the steel beam
• Width of stiffener does not have to cover the entire outstand of the flange
(select standard flat bar widths)
• Thickness of the stiffener needs to be robust as they can be easily
damaged during construction - so while a 5 or 6mm thickness would
theoretically do the job, when dealing with large beams or plate girders
increasing the thickness for durability is sound engineering
• When dealing with the spread of the reaction forces into the web
acknowledge that the end of the beam prevents spread of the load
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SERVICEABILITY CHECKS - During long term and short term service

Refer Student Code 13.1.2.6 – requires that long term serviceability checks need to take
into account creep, shrinkage and interfacial slip. Ponding should also be allowed for,
both in strength and deflection checks where the beams are substantially unpropped
during the wet concrete phase and precamber not considered. In addition, if the steel
beam has been unpropped during construction a check needs to be made of the
maximum stress in the bottom flange of the steel beam during serviceability
In this respect the stipulation to prop or leave unpropped steel beams during
construction of composite steel beams is crucial to the design. Code Cl 13.1.2.3 (c)
suggests that for steel beams propped at two or more places within the span and then
removed after curing, all permanent and imposed loads on the structure can be
considered to be taken by the composite section. In the case of one prop at midspan
the action of removing the prop should be taken into account in the design calculations.
Where no props are provided the design must follow the actual construction sequence
and carry most permanent loads on the bare steel and any further long term and short
term loads carried on the composite section
The calculations for serviceability described below assume (for the bridge) that only
those loads applied after curing are carried by the composite section and form the basis
of engineering judgement on deflections. The initial deflections during construction
under the permanent loads at the time do not form part of this calculation. If indeed (as
is normal practice in building construction) the beams are propped then the permanent
wet concrete loads are considered in the long term in service serviceability check
If beams are unpropped the beam deflects under wet concrete conditions. An estimate
can be made of the deflection under the wet concrete and the following strength and
deflection calculations can allow for an additional UDL about 2/3 of the peak effective
additional ponding load
SERVICEABILITY - During Construction Load combination SLS
- Note these loads are taken for a bridge beam example which are quite different
to building serviceability loads. They are used here to illustrate application of the
composite beam design procedure. During Pouring, the bare steel carries self-
weight, formwork, falsework, wet concrete (150 thickness) – Note - Construction
live load is transient and does not contribute to ponding
- Beam Spans 10 m - Self Weight Beam say ~ 0.54 kN/m

carries 2.5 m of slab Formwork say ~ 0.3 * 2.5 = 0.75 kN/m
Allow for wet concrete + 10% for ponding
24 kN/m2 * 150 * 2.5 * 1.1 = 9.9 kN/m
G + wet Concrete = 0.54 + 0.75 + 9.9 (150 thick) = 11.2 kN/m
E = 200 GPa Is = 188E6 mm4 (410UB54 Steel Beam only)
5𝜔𝜔𝐿𝐿4 5∗11.2∗100004
Then Elastic Deflection ∆ = 384𝐸𝐸𝐼𝐼𝑠𝑠
= 384∗200𝐸𝐸3∗188𝐸𝐸6 = 31𝑚𝑚𝑚𝑚 (1 𝑖𝑖𝑖𝑖 320)
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General acceptance at about 1:300 - eg a 410UB60 would deflect 27 mm (1 in 370), a

460UB67 would deflect 20 mm (1 in 500)
Or alternatively, and in order to minimize ponding, provide precamber to the steel beam.
This is becoming much more common with modern fabrication capability. However there
are some practical limitations as given below:
• Do not precamber hot rolled elements less than 20 mm

• Do not precamber welded girders < 10 mm
• Calculate and round down precamber to nearest 5 mm
• Largest size for cold precambering is 460UB67
• Precambering can be achieved by cutting a UB and rewelding two tees together
Strength - During Construction – Check steel beam on its own …
Check any lateral support required to the steel beam during construction (no propping
over 10 m) – assumed fully laterally supported at each end of the bridge, that is held
vertically and laterally in position but not in rotation. Assume formwork does not provide
any lateral support to the beam and wet concrete is unstable in the lateral direction
Construction Formwork – ~ 30 kg/m2 acts simply supported between beams
Allow say 10% additional thickness (as above) full span to cater for ponding, verify in a
later calculation
Construction Live Load – 1.0 kPa acts simply supported between beams
Beam Spans 10 m - Self Weight ~ 0.54 kN/m

carries 2.5 m of slab
Formwork ~ 0.3 * 2.5 = 0.75 kN/m
Wet Concrete ~ 0.15 * 24 * 2.5 * 1.1 = 9.9 kN/m
Pumped Concrete Construction LL ~ 1.0 * 2.5 = 2.50 kN/m
Design Beam (unpropped) for the following bridge load combination.
1.35DL + 1.49CN) = 1.35 (0.54 + 0.75 + 9.9) + 1.49 * 2.5
= 18.8 kN/m (Ultimate figure)
M* = 18.8 * 102 / 8 = 235 kNm
V* = 94 kN
Analyse 410UB54 - which is compact -
Ms = fy . s = 300 * 1060E3 = 318 kNm
Mb = αm . αs . Ms αm = 1.13 Table 5.6.1

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______________________________________________________________________
Unrestrained over 10 m * 1.4 (kl refer table 5.6.3(2)) = 14 m
410 UB 54 αs = 0.136 (by calculation since Table A1 only goes to 10 m)
φ Mb = 0.9 * 1.13 * 0.136 * 318 = 44 kNm << 235 kNm No Go!
Therefore Beam requires lateral support –
Require strength to be increased ~ 44 → 235
Therefore αs must increase 0.136 → 0.726 – from Table A1 - say restrain at 2.0 m
(equal) centres, Then Le = 2.8 m
So From Chart try at (2 * 1.4) = 2.8 m cs - αs = 0.741
Then φ Mb = 0.9 * 1.13 * 0.741 * 318 = 240 kNm > M* = 235 kNm OK
ACCEPT 410UB54 Restrained at 2 m centres (fifth points) ← or other options including

increase to a larger member size/weight and precamber (Note precamber does not
affect strength of the beam)
Check Shear, Web Local effects such as crushing and buckling? –
It is normally unnecessary to calculate the web in shear and the local effects on the web
due to crushing and buckling as the steel beam / web on its own is required to carry all
the in-service loads which are generally well in excess of the loads during construction
Suppression of bottom flange yielding under serviceability conditions -
In order to ensure stresses in the bottom flange during its life stays within the elastic
range a check is done during design to ensure this. This means that if the beams are
unpropped, the bare steel section supports the wet concrete loading which introduces
“locked in” tension stresses into the bottom flange. Once the composite section is fully
effective, the applied load causes increased tension in the bottom flange. Under
maximum serviceability loading, the combined tensile stress in the bottom flange must
not exceed 0.9 fy , in order to keep the beam behavior in the elastic range
Accordingly the Student Standard Clause 13.1.2.6(a) requires that elastically under
combined stresses (during construction with steel only acting , plus during service with
composite section acting) that the bottom steel fibre does not exceed 0.9fy under the
relevant bridge serviceability limit state load combination (Group 1A DL + 1.35IQ)
Determine Wet concrete forces (including ponding) – Moment under G (Steel beam only)
ω = 0.54 + 0.75 + 9.9 = 11.2 kN/m
Mc* = 11.2 * 102 / 8 = 140 kNm

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______________________________________________________________________
Determine loads on the Composite section in service M*Q that is the additional
superimposed permanent loads and the short term imposed loads (ΨS = 1.0)
ωQ = 5.5 * 2.5 * 1.35 * 1.3 = 24.1 kN/m (Includes SLS Load factors)
M*Q = 24.1 * 102 / 8 = 302 kNm
Combined Tension stresses in the bottom flange given by
M*c + M*Q = 140E6 + 302E6

zx zb 933E3 1557E3
= 150 + 194 = 344 MPa >> 270 MPa NG!
Allowable – 0.9 fy = 270 MPa No go! Require to go to a heavier or next section up – say
410 UB 60 or better still a 460UB67 for little additional weight
In continuous beams at negative moment regions, this criteria needs to be checked for
the yielding of tension reinforcement only in the slab – refer Student Standard 13.1.2.6
(b)
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TRANSFORMED SECTION SERVICEABILITY under the loaded conditions applied

after curing – Transformed into elastic steel properties
During Construction – Note Ec = 25084MPa (see Page 12) and b1 = 1800mm (see Page
14)
Section Transformed Dist from Ay Ay’2 Ilocal

Area A top of slab *E3 *E6 *E6
Steel Units y where y’ = ybar - y
Concrete 33750 75 2531 74.6 63
Steel 6890 351.5 2422 364.5 188

_____________________________________________________
Sum Totals 40640 4953 251
Therefore ybar = 4953E3 / 40640 = 122 mm < 150 mm
Neutral Axis within the Concrete Section
It = ∑ I local + ∑ A . y’2 (where y’ = 122 – 75 and 351.5 – 122)
= 251E6 + 33750 * 472 + 6890 * 2802
= (251 + 74.6 + 345) E6 mm4 = 671E6 mm4
Zb (section modulus for the bottom fibre of the steel beam)
= 671E6 / (403 + 150 – 122) = 1557E3 mm3
DEFLECTION ESTIMATES – Serviceability requirements Section C13.1.2.6
Construction – Wet Concrete - ∆c = 37 mm – assumed taken up with precamber or

acceptable (with ponding) at time of curing
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Deflection due to concrete shrinkage – induces a uniform sagging moment

throughout the span creating a positive downwards deflection
Concrete Slab shrinks P

Sagging Moment is a
uniform moment which
ebar induces sagging
Centroid of Composite
deflection
Section
Beam experiences sagging moment
∆L P ∆L
Hooke’s Law - = OR P= EA
L EA L
Force P at centroid of Concrete acts ebar above centroid of composite section inducing a
moment Msh on the combined section
Msh = ebar . Є . Ec . Ac where Ac is actual concrete area b1 * tc
ebar = 122 – 75 = 47 mm Line of action of the shrinkage force through centre of the
slab
Shrinkage strain generally allow = 300E-6 ( = ∆L / L = P / EA – Hookes Law)
∆sh = Msh . L2 = ebar . Є . Ac . L2 where n = Es / Ec ……….Equn C13.1.2 (2)

8 Es . It 8 . n . It and Ec is the short term value
Then ∆sh = 47 * 300E-6 * 150 * 1800 * 10,0002

8 * 7.97 * 671E6
= 9 mm ( Insignificant ! 1 in 1100 )
Deflections under additional permanent loads applied after curing and Normal Live Load
(UDL and/or Point Loads) – SLS Group1A G + 1.35IQ
Slip which occurs at the steel/concrete interface reduces the stiffness of a composite
members and this is based on the % of Partial Shear Connection
In this instance – % of Partial Shear = 1034 / 2067 = 50%
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______________________________________________________________________
In lieu of tests – Standard provides a formula for effective stiffness Ie of
Ie = Is + 0.85 (p)0.25 (It – Is) ……………Eqn C13.1.2 (1)
P = 1 for full 100% shear connection
In this instance - Ie = 188E6 + 0.85 (0.50)0.25 . (671 – 188)E6
Ie = 533E6 mm4
All serviceability deflections post curing are based on this effective second moment of
area of the composite section. All long term loads such as permanent loads applied
after curing and long term imposed loads (as required in the loadings code for say
buildings) attract a further 15% of the elastic deflections to be added for creep effects
(Refer Student Code Commentary C13.1.2.6 page 175). For bridges the imposed loads
are assumed transient and so creep effects are not applied here
Check LL Deflection – LL Q ~ 5.5 kN/m2 * 2.5 * 1.35 * 1.3 (from Serviceability Table of
Load Combinations) OR Point Serviceability Load Deflections if point loads occur
= 24.10 kN/m
Then if only UDL is applicable - ∆LL ~ = 5 24.1 * 10,0004

384 200E3 * 533E6
= 29 mm or 1 in 340 OK
ACCEPT 1 IN 300 UNDER SHORT TERM LIVE LOAD.
If truck axle or point loads occur on the bridge then different serviceability formulae are
required. Use SAP or hand calculations for deflections under the various point loads
under consideration with the point loads arranged for the worst action adopting the
effective stiffness shown above
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CIVIL 713

SEMESTER 1 – 2020
SECTION 5 – Checking Structures
5.1 Checking structures techniques

5.1.1 Moving loads
5.1.2 General items
5.1.3 Structural steel
5.1.4 Reinforced concrete
5.2 Checking structures examples

5.2.1 Reinforced concrete
5.2.2 Structural steel
5.3 Checking design
5.4 Checking calculations
5.5 Checking drawings and documentation
5.6 Structural assessment and recognition
5.7 Special note on examination and assessment technique

for application in this course
5.8 Final advice and conclusion
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CIVIL 713 – 2020: Section 5 Checking Structures and Moving Loads
CIVIL 713 - STRUCTURES & DESIGN 4

Semester 1 2020 – Checking Structures
5.1 Checking Structures Techniques –
Checking Structures is taught so that the student develops a critical eye to

drawings and designs which they may come across in their design life and also
are aware of their own designs and the drawings which may be created from
them – they may not reflect what you want them to be and you should be skilled
at looking for the problems which can arise more often than not while the job is
under construction
Structural Checking is different to techniques used in design and is a quick

assessment of an existing situation (dealing with an existing construction being
used for another purpose in which very little can be done about what you have in
place) or a previous design which hasn’t been constructed yet (in which
opportunities can be made to rectify errors in design or the drawing before it gets
constructed). Both skills are very much needed in the current climate of
engineers often not being given enough time to do a proper job, or building
occupiers wanting assessments done on parts or all of a building for various
purposes
The structural checking process is a first up, preliminary look at capacity of a

structure to withstand code loadings, to check that the drawings either reflect
what the engineer has designed or what has been constructed and to advise a
client or a colleague within your Firm, which areas may be of concern and require
further attention. The process is NOT meant to detail nor determine an
innovative solution or strengthening measures needed – this is for a future
exercise when the client gives you authority to proceed to the next more costly
approach. However in an assignment or examination, suggestions may be
sought as to what you may do as remedial work without actual design and
detailing
Different scenarios exist –
a) Does this design (still to be constructed) meet fully the requirements of the
current building code and design standards? In this instance the checker
requires to independently check the calculations and drawings to ensure that all
elements of the Design standards are being met
b) An existing building is likely to be used for a different purpose or being modified
where the Council is requiring upgrading to current standards. Does this design
(as constructed) meet current design standards (or modified standards as
conceded by discussion with Council)? In this instance the checker needs to
obtain as-built drawings, satisfy themselves (as best can) whether or not the
structure was constructed in accordance with the design, and then to structurally
check each item against the current (or modified) design standards
Section 5 - Page 1
c) An existing building is to be purchased and the future Owner wishes to have an

evaluation carried out on the structure
d) An existing structure is being asked to carry a one-off operation (such as a crane
lift or a machinery shift) where a nonstandard load is being applied to part of the
structure. In this instance the checker needs to obtain as-built drawings, satisfy
themselves (as best can) whether or not the structure was constructed in
accordance with the design, and then to structurally check each item against the
one off operation pointing out areas which may be inadequate and eventually
advising the client what remedial measures can be taken to allow the operation
to occur. One option of course is to tell the client it can’t be done without
substantial and costly reconstruction which may not be viable as other options
could be explored (look outside the square!) – You can’t do the impossible!!
The techniques generally deal with –
• Being rigorous in reading a drawing and inferring those things which can
arise which affect the strength or stability of the structure. Many of these
things do not arise from calculations but from bad drafting, poor detailing,
improper construction or misunderstanding of where the load path travels
• Understanding that in many instances an existing structure cannot be
readily and cost effectively modified. For example in an existing reinforced
concrete beam you can’t reduce the stirrup spacing, increase the size of
the tension reinforcement or increase the thickness of an embedded plate.
The engineer needs to adopt other methods to strengthen using external
bolt or glued on features
• Dealing with the strength of an item of structure and following the load
path through – that is, it is not good enough to find that one part of a load
path is not able to take the load therefore ignore the rest. It is expected
that the checker thoroughly deals with each item of load path to ensure
which items are OK and which need attention or strengthening measures
to bring the whole structure up to the standard required
• Don’t be impatient about suggesting remedies as you proceed through the
load path. Wait until the whole structure is assessed and then spend time
on the items of most concern. For example it is trivial to worry about the
thickness of a cleat at the end of an overloaded beam when the
inadequacy of the main column or beam requires major reconstruction
• Inspecting existing elements of a structure for agreement with the intent of
the original designer, noting deterioration in-service (by corrosion, poor
original construction, or wear or abuse) and taking this into account during
the assessment process
• A summary is often provided giving percentages under or over strength of
each part of the structure in order to enable the reader or client to
appreciate the scale of a problem and the areas of most concern.
• Your analysis may also find that the current structure is inadequate to
carry current design actions. This should attract an immediate discussion
Section 5 - Page 2
with your supervisor, a thorough check of your assessment and then when
proven, an immediate discussion with the current owner or occupier who,
based on your investigation (which has discovered a health and safety
issue), implies they are operating an unsafe structure which requires their
immediate attention
• Serviceability does not usually come into checking structures and while
these skills are not taught if a structure is sensitive to deflections and/or
the proposed use which is important to the user of the structure then
serviceability should be checked at some time. Initially we look at strength
only first and then proceed from there
• Techniques also, in the first instance, make broad assumptions usually on
the conservative side of load paths and strengths – it is not expected that
a checker spends weeks determining rigorous analyses in order to get a
structural assessment – using common sense and engineering judgement
an engineer should be able to ascertain those items which clearly are
borderline quite quickly and then if the project is to proceed can then start
looking closer to those aspects of the structure
• It is a common practice amongst some students to divide the M* and/or V*
by φ, and then check the ULS capacity. On the expectation that some
values of φ change with respect to the design element being undertaken
(weld category, bolt in shear, bearing, corbels etc) the student needs to be
cautious that the correct φ is being applied. It is much safer to identify M*,
R* or V* and allow that to remain as such, and to apply the correct
strength reduction factor φ appropriately when that part of the structure is
being designed. That is compare φMb with M* not Mb with M*/φ
• At the end of the day you as engineers must be confident in advising a
client that the structure is sound. If there is doubt either say so and explain
the risks or better still (in order to remove doubt) recommend modifications
which are sure and measurable
• In conclusions and recommendations be careful that you do not take onto
yourself decisions that should rightly be made by the Client or pass onto
the Client the decision of the level of risk they should or shouldn’t take
o Do not surmise what is too expensive – it is the Client’s decision to
spend or not spend the Company’s money – you as the engineer
may provide them with options and cost estimates – let them
decide for themselves what to do
o Do not pass your concerns or fears onto the Client – they will not
understand the level of risk they should or shouldn’t take. You as
the Engineer must advise facts and alternative actions which in
your opinion meet design standards – they have the prerogative to
ignore your advice
5.1.1 Moving Loads
When designing a structural system for moving loads, care must be taken to
ensure the critical design actions are determined. This may involve running
Section 5 - Page 3
sufficient load cases until the critical one is obtained. This is a very large topic in
itself and for this course, coverage is confined to some general recommendations
on how to deal with moving loads:
1. Know the positional limits associated with a moving load. For example, in
roading network bridges, loads are applied in specified lanes as determined
by the Transit NZ Bridge Manual. This determines how the load is shared
between supporting members.
2. Know the contact areas and dimensional layouts of a moving load. All moving
loads will have specified contact areas over which the load is applied to the
supporting surface. For example see the Structural Design Actions 1170
Student Standard Clause A3.2 on concentrated loads.
3. On a beam, the maximum shear from a moving load will be determined by
placing this load as close to the end support as possible. If the moving load is
applied in more than one position, ie a set of wheel loads from a design
vehicle loading, then all the loads should be on the beam being designed.
4. On a beam, the maximum bending moment is typically generated by having
the load applied as far away from the supports as practicable. However, if the
moving load consists of a load group with more than one applied load point at
a fixed spacing, applying the centroid of the load group at the centreline of the
beam may NOT give the maximum moment. You need to also check the case
for one load at the centre of the beam and the other(s) at their fixed spacing
point of application. Even those two may not give the maximum moment in
the beam.
5.1.2 General Items –
1. Understand the load path and follow it/them through to ground

a. Example – Hyatt Regency Collapse – where assumptions and
directions given in drawings weren’t followed on site. Without
referral, changes were made creating failure and loss of life
2. Ensure it is clear where lateral restraints are required – it is often the
secondary forces which are not dealt with particularly carefully that can
create failures
3. Know quick methods of analysis, moment distribution methods for
continuous beams and moment area techniques for displacements of
beams under simple load conditions
Section 5 - Page 4
4. A common error by engineers is to assume the shear force on either side

of a support in a continuous beam is the sum of the shear forces on each
side. This is incorrect. The shear force design action either side of an
internal support remains as is and is usually different either side of the
support. The sum of the shear forces is equal to the reaction of the
support (taking full account of the direction of the forces)
5. Checking Design, Calculations and Drawings – Refer to Learning from
Failures Section 4 Pages 3 and 4 for guidance on detailed checking of
these features of design
5.1.3 Structural Steel :
Steel is relatively easy to measure and assess – it is usually visible, usually

the sizes are standard UB or UC sizes which, if necessary, can be checked
against property charts of many years ago. Bolts can be tested, welding can
be inspected and the general condition of the steelwork can be assessed
1. Cleats can be checked for pin or bolt capacities, end and edge
distances, bolt tear out, welding and tension compression. Note
whether or not the applied load is causing unacceptable secondary
forces from eccentricities
Section 5 - Page 5
2. Baseplates, endplates and all connections can be viewed, inspected,

checked to see whether or not “as-built” is the same as “as-drawn” or
intended by the original designer. Note Prying arises in steel on steel
connections where a plate flexes under load, creating very high point
bearing loads which adds to the total load on a bolt as in moment
endplates. In concrete/steel connections it is generally accepted that
the high bearing loads at the point at which the steel plate wants to dig
into the concrete just crushes the concrete and therefore prying is not
an issue
3. Members can be analysed for the design actions taking full account of
actual lateral stability
4. Anchor bolts to concrete can be analysed for the “as-built” condition
although assuming what is drawn, was built, particularly inside a
concrete beam or footing, can be risky. If the risk is too great then
testing to full load may be warranted or removing the risk of guessing
and provide strengthening measures taking zero contribution from the
internal reinforcement or details. Engineers have for decades detailed
anchor bolts or steel assemblies in concrete for lateral load or tension.
Details all vary and each engineer has their special way of transferring
load. Carrying the load path through to conclusion needs to be
carefully understood for example a pair of bolts in tension cast into a
reinforced concrete beam can fail by
a. The threaded part failing
b. Tension in the steel bolts
c. Anchorage into the concrete via development length or physical means such
as bearing plates
d. Connection of bolt to the anchor plate (tearing through)
e. Bearing on the concrete
f. Ability of the concrete cone to sustain the load (punching shear or pull-out)
The technique is to check the capacity of each element in turn and
compare that with the applied design action. All elements should be
checked to find “the weakest link”. If this element can be easily and cost
effectively strengthened then the next element becomes the weakest link
etc
5.1.4 Reinforced Concrete :
Reinforced Concrete can be a problem with checking an existing structure in

ascertaining what is on site versus what is shown on the drawings. Gross
dimensions can be measured – whether or not the reinforcement has been
placed as-drawn is in doubt and some caution must be taken when large loads
and physical safety is at risk – in addition the condition or strength of the
reinforced concrete and grade of reinforcement may not be shown on the
drawings or if shown whether or not the structure was constructed with that
material strength
Section 5 - Page 6
1. Making reinforcement fit is a

problem which should be sorted out in the
design office and NOT on site. The
photos above show good examples of
congested reinforcement and the possibility sometimes of reinforcement
being unable to fit into the beams, columns and beam-column joints.
Some draftspeople are very good at sorting these problems out on the
drawing board BUT it is up to the design engineer to ensure this is done
and thought about when determining the size, shape and spacing of
reinforcement in concrete
2. Old drawings may not tell you the whole story
Section 5 - Page 7
3. Confusion can arise about concrete cover to reinforcement or the grade of

bars used. Even if it states “cover to reinforcement to be 35 mm”, does
that mean to the main reinforcement or to any steel (ie outside of stirrups)
4. Cover of 50 mm to main reinforcement means to the main longitudinal
bars – watch for this – cover to reinforcement can be described in several
ways – “minimum cover to all reinforcement” means this is the minimum
cover to the outside face of all stirrups and bars throughout wherever they
are placed. Often cover is given (like maybe in your assignment) to the
main reinforcement which allows the stirrups to enter in to the cover
concrete
5. Development lengths of reinforcement are an issue which is often ignored
or considered as an afterthought – moreover it is hazardous to assume
the beam, slab or column has been properly constructed in accordance
with the drawings (however we have to start from somewhere). Students
need to understand the significance of development length in the moment
carrying capacity of a beam or column and to check accordingly. While
development or lap lengths may be detailed as quite short, at that
particular point in the structure the requirement for 100% development
may not be needed and therefore may be OK or acceptable
6. Often lap lengths have not been dimensioned, sometimes laps aren’t
shown at all allowing the contractor to decide for themselves where laps
can occur (not good practice!!). Discuss random lapping.
7. Before the latter part of the 20th century “high yield reinforcement” was
generally non-ductile and achieved an fy of 380 MPa. Mild steel fy of 275
MPa (cf 500 MPa and 300 MPa in current standards) so be careful when
checking older structures. If it is critical carry out mechanical property
tests on the materials by sampling
8. Know how to convert imperial units to metric and don’t make a
fundamental error!! – 12” can be converted to 305 mm, ¾” bar can be
converted to 19 mm dia, 3000 psi can be converted to 20 MPa
9. Given that you believe the drawings or can assess the amount of
reinforcement present, simple analyses can obtain capacities of the
section in flexure and shear
10. Connections of steel to concrete can be thoroughly checked as 5.1.2 item
4 above
11. Potential areas of early failure is in parts of the structure which can attract
congestion in steel eg corbels, beam column joints, half joints and
anywhere where water has collected or flowed and corrosion can have
started or be well advanced. Be aware that reinforcement because of
congestion on site can in reality be cast in quite different positions thereby
reducing considerably the capacity of the section
12. If you are checking someone else’s drawing ensure you understand the
concept of congestion and how difficult it is for a fabricator to create a
cage from a myriad of lines on paper. Many heavily reinforced joints
should be looked at in three dimensions on the drawing board
Section 5 - Page 8
13. Punching shear or pull-out is an issue in slabs, sometimes walls and in all
cases where very high loads are concentrated in one place. If you come
across a structure which has very high loads impinging on or within
reinforced concrete be very wary. In addition if you have specified laps
and/or hooks, ensure there is room to fit these in. When lapping bars it is
preferable not to have the laps all in the one place
14. Note with existing reinforced concrete structure you can’t modify details
which are embedded in concrete, eg size or spacing of existing stirrups,
development length or number of tension bars – it is impossible without
destroying the existing structure
5.2 Checking Structures – Examples

5.2.1 Reinforced Concrete – The NZ Standard 3101 has many
requirements for the design and detailing of reinforced concrete. Items such as
durability, concrete cover, guidelines for edge distances, bolt centres, stirrup
centres, minimum reinforcement and buckling of compression bars during a
seismic event are all covered substantially in the standard and new designs are
required to conform with these. In many instances existing (and particularly old)
structures may not conform with these requirements. However not complying
with these does not mean the structure may not be structurally adequate to carry
a one off load or be used for a different purpose for a short time. Addressing
these issues is not part of checking structures procedures and can be left to a
further study if demand or the particular problem requires it
5.2.1.1 Strength in Flexure (Beams or columns) –

Without axial load the calculation is simply checking the capacity of the tension
reinforcement whatever it may be to carry an applied ULS moment at the section,
checking, of course, that the reinforcement location and anchorage can develop
the full strength of the bar
Given As, b, d, f’c, fy or values assumed as conservative for this instance –
As f y
a= (depth of compression block)
0.85 f c'b
a
φM n = 0.85 As f y (d − )
2
For example, Determine the flexural capacity of a 750 deep by 350 wide reinforced
concrete simply supported beam in a warehouse built in 1958 by referring to a drawing
showing 4 – 1” diameter deformed mild steel bars in a single line (no cover shown) at
the bottom of the beam
Assume f c' ~ 20 MPa (old 3000 psi) , f y = 275MPa (old strength)
Section 5 - Page 9
As = 1810 mm2 (M24 bar) Cover say 40 mm (conservative)
Then d = 698 mm to main reinforcement (one layer)
As f y 1810 x 275
=a = '
= 84mm
0.85 f c b 0.85 x 20 x350
a
ϕ M n 0.85 As f y (=
= d − ) 0.85 x1810 x 275 x(698 =
− 84 ) 278kNm
2
2
Report that provided the beam was constructed in accordance with the old drawings and
(using engineering judgement) taking full account of conservative values for strengths of
materials the expected ULS of the beam in flexure is 278 kNm OR a percentage of the
required ULS strength. Not that this means anything to a non technical Client. It is
better to compare this with the expected M* and advise whether or not the structure is
under or over strength to carry the duty
If axial load in compression is present (and the designer is confident it is present) then
this force enhances the moment capacity of the section. In checking structures techniques
in the form of a first “take” the axial compression force could (conservatively) be
ignored. A good rule would be to check that the axial ULS compression load does not
exceed say 5 or 10% of 0.85 * 0.85 * f’c * A. If the applied load is smaller than this it
can be ignored in the above calculation as its effect has little difference to the moment
capacity. If the axial load is necessary to make the section work a more rigorous analysis
can be carried out which enhances the moment capacity. However, designers must be
satisfied that the axial force being assumed in the design does actually occur
If there exists an axial load in tension then this reduces the capacity of the section to carry
moment and must be taken into account in the moment calculation by …
Assume that the reinforcement yields (ignore compression steel)
As f y − N nt
a=
αf c'b
 a h a
M n = As f y  d −  − N nt  − 
 2 2 2
where Nnt = ULS tension force
For the above problem if an axial tension load of 135 kN was present in the beam, using
these equations a → 61mm and ϕMn → 242 kNm
Section 5 - Page 10
5.2.1.2 Strength in Shear (Beams) – Section 9 of the Concrete Standard
Under normal circumstances (not in a plastic hinge region) designers can carry
shear on the concrete and once the capacity is exceeded can carry the balance
of shear forces on any stirrup or diagonal bar reinforcement present. In plastic
hinge regions ALL shear must be carried by reinforcement with no shear being
carried by the concrete
Given Av (area of total number of stirrup legs across a cross section) and s
(spacing of stirrups), normal aggregate concrete (ka = 1) and d < 400 mm (kd = 1)
vc (min) = 0.08 f c' (conservative) or vc = (0.07 + 10 ρ w ) f c'
Av f y
vn = vc + . Then φVn = 0.75.vn .bw .d
s bw
For example, say our beam in the previous example shows 2 leg 3/8” dia stirrups at 12”
centres – 1” = 25.4mm, then 3/8” = 9.5mm, 12” = 305mm. Generally assume all stirrup
steel unless shown otherwise is mild steel ~ fy = 275 MPa
Av = 9.52/4 * pi * 2 = 143 mm2, s = 305 mm, fy = 275 MPa
vc (min) = 0.08 20 = 0.36MPa or Equn 9-4

10.1810
10 ρ w = 0.074
=
698.350
= =
vb 0.64 =
and kd (400 / 698)0.25 0.87
Therefore vc = 0.57
Av f y 143 275
= . = . 0.368Mpa Then vn =0.57 + 0.368 =0.94 MPa and
s bw 305 350
=ϕVn 0.75
= x0.94 x350 x698 172kN
Taking no account of the main Tension reinforcement
= =
vc 0.36 MPa and ϕVn 0.75
= x0.73 x350 x698 134kN
Can now compare that with the actual design actions occurring on the beam and report
that provided the beam was constructed in accordance with the old drawings and (using
engineering judgement) taking full account of conservative values for strengths of
materials the expected ultimate limit state of the beam in shear is 134 kN or taking
account of the main tension reinforcement could rise to 172 kN
Section 5 - Page 11
5.2.1.3 Punching Shear in reinforced concrete two-way suspended slabs –

Section 12.7.1 (b) of the Concrete Standard
NOTE: that 12.7.1 (a) provides for the slab acting as a beam (for example when
a slab sits atop a column and the reaction is punching upwards through the slab
– the sketch below but upside down) – refer further information ex Park &
Gamble in 1982. Designers should determine the strength governed by the more
severe of the two conditions. These notes address the two-way punching shear
action described in para (b) only
Whenever large forces occur over small areas on a suspended structure very
careful consideration should be given to the likelihood of the slab or structure
failing in local punching shear – effectively like a stiletto heel punching a cone of
concrete or hotmix seal through a thin layer. The model normally assumed is a
45o cone created by the applied load over the area of the footprint and the
concrete fails in diagonal tension across the 45o shear plane see below
This model (conservatively)

ignores the contribution made
by the horizontal reinforcement
in the slab
The perimeter of the shear

plane termed bo is calculated at
a distance d/2 from the footprint
forming a rectangle in plan. If
the applied load is circular the
normal practice is to convert this to a rectangle of the same area
For the example shown above if the load is applied through a 150 mm SHS
(square hollow section) and clear cover to the main bars is 25 mm, d = 184mm
The perimeter bo would be (75+92) * 2 * 4 = 1336 mm
If the SHS was a 230 mm dia CHS (circular hollow section) the perimeter would
be a circle with a diameter 414 mm and the equivalent square in this instance
(equating areas) is 367 mm square – Perimeter = 1468 mm
Given an ultimate limit state axial load of 270 kN acting on the slab through a 150
mm SHS (rigid baseplate) determine the capacity of the slab to sustain the
punching shear assuming 28 day strength of concrete is 30 MPa Ref 12.7.3.2
𝑘𝑘𝑑𝑑𝑑𝑑 = �200�184 = 1.04 > 1.0 Adopt 𝑘𝑘𝑑𝑑𝑑𝑑 = 1.0 𝛽𝛽𝑐𝑐 = 1.0
1 2
Then 6
�1 + 𝛽𝛽 � = 0.5 > 0.333 Equn 12-6 so use Equn 12-8
𝑐𝑐
Section 5 - Page 12
φVn = φ .vc .bo .d where φ = 0.75 and vc (min) = 0.33 f c'
d = 184 mm and bo = 1336 mm νc = 0.33*1.0*√30 = 1.81 MPa
Then φVn = 0.75 * 1.81 * 1336 * 184 = 334 kN > 270 kN OK!!
No special reinforcement necessary
If this calculation fails the designer needs to reduce the point load, or increase the
footprint area, or if this is a new structure, increase the depth of the slab, or put in special
shear reinforcement
5.2.1.4 Pull out in Beams (a form of Punching Shear) – Anchors in a wall in

tension or in a beam (horizontal forces from a moment connection (push –
pull)
For anchors in a wall or, say. Buried anchors in a beam the designer needs to
check:-
• Note that the shear forces may be different on each side of the tensioned
member – you need to look at the geometry of the applied action
• Bolt in tension (dealt with as steel member or threaded member in
tension)
• Capacity of the back plate (thickness/size)
• Bearing on the concrete under the plate (dealt with in 5.2.1.5 below)
• Punching shear through the wall carried by the concrete only as a 45o
cone of failure (dealt with in 5.2.1.3
• Punching shear through the beam as a 45o cone of failure but prevented
from failure by tension in any stirrups available within the cone of failure.
This assumes zero shear carried by the concrete (conservative)
Section 5 - Page 13
In a beam such as shown in the sketch above the failure plain is very localised.
Engineers generally ignore any punching shear carried by the concrete and
assume the cone of failure is prevented from occurring by calling on those
stirrups intersecting the cone to carry the load in tension. The line of failure is
sketched to scale at 45o from the plate (as shown). NOTE: this assumes the
shear is equal both sides of the tension member – be careful to understand this
should be considered in terms of the shear span. Not knowing where exactly the
stirrups are placed but knowing or having an expectation of their spacing the
stirrups can be adjudged back and forth to show a conservative number of
stirrups intersecting the 45o shear plane (in this instance three sets of two leg
stirrups as a possible worst case). A stirrup which intersects the bearing plate
itself should not be considered as acting. A comparison is then made of the ULS
capacity in tension of the six legs of stirrups versus the ULS applied load on the
plate (see example after 5.2.1.5)
5.2.1.5 Bearing on reinforced concrete – Section 16.3.1 of the Concrete

Standard
φN b = φ * 0.85 f c' * A1 where φ = 0.65 and A1 is the area of the

applied footprint
For a slab or a beam or surface wider than the bearing plate some spread of the load
footprint (2h : 1v) is allowed (to a given maximum), but be careful of edges or depth of
slab which limits the spread. The “spread” area is designated A2 and the allowable
A2
bearing strength can be increased by the ratio (maximum value of 2.0)
A1
Examples of situations where the edges may affect the area A2 or where bearing plates
impinge on each other will be given in class
Section 5 - Page 14
Given the 28 day strength of concrete for the above beam and slab is 25 MPa,
determine the maximum ultimate bearing capacity under the bearing plates as
shown – bearing plate sizes are 250 by 150 in the first instance (beam) and 150
by 100 in the second (slab) – say ULS N* = 245kN for both cases
Beam - Plate is 250 across the beam and 150 along the length of the beam -
Allowable spread is at 2:1 – across the beam the load spreads only to 350 wide
(confined by the edges of the beam which is 350 wide) that is only 25 mm deep
→ 50 each side + 250 = 350
Given that restriction the spread along the beam can only be 25 deep or 50
increase in width both edges, spread is therefore 150 + 2 * 50 = 250
A2 = 350 by 250 and A1 = 250 by 150
A2 87500
= = 1.53 < 2.0 OK – Then using A1 for the area and the
A1 37500
multiplier -
φN b = 0.65 x0.85 x 25 x 250 x150 x1.53 = 792kN > N * = 245kN
A2
Two way Slab – Plate is 150 by 100 - Spread in both directions 2:1 such that = 2.0
A1
Then φN b = 0.65 x0.85 x 25 x150 x100 x 2.0 = 398kN > N * = 245kN
NOTES: -
• The 2:1 spread suggested in the Standard is given assuming that sound concrete is close
to the bearing plate. This spread must not be confused with the general acceptance of a
45o spread of the load as it passes through the concrete
• The basic equation provides the minimum requirements, higher loads can be accepted if
confinement of reinforcement is taken into account (refer NZS3101)
• Bearing is one method of failure of a high concentrated load on the surface of or buried
within concrete and must not be confused by the other mode of failure which is to
punch out a shear cone of concrete as covered in 5.2.1.4 above
5.2.1.6 Example
Showing a set of calculations for a buried plate sustaining an ULS force. The
system needs to be checked for various possible modes of failure.
Section 5 - Page 15
Note that this is only part of an exercise where in this instance the entire beam
may have to be checked with regard to flexural and shear capacity, and end
supports
Require to check the capacity of
a single M24 Grade 4.6 (mild
steel) threaded rod in tension
buried within a concrete beam
(400 by 250) anchored by an 80
by 80 by 10 plate underneath the
top reinforcement. Assume
reinforcement is 4HD20 main
bars top and bottom and R10
two leg stirrups at 180 mm cs.
Assume cover to main bars is 40
mm, fut of mild steel is 400MPa,
fy is 300 MPa
A. Check threaded rod in tension – Tensile Area of a threaded M24 Grade

4.6 threaded rod is 353 mm2
ϕ N t = ϕ . As . f ut
ϕ N t = 0.8 x353 x 400
= 113kN > N=
* 110kN
B. Check Plate in bending – (conservative as plate is held rigidly by

concrete both faces). Assume plate bends in one direction as shown next
page, with the bent edge failing at the edge of the nut. Check N*
Pressure on the plate ~

110000 N
= 18.74 MPa = 18740kN / m 2
π .26 2
(80 x80) −
4
The width of the nut is 36 mm – assume bending of the plate beyond the nut
resisted by bending by the full width of the plate
P’ acting over 22 mm by 80 mm = 33 kN
Then M* = 33 x 0.011 (over 80 width) = 0.36 kNm
Section 5 - Page 16
of plate ~ ϕ . f y .z 0.9
Capacity = = x300 x80 x102 / 4 0.54kNm > 0.36 kNm OK!!
A 10 mm plate is acceptable under this model…at this stage this plate can sustain ULS
load of 113 kN > N* = 110 kN
C. Check Bearing – Maximum width of 2:1 spread through beam is 250 mm
A2 250 2
Then A2 = 2502 and = = 3.13 > 2
Aa 80 2
Adopt multiplier of 2
π .26 2
Now - A1 = (80 x80) − = 6400 − 531 = 5870mm 2
4
Then φVb = 0.65 x0.85 x 20 x5870 x 2 = 130kN > 110kN
Connection OK in bearing
D. Check punching shear (or ability of stirrups to achieve strength in punching

shear) – say calculations by others show the existing stirrups at the position of the
lifting eye are up to 40% utilized under 1.2G and 1.5Q
Drawing the elevation of the beam to scale it appears that at least four stirrups may
intersect the 45o cone of spread accepting that the whole cone needs to fracture before
failure occurs. In this instance let’s try three stirrups being conservative…
The residual (60%) capacity of three by two leg R10 stirrups in tension is given by
102
ϕ N t ϕ . f=
= y . Ast 60% of 0.85 x300 x6 xπ x= 72kN < 110kN
4
Section 5 - Page 17
Summary Bolt capacity 113 kN

Plate Bending Capacity 127 kN
Bearing Capacity 130 kN
Punching Shear 72 kN
Based on as-built drawing Ultimate

Capacity of Bolt Connection is 72 kN < N*
= 110kN
Other items may be more critical but this

connection can’t be relied upon to carry the
duty specified
Note : If two or more are attached through a

buried plate then take the total N* across all
bolts and apply it to the plate. In sketch or
drawing terms the description of a group of
bolts may be described, say, as 2 X 2 M16 4.6/S bolts. This implies two rows of two
bolts making a total of 4 bolts in the connection. The elevation may show only two bolts
but it is a group of four acting
5.2.2 Structural Steel –

5.2.2.1 Strength of Steel Cleats in shear / tension –
Items to consider when checking Cleats
1. Bearing on the cleat from the bolt – refer Cl 9.3.2.4.1

2. Tear-out (vertical and horizontal) – this refers to Eq 9.3.2.4 (2) for edge
distances
3. Tension in the cleat – various options
4. Moment on the cleat
5. Weld in shear and carrying moment (assume as push-pull forces
conservatively)
Note this ignores fastening bolt itself whether in single or in double shear
Example of a weld check – Assuming a 60 mm deep by 75 mm wide (30 edge

distance) by 10 mm thickness cleat supporting a single bolted connection
required to transfer an Ultimate vertical upwards force of 18 kN. Check the
capacity of the 5mm GP fillet weld both sides of the cleat
Section 5 - Page 18
Approximate methods as preliminary check – Assume 20mm of weld top and

bottom carries the moment in push/pull and the remaining 20 mm carries the
vertical force in shear
Applied Moment to weld = 18 x 45mm = 0.81 kNm
Length of weld available in 20 mm = 40 mm (both sides fillet weld)
Capacity of 5 mm (GP 0.6) fillet weld is 0.522 kN/mm
The capacity of 40 mm of weld = 40 x 0.522 = 20.9 kN

Distance between centres of 20 mm weld is 40 mm
Then capacity of weld group to sustain moment under this model is given by
20.9 x 0.04 = 0.835 kNm > 0.81 kNm OK
Shear Force in remaining 20 mm given by 20 x 2 x 0.522 = 20.9 kN > 18 kN OK
Further checks available using normal design criteria are tension in the cleat,
tear-out, bearing on the cleat, capacity of the bolt
5.2.2.2 Baseplates in tension – Simplified methods which are

conservative but quick. Further checks must be made using yield line theory if
quick (conservative) methods do not provide sufficient strength for the duty of the
connection – Refer Charles Clifton’s Notes of CIVIL313 Section 10.3
Applied design action includes an allowance for prying
The plastic moment per unit width of the baseplate is given by
ϕ . f y .t p2 202
m p ϕ M=
= s ( Plate ) = 0.9 x =
300 x 27 kNmm / mm
4 4
Section 5 - Page 19
Assume a semi-circular yielding pattern – Leff - πm per bolt where m is

(conservatively) the distance from the web to the centreline of the bolt (Table 2.4
HERA Report R4-142, which is included in the section on connections)
Given m = 55mm , Maximum allowable ULS design axial force from tee stub
yielding is given by
mp
=
Pr = 2=
Leff where Leff (two bolts) π m 345mm
m
Then Pr = 27 . 345 / 55 = 170 kN > 135 kN OK
Check strength of weld – say 130 mm of fillet weld in tension each side of the
web - Say 5 mm GP E41XX fillet weld – Capacity 0.522 kN/mm
For 130 mm length – Capacity in tension = 0.522 x 260 = 136 kN > 100 kN OK
5.2.2.3 Braced Frames – Getting design actions right -
Despite the compression

diagonal being there it cannot
come into compression
(extremely weak in compression)
and therefore carries very little, if
any, load – safest to assume
zero and carry entire horizontal
load with the tension diagonal
Items which can be checked –
• 150UB in compression
• Rod in tension
• Weld - rod to cleat
• 200 by 50 cleat in tension
• Bolt bearing on single
cleat / tear out
• M20 Bolt in double shear
• Double cleats in combined
actions
• Weld - cleats to 150 UB
• Uplift of baseplate (Weld,
plate)
• Anchor bolts in baseplate
Section 5 - Page 20
5.3 Checking Design
Here are some pointers on checking a design.

• Requires overview, looking at the whole picture, experience, double checking base
information
• Requires thorough documentation and confirmation by the Client of what is wanted
• Ensuring staff are capable of carrying out the design or at least prepared to discuss
and ask questions
• Ensuring thorough procedures are in place to document changes which can arise
during design so that all parties to the contract become aware of the changes
• Discussing early and agreeing principles of the structure, how loads are transferred to
ground (loads paths) and the basis on which the applied loads are identified
• Errors and failures can occur not when an engineer has got the calculation wrong but
when something has not even been considered – eg is the structure susceptible to
o differential temperature such as solar radiation, thermal movements
o fatigue
o creep and shrinkage
o foundation settlement
o heat of hydration problems in concrete structures
o unusual load paths
• Is a progressive and identified collapse mechanism apparent if one element fails?
What is the consequence of a failure in a structural member or detail
• Are appropriate construction materials being used and specified – well tried
performance records are much better than the claims of the “latest and greatest” from
a salesperson
5.4 Checking Calculations

• As covered in Notes Section 1 your own calculations must be clear, must tell the story
and be totally unambiguous
• Being astute at all times, not assuming anything until thoroughly researched
• Being thorough and double checking computer output for accuracy
• Can check another person’s calculations by going through page by page but this has a
danger in the checker being lulled into the same error the original designer (OD)
made. It is also tedious
• Better to take a fresh look at the design calculations and carry out some approximate
check calculations which should get within 5 or 10% of the actual figures obtained
rigorously by the OD
• Method –
o Check input, base loads, dimensions for accuracy
o Perform an approximate independent structural analysis to assess approximate
design actions
o Check member sizes and reinforcement for both strength and serviceability
o Check connection layouts, design
o Check consistency between drawings and calculations
Section 5 - Page 21
• In practice, these last two bullet points are not commonly carried out – it relies on the
profession to ensure it is and this will be checked as part of the steel project
5.5 Checking Drawings and Documentation

• Do the drawings reflect what the calculations say?
• Is the structure as drawn adequate to satisfy the intent of the engineer?
• Are the drawings complete – that is, is there sufficient information on the plans and
details to enable the contractor to understand all things necessary for the
construction?
• While engineers may sometimes get caught up in the detail and getting things right on
the page, the basics are sometimes forgotten which may be applicable to the whole
job eg what strength of concrete, grade of reinforcement, concrete cover to main
rebar, grade of structural steel, priming and corrosion protection, grade of timber
treatment
• Is there consistency between the specification and the drawings?
• Are all details constructible?
• Are the drawings consistent within themselves? – it has been suggested that
information should only be specified/drawn once – that is if a dimension or size of bolt
is given on one sheet or one detail then any other information on that sheet or on
another sheet referring to that same detail should not repeat the information. This
means that if a change is made then one dimension or size needs changing and
nobody need trawl through the entire set of drawings to change each time that
dimension is shown and change accordingly
• Method (a well-respected procedure ensuring thoroughness) –
o Check completeness – is everything there, clear, concise with no ambiguities
o Score through with a colour say green pen every detail and item which is
correct and been checked
o Score through with say red pen errors discovered and possibly with the correct
figure/detail shown
o When the error is rectified the red is overscored with say blue pen to provide at
a glance a confirmation that that detail has been rectified
5.6 Structural Assessment and Recognition
It is important for the structural checker to be able to assess the validity or otherwise of : -
a) Base information, loads from Client, NZ Standards, other sources, dimensions,

contours, clearances (say doors), openings, column positions, special requirements of
the Client
b) Design actions from BM’s, shear forces, reactions
c) The flow of forces through a detail or through the structure to ground
d) The displacements imposed on a structure
Checklist – In reality this rigour of checking is NOT common in many design offices which is
a pity. Students therefore must be prepared for facing what may be lack of checking
Section 5 - Page 22
procedures in their working environment and to respond as they consider appropriate. I can
only suggest that you as a graduate engineer can influence your supervisors into returning to
a sound checklist for ensuring the quality of structural engineering remains high in your work
A. Base Information
o Trawl through Base information on file to check completeness of brief, clear
instructions from Client on special features of the design, records of visits to
site, geotechnical review, survey and levels, working off latest issue of drawings
B. Applied Loads
o Check permanent loads, unit weights of materials, self weight, superimposed
permanent loads such as fittings, surfacings, computer floors, ceilings, services
o Loads from Plant, machinery (vibration??) and special storage
o Imposed loads ex NZ Standards, vehicle and pedestrian loads
o Transient loads such as wind, snow and seismic
C. Analysis
o Verify the type of analysis is appropriate, concept of load carrying systems,
modeling of joints (fixed versus pinned, continuity or not), flow of forces through
the structure
o Verify applied loads, load combinations
o Independent check on validity of results (BM, SF, Axial Loads, Reactions)
D. Design
o Correct interpretation of analysis and output
o Check analysis of final beam and column sizes
o Flow of forces through a connection, check calculation on connection elements
o Systematic check on serviceability deflections
Example – Bending Moments – For the continuous beam shown below, supported on
pinned foundations, is the shape and values of the bending moment realistic?
Points to raise : -
1. Loads not defined should be sought if possible
Section 5 - Page 23
2. Shape looks OK but BM’s in outside spans could be different depending on value of
applied loads
3. Something is wrong with the values of BM in central span – what governs your
judgement on that?
4. If the reaction at Support B was given as 300 kN would this be realistic? On the
expectation that the external loads are shown as being lower than the internal span,
the figure of 300 kN is not sustainable – why is that?
Example – Flow of Forces in a member – Often it is important to be able to assess the flow
of forces in a member which can help to identify where reinforcement is required –
Consider a beam or column for example subjected to a central high value concentrated force
– this can arise from a prestress anchor (in a beam) or a support to heavy load (in a column)
Example – Flow of Forces around an Opening in a Wall –
Example – Flow of Forces in a structure – Longitudinal windloads on a building
Section 5 - Page 24
Matters to address –
1. Are the design actions calculated for the wind truss derived from wind on one end or both
ends of the building – that is are there one or two braced bays in the structure?
2. Has the designer included frictional drag in their calculations?
3. Does endwall column carry gravity loads from roof AND lateral wind?
4. Is the roof (and wall) bracing designed as a tension structure only or do the opposite
diagonals require designing for compression?
5. Do the struts B, D & F carry restraint loads as well as wind loads?
6. Are the portal columns capable of carrying this additional compression under transverse wind
load
7. Are the portal rafters capable of carrying the additional (tension/compression – truss chord)
forces under this load condition
8. Have the foundations been designed for the additional loads being imposed under this load
condition
5.7 Special Note on Examination and Assignment Technique –
In any problem associated with checking structures you need to demonstrate to

the assessor your breadth of knowledge. If you are asked to check six items
within the structure for capacity procedures don’t do six checks of shear in the
reinforced concrete beam in six different positions – you may only get marked on
one. You need to show your ability to know what to look for and to demonstrate
skills in as many areas as you can show
5.8 Final Advice and Conclusion
The above items are incomplete but provide for you as a professional engineer some incite
into the number of items which can go wrong in the process of communication of “Design” to
a contractor for construction. Many items which require checking are not covered by a Code
Section 5 - Page 25
of Practice and that is where the professional engineer should develop and have experience,
expertise and skills over a technician
• In your early years ask many questions and get alongside the senior engineers, steel
fabricators, reinforcement placers, grader drivers and contractors who have buckets of
experience to share – listen and learn
Above all common sense alongside an astute enquiring mind which looks at the
wider view stands well above following rules and regulations
Section 5 - Page 26
CIVIL 713

SEMESTER 1 – 2020
SECTION 6 – Non Specific Design of Buildings
6.1 Design of Timber Framed Buildings Using NZS3604
6.2 Overview of Light Steel Framed Building Design

to NASH New Zealand Guidelines
This page is blank
CIVIL 713
SECTION 6.1
DESIGN OF TIMBER FRAMED BUILDINGS

USING NZS3604
SEMESTER 1 – 2020
COPYRIGHT WARNING
This set of notes may be used only for the University’s educational purposes. It includes extracts of
copyright works copied under copyright licences. You may not copy or distribute any part of these notes
to any other person. You may not make a further copy for any other purpose. Failure to comply with the
terms of this warning may expose you to legal action for copyright infringement and/or disciplinary
action by the University.
CIVIL 713 – Structures & Design 4
Timber Framed Buildings – 2020
CONTENTS – PAGE No
Section 1 - Introduction & Limitations 1
Section 2 - Durability 3
Section 3 - Definitions 4
Section 4 - Design for Horizontal Loads
4.1 Wind 6
4.2 Earthquake 11
4.3 Satisfying Demand 13
4.4 Subfloor Bracing 17
Section 5 - Design for Vertical Loads
5.1 Foundations and Subfloor 18

5.2 Floors 22
5.3 Walls, Lintels & Plates 26
5.4 Roofs 35
Supporting pages from NZS3604 and other design

documents for section 6
LIMITATION TO THESE NOTES -
It must be understood that the material delivered in these four lectures is only a
very small part of the design (Using NZS3604) of residential and light commercial
buildings. The Standard is extremely comprehensive and covers a wide range of
design procedures for most elements in these buildings. This material is given for the
express purpose of an introduction to the use of the Standard only and is not meant to
be a complete design guide for structures of this type
________________________________NZS 3604 Timber Frame Buildings – 2017
NZS 3604:2011 - Timber Framed Buildings (all references to “the

document” or “this standard” mean to NZS 3604)
Introduction and Limitations on Use of the Standard (Section 1)

including amendments arising from the NZ Building Code in August
2011 (DBH NZBC B1/AS1 Section 3.0 Timber)
NZS 3604, (first published in 1978), was intended to set down procedures for
the non-specific design and construction for buildings categorized under
Importance Level 1 and 2 only thus eliminating or at least minimizing the need
for specific design (the role of the structural engineer) for many buildings. In
2011 and after the Christchurch earthquake events, the Department of
Building and Housing (DBH) made significant changes to NZS3604 which to
date (December 2013) have not been issued as amendments to NZS3604.
It cannot be emphasized enough the importance of recognizing the limitations

expressed in Section 1 of this standard. The student should read carefully
these limitations before embarking on a design using the standard.
Importance Level 1 & 2 buildings are those described in Table 1.1. The
examples cover many family dwellings, commercial, healthcare, educational
and public buildings of a certain size or accommodating a maximum number
of people and of total height not exceeding 10 metres. Some other limitations,
for example, include founding on “good ground”, limited snow loads (not > 2
kPa), limited floor loads (not > 3.0 kPa and /or concentrated loads not > 2.7
kN) and roof slopes not exceeding 60o to the horizontal. The standard does
not cover buildings with operating theatres or heavy equipment or libraries
where floor loads exceed 3.0 kPa. In all of the cases exceeding these
requirements specific design and the engagement of a certified professional
structural engineer is required
Section 3.1 (NZBC B1/AS1:NZS3604 Timber) in the Acceptable Solutions

area of the NZ Building Code (B1 Structure – Pages 23A and 23B refer
Notes) provide recommendations from DBH on changes to the Standard
particularly related to Christchurch Area. Engineers and students carrying out
designs in this region should be very careful to follow these
recommendations, discuss with the local Authority any local requirements
they may need, and ensure their designs comply in all respects with the
myriad of standards now facing structural engineers in their specific and non-
specific designs
The definition of “good ground” (ref p 1-16 of the document, Cl 3.1.14 NZB
B1/AS1, and P17 of these Notes) has come into question particularly following
the Christchurch earthquake where liquefaction and lateral spread was
evident in many areas of damage to dwellings. Substantial discussion is
currently underway in dealing with this issue and clearer definitions may be
necessary to alert “designers” using this document to be wary of what “good
ground” actually means. Refer comments in the Foreword to the document
pages 9 and 10
______________________________________________________________
__
1
NZS 3604
• Adopts grades of timber being SG 6, SG 8 and SG 10 (for dry timber)
and SG 6 (wet) and SG8 (wet) for wet or green conditions. No1
Framing timber comes under the category of SG 6 and the figures
relate to the appropriate VSG or MSG grade defined in NZS3603
• Requires the use of a lower bound modulus of elasticity for members
that do not act as a group of four or more members
In this course the student will be asked to adopt SG 8 as the “standard” grade
used for design of residential or commercial properties governed by this
standard. Under certain circumstances for example in design of an external
deck, external treatment to No 1 framing timber may be an appropriate
selection in which case SG 6 would be used in design. Failure to
acknowledge this may result in an unsafe design. SG 8 is the most readily
available gauged timber grade although in the design office you may have
need or wish to specify the higher grades in which case the design can be
carried out using the higher value properties but CAUTION – ensure you
specify the Grade well and ENSURE you obtain that grade on site!!
Acknowledgement is made to skilled and practical engineers who have been

instrumental in assisting me and the design profession in Timber Engineering.
In particular Dr Richard Hunt initially taught this course in the early part of the
century and it was on the basis of his coverage and original notes that the
course as you see it today is delivered.
Charles Clifton, who compiled these notes for the 2017 issue,
acknowledges the excellent role of Colin Nicholas in creating these notes in
2015
______________________________________________________________
__
2
DURABILITY
Much of New Zealand is near the sea and humid and as a result experiences
considerable durability problems with corrosion of fasteners and steel plated
connections. Generally buildings are required to sustain a 50 year life and as
such require special consideration for metal items which may corrode and
therefore substantially shorten the design life of the structure
Figure 4.2 provides a map of three designated “zones” which are used as the
starter for determining protection requirements for various construction items.
The Standard also gives guidance on cover to reinforcement in masonry,
reinforced concrete and specified strengths for the various corrosion zones.
Recommendations for flashings, bolts, veneer ties and underlay are also
covered.
An excellent Material Compatibility Chart (in previous versions of NZS3604

but now eliminated) is provided which gives clear guidance on ensuring non
compatible materials are not touching nor the chances of water running off
one material to another causing early onset of corrosion
SURFACE WATER & EXTERNAL MOISTURE
In previous issues of NZS3604, a section (Ch 11) was devoted to the building
envelope and provided sketches and details principally drawn in order to
prevent the entry of moisture. Flashing details and specifications for roofing,
wall claddings, vapour barriers were given which along with the current
standards recently published after the “leaky building” problems arose provide
the practitioner with substantial guidance on details required to achieve a
problem free building envelope
These details are now provided in Dept of Building and Housing Compliance
Documents to the NZ Building Code or in NZS 4229
A copy of the Compliance Document E2/AS1 will be dropped onto CECIL for
student use and interest
______________________________________________________________
__
3
DESIGN -
DEFINITIONS –
In order to create a document which could be used by non technical people it

is important to create simple descriptions to describe light or heavy roof, light
or heavy cladding. These need to be defined in order for the user of the
document to adopt correct interpretation of the various design charts.
Gravity Loads on structures are defined as “Permanent” such as self weight,

cladding, roofing, partitions or “Imposed” such as pedestrian or user loads,
traffic, snow, fittings, furniture, shelves cupboards, storage systems and their
contents.
Other “Imposed” loads include the natural loads such as wind and seismic
(earthquake) loads which tend to apply horizontal loads to the structures.
This Standard attempts to allow the user to determine each element of a

timber structure and to have confidence that the selection of a timber element,
size, shape and spacing will carry out the duty for which it is designed. What
is most important is that the user understands the limitations of the code,
understands the basis on which he/she enters the Standard tables and uses
the tables properly in order to achieve the result. Too often assumptions are
made at the early stage (and the building may not comply with the
requirements of the Standard) or the user applies the wrong table or reads the
wrong figure from the table any of which can provide heartache and problems
down the track with local authorities or a disgruntled owner who has a leaky
building, a bouncy floor or is faced with a major addition in cost because
something was under-designed
For determining design for lateral seismic loads (Refer Definitions 1-16 ff) –
Cladding is classified as three types : -
a) Light – having a mass < 30 kg/m2 – Typically

weatherboards
b) Medium – having a mass between 30 & 80 kg/m2 –
Typically stucco cladding or plaster finish
c) Heavy – having a mass > 80 kg/m2 but not exceeding 220
kg/m2 – typically brick and concrete veneers
Roofing materials (roofing and sarking) are classified as two types : -
a) Light – (includes sarking) having a mass < 20 kg/m2 –

typically metal roofing, 6 mm cellulose cement tiles
b) Heavy – (includes sarking) having a mass between 20 &
60 kg/m2 – typically concrete tiles, slate etc
______________________________________________________________
__
4
Gravity Imposed Loads
Nominally 1.5 or 2.0 kPa which in accordance with NZS 1170 covers most
residential and some institutional applications. Charts in Ch 14 are available
for 3.0 kPa floor loadings which extend the Standards use to many
Commercial and Educational applications. Note the restriction on
concentrated loads of not > 2.7kN which prevents for example a timber
garage floor being designed under this standard. If you are at all concerned
check thoroughly the early sections on the limitations of the Standard and
whether or not the application comes under the guidelines.
Take care that applications such as libraries, storage rooms, assembly areas,
worship centres, commercial kitchens, residential garages and plant rooms
attract minimum design imposed loadings in excess of 3.0 kPa or 2.7 kN point
loads and would therefore require specific design by a structural engineer
Snow loads generally up to 0.5 kPa and special cases up to 2.0 kPa are
covered in the Standard (Ch 15)
An example of the way the Standard works is shown on the next few pages
using the example of Floor Joists – Note these are not the notes related to
Floor Joist Design which are on Pages 22 – 25, but are only included to
illustrate the way the complete standard is set up and the various charts
applicable to the different Grades of Timber.NZS3604 is available to view by
students refer CECIL
The first page of Floor Joists give technical descriptions of the limitations and
extent of the guidelines and describes in general terms the layout,
construction and fixings within which the tables are defined
What follows, are easy to apply tables of spans for various spacings for
various sizes and strength grades of timber for particular floor loads. The
charts are colour-coded (yellow SG 8, blue SG 6 Green SG 10). The way in
which the Tables work will be covered later and all tables given in this course
refer to Timber Grade SG 8 but (as shown here by the similar looking tables
but very different in content) when you are referring to the Standard in the
Design Office take particular care to ensure that you are reading off a chart of
the Grade of timber you are specifying. In the standard the SG 6 and SG 10
tables are given in an Appendix to chapter 7 (shown here)
In this course (and probably in the Design Office unless for a very
special reason) use SG 8 timber grades throughout – this is because
this grade of timber is the most readily available, and is therefore, in
most instances, the least cost. Other grades such as SG 10 are usually
unavailable, special runs and reasonably expensive
______________________________________________________________
__
5
DESIGN FOR HORIZONTAL LOADS (SECTION 5 BRACING DESIGN)
Section 5 of the Standard is devoted to the provision of lateral load carrying

elements designed to resist loads imposed from the natural elements such as
wind and earthquake. Depending on the availability of walls suitable for
bracing, within a structure, the standard allows the designer to assess the
severity of the wind and/or earthquake on the various levels of the structure
and to then calculate the strength of elements of bracing systems such as
braced walls, within the structure which resist the loads at each floor level
If walls are not available or are expected to be removed/altered/modified

during the life of the structure then guidance may require to be sought from a
structural engineer as to how to handle the lateral load carrying requirements.
Often specifically designed flat top portal frames in steel or steel bracing may
be necessary to provide the system required to carry the lateral loads
The method of analysis is based on the concept of Bracing Units, the values
of which describe the level of severity of the earthquake or wind and
additionally describes the capacity of a bracing system to resist the imposed
loads
Nominally 20 bracing units equates to 1 kN of lateral load
Bracing Design –
Is required for each floor of the structure and for the subfloor (if one is
present), the effect being to transfer horizontal imposed natural element loads
through the structure and into the ground without damage to the structure.
Note that the “design” of each floor does not mean that you focus your
attention JUST to that floor – using the principle of “following the load path”
the size and shape of the upper floors may govern the requirement for lateral
capacity of the floor in question – eg wind and seismic for varying floor
footprints
Design for Wind –
The level of natural wind on a building structure is dependent on a number of

factors and the Standard addresses each of these in order for the designer to
reach a conclusion as to the “Wind Zone” for the building : -
a) Location – determine the wind region from the map of NZ in Fig 5.1 –
two regions are offered A or W
b) Location – determine whether the building is in a lee zone – if the
building is in one of the hatched areas of Fig 5.1 map, special attention
is given by notation in Table 5.4 – Low wind becomes High etc
c) Location – determine ground roughness from the description given in
Standard 5.2.3 – the Standard offers Urban (typically 3m high houses,
trees, buildings with density greater than 10 such obstructions per ha),
or open (typically rolling country, farmland with very few trees, airfields
or coastal where the wind has little obstruction)
______________________________________________________________
__
6
d) Location – considers the local effects of shelter or shielding from

adjacent permanent buildings all round – at least two rows of similar
sized buildings at the same ground level all round attracts a site
exposure of Sheltered (typically urban areas). Otherwise Exposed
where these obstructions are not present or may be temporary during
the design life
e) Location – the topographic class of the building assesses the effect of
wind as it passes over a hillside, the position of the building to the
crest, whether or not the hill continues or curves over beyond the
building site, and the gradient of the hill. Table 5.2, 5.3 and Fig 5.2
provide clear details of the determination of T1 to T4 the Topographic
Class of the site
Armed with the information determined from above the designer enters Table
5.4 which allows him/her to select a Wind Zone given by the letters L (Low), M
(medium), H (high), VH (very high), EH (extra high), SED where Specific
Design is necessary
Often designers may elect to approach the council who may be able to
indicate what Wind Zone a particular site is, given the address. This may be
helpful but the Designer should verify that information by carrying out a
separate calculation and checking with the Territorial Authority prior to
proceeding. CAUTION – while your calculations may indicate a particular
wind zone it is prudent to check with the Territorial Authority who may elect to
override your assessment (based on local knowledge which you may not
have) and this could save you time and effort by clarifying this at the start
Example – refer Contour Map over – Dwelling is located with long (14.3 m)
faces facing N-S
Area is farmland with few obstructions such as trees > 3 m in height ► Open
Terrain and is NOT in a Lee Zone
Dwelling located in Ponga Road Papakura (no close permanent structures) –

refer to contour map ► Exposed site ►Ground Roughness = Open ►Wind
Region = A
The building is located on an exposed site on the side of an escarpment at an

elevation of 180 m. The valley floor is at an elevation of approx 40 m and the
crest of the hill is at an elevation of 292 m. The total rise from the floor is H =
252 m over a distance of about 3 km (3000 m) so the Gradient is low (1 in 12
or 0.083). The building is located 260 m from the crest and therefore is just in
to the outer zone. Using Table 5.3 the Topographic Zone is T1.
Entering Table 5.4 with A, Open, T1 and Exposed gives Wind Zone H
Checking with Local Authority (say) confirms wind zone H
______________________________________________________________
__
7
Contour Map of Ponga Road Building Site
Bracing Demand for Wind – How to use the Wind Charts
NZS 3604 provides Charts for the determination of bracing requirements

under wind for each level of a building including subfloor (foundation
structures). Tables 5.5, 5.6 and 5.7 give values in bracing units per metre
width of building or roof length, across or along the ridge, for a HIGH Wind
Zone, height to apex of roof (H) from the various floor levels, and height
(above eaves) of pitched roof (h). Multipliers (to the chart figures) are
applicable for other Wind Zones – refer Notes below each chart. NOTE also
definition of H (different between Tables 5.5 and 5.6)
Figures are given in the chart for “across” the ridge or “along” the ridge (Refer
Fig 5.3). Depending on the proportion of roof height to overall height, across
the ridge figures provides a different figure for wind on the top storey of a
building as the charts assume a gable end (as opposed to a hip end). So for
hipended buildings for both directions of wind, adopt ‘across the ridge’ figures.
For the end of a building which has a gable end (and depending on the
proportions, more or less area of wall to face the wind) adopt ‘along the ridge’
figures
The value of bracing unites required for wind depends on the width of the
building perpendicular to the wind force. Depending on the shape of the
building this will vary. The maximum width is required taking full account of all
parts of a floor and whether or not floors ABOVE may have larger widths and
therefore higher wind forces
The figure obtained from the chart should be multiplied by the dimension of
the face of the building perpendicular to the wind (for roof slopes 250 or less.
______________________________________________________________
__
8
If the roof pitch is greater than 250 the roof length dimension should be used
as the multiplier. NOTE that wind values may depend on the length of the
wall or floors ABOVE the level of bracing particularly if any floor above is
wider – refer sketch
No matter how many parts of a complicated shape of the building is, the FULL
width is the measure of the windage to be calculated for the structure – in ALL
directions – refer above. Referring to the elevation, the bracing for level 2
should be based on the length of the parapet; the bracing for ground floor
level 1 could be based on level 1 width or conservatively the parapet length
The same applies for bracing units under seismic design actions where the
AREA of the floor/s above should be used in calculating the demand
Example – For the rectangular, two storey, gable ended farm building in
Ponga Road, determine the wind bracing demand in both major directions of
the top storey, ground floor and subfloor levels assuming the following :
Building is 12.5 degree pitched roof, 14.3 m by 8 m in plan, ground floor to
first floor level 2.8 m, ground level 800 mm below ground floor level, top floor
to ceiling level 2.4 m. All floors are timber framing and plywood flooring or
equivalent
______________________________________________________________
__
9
Building is in High Wind Zone, Height to Apex of building is given by
H = 0.8 + 2.8 + 2.4 + 4.0 tan12.5O = 6.9m GrdL / 6.1m lower FL/ 3.3m Upper
FL
Height of roof is 0.9 m (4.0 tan 12.5O) Chart is already for Zone H – no
multiplier – linearly interpolate between parts of each chart
Subfloor – Refer Table 5.5 – Across (NS) = 130 BU/m X 14.3 m = 1860
BU
( H = 7, h = 1 )
Along (EW) = 135 BU/m X 8.0 m = 1080 BU
Lower of 2 storeys – Table 5.7 - Across (NS) = 84 BU/m X 14.3 m = 1201

BU
( H=6.1, h = 1 )
Along (EW) = 93 BU/m X 8.0 m = 744 BU
Top Floor – Table 5.6 - Across (NS) = 33 BU/m X 14.3 m = 472 BU

( H = 3.3, h = 1 )
Along (EW) = 38 BU/m X 8.0 m = 304 BU
This indicates the impact of area of wind face and the high demand from wind
as the building gets higher
______________________________________________________________
__
10
Bracing Demand for Earthquake – How to use the Earthquake Charts
NZS3604 provides Charts for the determination of bracing requirements under

earthquake for each level of a building whether single or two storey and
whether supported on a concrete slab on ground or subfloor framing. Tables
5.8, 5.9 and 5.10 give values in bracing units per square metre of that building
floor or roof supported by the bracing, for earthquake zone 3 (Refer Fig
5.4), soil type D/E, weights of combinations of cladding & roof, and roof pitch.
Multipliers are given for various combinations for other zones and soil types
Note that with the recent earthquakes in Christchurch adopt Zone 2 for the
Christchurch region (refer Fig 5.4). Check with the local authority as we
expect local rules will also apply which may very well override the current
Zone 2 designation in areas surrounding Christchurch
The figure obtained from the Chart should be multiplied by the area of the
floor or roof above that level to achieve the total bracing demand for that
level. The figure will be the same for each direction as seismic attack is not
directional and the mass of the building at any level is constant
Example –
For the building in Ponga Road, determine the earthquake bracing demand in
each major direction of the top storey, ground floor and subfloor structure
assuming the following ; Roof is metal on timber purlins and in-plane metal
strap bracing, cladding at first and ground floor levels is plaster stucco.
Founding material is shallow soil site
Building is in South Auckland Earthquake Zone 1
Roofing is classified as Light – Cladding & subfloor is Classified as Medium
Area of each (timber) floor level is 14.3 X 8.0 = 114.4 m2 ; Aroof = 126 m2
Roof Pitch is 12.5O Soil Type is Class C – Multiply chart figures by 0.4
Select Table 5.9 – for Buildings on Subfloor framing – 2 storey
Read off Bracing Demand for –
Subfloor Structure 30 * 0.4 = 12 BU/m2
Lower storey Walls 26 * 0.4 = 10.4 BU/ m2
Upper Storey Walls 13 * 0.4 = 5.2 BU/ m2
______________________________________________________________
__
11
Total Bracing Demand under Earthquake =
1373 BU for Subfloor in each direction (cf Wind 1860 & 1080)
1190 BU for ground floor walls in each direction (cf Wind 1201 & 744)
655 BU for top floor walls in each direction (cf Wind 472 & 304)
This shows that in Zone 1 earthquake remains a factor in design and that
wind in this instance only governs in 2 instances
Bracing should be provided within the building to cover the demand indicated
from either Earthquake or wind whichever is the more demanding. If the wind
and earthquake demands are similar the designer needs to be careful to
check for the worst case as the capacity of bracing elements in earthquake
can be substantially lower in seismic than in wind (see later under GIB
Bracing Plasterboard Ratings
Summary of Bracing Demand for Ponga Road Property –
North South East West
Subfloor Framing 1860 (W) 1373 (E)
Ground Floor 1201 (W) 1190 (E) 1190 (E)
Top Floor 655 (E) 655 (E)
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12
Satisfying Bracing Demand from Horizontal Forces
Buildings require to transfer horizontal imposed loads from the elements

through to ground via various forms of bracing, frame action or structural
systems. For properly designed and detailed roofs and floors it can be
assumed that the roof structure (cross bracing, metal strap bracing, ply
diaphragm action) and the floors (timber diaphragm action, concrete floors)
can transfer the horizontal forces to the various walls provided the
recommendations of the Standard are applied
Generally within the building structure above the ground floor, horizontal
imposed loads are transferred through walls which, given various forms of
cladding or diagonal structure, will carry the loads from floor to floor.
Designers generally select two directions at right angles to each other, parallel
to the external walls and create imaginary “bracing lines” at the external walls
and internally at not more than 6 metres apart (5 metres for sub-floor bracing).
Where floor diaphragms are used wall bracing is required on all four sides of a
floor diaphragm (Cl 5.6.2). In addition the walls selected should if possible be
evenly distributed to ensure the response to a wind or earthquake force is
resisted without unacceptable torsion loads being imposed on the building.
This requires some skill and practice to achieve an understanding of how this
can be achieved in a real situation. External corners should always be
braced in each direction as these areas often attract high forces under wind or
seismic actions
General requirements for bracing along these lines are : -
Each external wall must have at least 15 BU/lin m of external wall or at least
50% of the total bracing demand for that level divided by the total number of
bracing lines whichever is the larger. Parallel walls not > 2m apart may be
considered as one bracing line
Each bracing line must have a minimum of 100 BU total within its length or
50% of the total bracing demand for that level divided by the total number of
bracing lines whichever is the greater. Walls parallel to and within 2 metres of
a “bracing line” can be considered as acting on that bracing line
Pedestrian doors, cupboard doors, windows, cavity sliders, walls backing onto
bathroom/shower units (attracting substantial plumbing) all interfere with
bracing and cannot be counted in the “length” of a braced wall
No wall bracing element shall be rated higher than 150 BU/m (fixed to a
concrete slab) or 120 BU/m (fixed to a timber structure)
External walls generally can only be braced on one side (external cladding is
not normally considered as adequate for bracing although plywood properly
secured to the external face of an external timber framed wall may be
impractical but could be considered as bracing on the external face
Internal walls have the facility of being braced with wall linings both sides,
indeed standard plasterboard fixed (in accordance with the recommendations)
both sides without special diagonal bracing and “hold down systems” can
considerably enhance the capacity of the building and in some instances of
single storey lightly loaded structures provide the entire horizontal load
carrying system
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13
Winstone Wallboards supply Plasterboard to the Industry. They have tested

several systems of plasterboard bracing systems and have published design
data for use in designing braced walls using their product. This course will
follow their recommendations and has been regarded as the “industry
standard” for many years.
Refer Gib Catalogue “Gib EzyBrace Systems” June 2011 , which we will be
using throughout and is available for downloading see www.gib.co.nz
(Products)
In particular Pages 10 and 11 which describe the various ratings for Standard
Gib and braceline systems offered by Winstone Wallboards. Refer also
Construction Details Pages 17 – 28
Notes to apply to the use of the GIB Bracing Rating Charts –
1. The Charts given in the assignment and Notes are particularly for 10 mm
Standard GIB and Braceline – The charts have amalgamated 10 and 13mm
thickness plasterboard – If intending to use 13 mm Gib, some savings may be
made by contacting Winstone Wallboards or using the software available
online
2. The values in the charts are based strictly on walls 2.4 m high (bottom of
bottom plate to top of top plate). For heights of braced walls in excess of
2.4m, the values in the charts must be reduced by 2.4 divided by the actual
height of the wall. For shorter walls the walls shall be rated as if they were
2.4 m high
3. Plasterboard is generally only applied to the internal faces of walls of a
dwelling. This means that external walls will only have plasterboard on ONE
side and most internal walls should have plasterboard BOTH sides – the
exception being say walls with cupboards where lining may be applied on one
side only – ensure you select the correct Type and use the correct figure for
BU’s/m based on single or both sides
4. Standard 10 mm GIB has some limited bracing rating and as such should be
used to the maximum because it is already in place. All GIB system will be
required to be fixed in strict accordance with the GIB Recommendations in
the June 2011 Manual and have the necessary (where appropriate) “Other
Requirements” such as hold downs for Braceline units
5. Braceline attracts a higher cost but achieves much more in BU’s per metre
and can be selected to replace “Standard GIB” in any place within a building
usually in multiples of 1.2 m (the standard width of a sheet) where that
dimension allows. “Other Requirements” such as hold-downs are required as
called for under the tables given
6. Note asterisked figures for Type BL bracing which limits BU values adopted
unless special care is taken to carry particular uplift forces
7. GIB Ezybrace Software available through the website is available for use by
Engineers and may deliver cost effective solutions. In this course it is
expected that students will demonstrate design from first principles and NOT
use Software
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14
Example :
Taking our Ponga Road property let’s look at the Provision of wall bracing on
the ground floor of the building to cover the demand from earthquake and
wind. Referring to Sketch the 14.3 by 8.0 metre rectangular floor, the plan
shows the layout of external and internal walls and the areas such as doors
and windows at which bracing cannot be applied. All flooring is Timber.
The process of determining suitable bracing solutions is as follows : -
• Divide the floor into bracing lines at less than 5 m spacing or 6 metres
if a double top plate is provided to the intersecting walls (Cl 5.4.6)
• Determine which walls are available for those bracing lines and
tabulate actual lengths of segments of each wall – note that tee
junctions need not necessarily require the bracing unit to stop and
start. This Board example takes lengths of wall each side of a tee
junction and the length as a clear dimension
• As a first estimate of Demand divide the Total demand for the floor into
the number of Bracing Lines
• Divide the “average” demand for each wall by the total length of wall
available for each line
• This enables the designer to see at what level you may require bracing.
If the BU/m is higher than 60 – 80 then “Braceline” systems will be
required. On the other hand if the BU/m is low (<50 BU/m) then some
walls can be neglected and the Standard Gib Bracing systems could be
adopted
• Based on the assessment above, select a suitable system for each
segment of wall and calculate the bracing demand for each element
and then for each wall
• Collate the total for bracing ratings for one direction, reassess and
recycle as necessary to refine the solution
• Note that bracing cannot go behind bathroom/laundry/showers units
• Plywood is a very good product as bracing when fixed to the timber
framing or backing to cladding but note that it cannot be used if a cavity
is required
Summary of Bracing Demand for Ponga Rd Property (From Page 11)
North South East West
Subfloor Framing 1860 (W) 1373 (E)
Ground Floor 1201 (W) 1190 (E) 1190 (E)
Top Floor 655 (E) 655 (E)
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Therefore – for the Ponga Road Property Ground Floor (only)–
1. By inspection Seismic governs in both directions. Wind governs in one

direction only and only for the subfloor bracing design – Design ground
floor for 1190 BU for both north south and east west
2. Divide building into four and three bracing lines as shown, note the
short 2.0 metre N/S wall in the centre of the building is not within 2
metres of either wall – at this stage ignore this but the floor could have
been divided into 5 bracing lines if it was necessary
3. Lines A, B, C and D – Average Demand for each line is 1190 / 4 = 298
4. From calculation of the available wall space BU/m ranges 66 – 78
BU/m which indicates Braceline may be necessary for one sided
(external) walls but that the internal walls may be OK with GS2(10)
5. Spreadsheet enables each Wall to be assigned a bracing element if
necessary – Note 2.8 m high walls for this floor level and timber
flooring throughout
6. Many combinations are available for mixing Bracing types and having
some walls braced and others not. With experience comes wisdom of
cost effective design and practical solutions
7. If there is not enough walls for bracing then advice from a structural
engineer needs to be sought
Designation for Bracing Elements varies from Company to Company –

information that is needed is : The type of bracing element, the length of
element and a cross checking element number which is unique to that
element and can be found in calculations - two examples are shown on the
sketch and CAD systems can be set up to produce these quickly
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Pong Road Property ‐ Ground floor ‐ Seismic governs ‐ Required 1190 BU's Total East‐West
Note ‐ Wall height is 2.8m Use 10mm GIB throughout & Timber floors
East‐West WALL P WALL Q WALL R

External Internal External
Target 397 BU's 397 BU's 397 BU's
BU's required per meter of wall available (all walls used):
57 BU's/m 33 BU's/m 44 BU's/m
Height factor: 2.4/2.8 = 0.86
These figures are <70 ‐ 80 BUs/m and therefore GS walls could be used for both internal and external walls.
If demand is not met, external corners could go to BL.
WALL P External WALL Q Internal WALL R External

Length No Total BU's Total BU's Length No Total BU's Total BU's Length No Total BU's Total BU's
Type BL1‐H Type GS1‐N Type BL1‐H Type GS2‐N Type BL1‐H Type GS1‐N
1.2 P1 108 0.6 1.5 R1 135
2.2 P2 113 2.4 Q1 175 1.5 R2 77
1.8 P3 93 2.9 1.9
0.9 2 1
0.9 P4 77 1.2 3.1 R3 279
2.8 Q2 204
Total for walls
selected: 7 m 185 206 11.9 m 0 379 9m 414 77
Total all walls: 391 BU's 379 BU's 491 BU's
Total demand required: 1190 BU's Total provided: 1261 BU's > 1190 BUs
NOTES 1. Figures of target BU's per metre indicate could use GS Bracing throughout
2. Could pack internal wall Q with GS2 but in this instance go to BL1 on external corner walls P & R
3. External walls at corners are susceptible to high demand and Braceline is often used for corners
Sub Floor Bracing Design (Section 5.5) –
Foundation and subfloor bracing are aspects of building construction requiring

careful design. Concrete slabs and footings poured direct to ground are
straight forward and there are many guides on the attention needed or
detailing to achieve a good solution -
For buildings where the subfloor height is very variable or high, specifically
designed pole frame structures may be cost effective
For the majority of structures with subfloor timber framing the horizontal loads
arising from wind and earthquake can be transferred to ground in a number of
ways. Choice of which is dependent on the type of building and topography
and geotechnical; aspects of the site
Subfloor bracing systems can consist of :-
1. A continuous perimeter foundation wall carried up to a timber plate

which is fixed directly to the floor joists of a structural diaphragm
2. An evenly distributed perimeter system consisting of braced piles,
anchor piles or sheet bracing
3. Corner foundation walls at four or more corners combined with anchor
or braced piles on external walls and an internal bracing system
consisting of shallow founded cantilevered piles
4. Driven or drilled and placed pole or square timber cantilever piles
Piled Foundations to Resist Horizontal Design Actions
There are four types of foundation specified in NZS 3604 designed for
horizontal Loads : -
• Ordinary Piles (either precast concrete or timber) – carry vertical loads

only – see later
• Cantilever Piles – driven timber poles
• Braced Piles – Excavated and placed in concrete, timber piles and
braced to each other
• Anchor Piles – drilled and placed timber poles
Examples of each are illustrated
Tables show the Bracing Units which can be carried by each Type of
foundation
Also illustrated is the types of fixings available for standard foundations
required to transfer the design forces described in the Standard
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DESIGN FOR VERTICAL LOADS
Foundations and “Good Ground” – Section 6
The essence of structural engineering is to carry loads created by mankind or

nature into the ground using adequate, safe and cost effective means. If the
ground itself is incapable of sustaining the loads then “design” is not complete
until appropriate measures have been specified to ensure it does, either by
modifying the subgrade material or providing alternative means of foundation
The standard specifies a 300kPa minimum ULS soil bearing capacity and this
can be checked via a geotechnical technician and advised or reported. The
standard specifies exclusions and as mentioned previously some further
recommendations may be arising from the liquefaction experienced in the
Christchurch earthquake. In all cases it is wise to obtain good advice here as
many problems have occurred by people having foundation problems with
long term settlement or slips occurring in adverse weather when judicial
geotechnical advice obtained early in the project was needed. Having
knowledge of adjacent buildings and their performance under their foundation
type is also useful information. Council often have substantial knowledge of
surrounding site problems and should be consulted early in the project. Rules
are -
• Ensure you have sufficient soil strength
• Ensure there are no “hidden” problems
• Ensure the topography and site conditions don’t lend
themselves to future foundation problems such as instability or
scour (due to flooding)
• Where appropriate and when you are not sure, advise the Client
to engage a geotechnical expert to give advice
• Consult the Council on any knowledge they may have in the
area
Light Commercial and Residential Buildings can be founded on concrete strip

footings (below reinforced concrete or masonry subfloor walls), concrete pads
supporting timber or concrete piles, driven or in-situ concrete/timber piles.
Generally lines of foundation elements on the building perimeter and through
under the building are required. Concrete footings, slab thickenings and
concrete slab on ground are very common with or without perimeter walls to
take up levels of the site
Tables are available within the Standard to determine size and shape of
discrete footings depending on the choice of bearers and joists and the
number of storeys of the building
Basis of design is to decide on perimeter and rows of discrete foundations
which may or may not coincide with load bearing partitions and therefore
govern the design and span of floor joists. Refer Board Example
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18
Selection of Pile Type –
For simple discrete foundations carrying gravity loads only - Ref Fig 6.2
Ordinary Piles – Select square, timber. Concrete or round piles placed on
circular or square footings. Guidance is given as to concrete strength
(minimum 17.5 MPA 28 day strength), depth below cleared ground, thickness
and plan size of footing refer Table 6.1
Floor joists carry the flooring and span normally on top of bearers which are
supported by foundation piles. The size and shape of foundation is
dependent on how much load is imparted to each, therefore the designer
needs to assess a suitable layout of floor joists, bearers and foundation pads
in order to select sizes
For example using Table 6.1, with floor joists spanning 3.3 metres and
bearers spanning 1600 mm as subfloor to a single storey dwelling, a circular
footing supporting an ordinary pile, would be chosen as 460 mm dia
Bearers can be supported by timber or concrete piles direct or the pile can be
extended by a jackstud (depending on the level of ground floor above
surrounding ground) – refer Fig 6.3
Table 6.3 gives maximum unsupported heights (lengths) of subfloor jackstuds

(Grade SG 8) (2 sizes 90 x 70 and 90 x 90) given joist spacing and span of
bearers
Table 6.4 gives (Grade SG 8) bearer sizes for two different floor loads and
various dimensions of the “loaded dimension of a bearer” – refer Fig 1.3 –
effectively the same as the dimension associated with “the contributory loaded
area” with which we are familiar. The “loaded dimension” for any truss, beam
or post is effectively the length of span being supported by the item. For
example if a floor joist spans 3.8 metres between foundation walls or bearers,
each wall has a loaded dimension of 1.9 m – however if the wall carries floor
joists from the other side which spans say 2.4 metres (on its own imparting
1.2 m span to the foundation wall) , the total “loaded dimension for the centre
foundation wall or bearer is 1.2 + 1.9 = 4.1 metres, the outside bearers “carry”
1.9 and 1.2 m “loaded dimension” – see and explain Fig 1.3
Referring now to Table 6.4 if the bearers in the example just given were
spanning say 1.3 metres and were required to carry a residential load of
1.5kPa, reading Table 6.4 would give
• 90 x 70 for the external bearer (carrying 1.2 loaded dimension)
• 90 x 90 for the external bearer (carrying 1.9 loaded dimension)
• 140 x 90 for the internal bearer (carrying 4.1 loaded dimension)
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Class Example of Subfloor Design – Refer Proposed House (See Sketch

over)
Single storey, “Light” roof, 15˚ pitch, “Light” wall cladding & Subfloor Cladding
Building Height ~ 5.0 m , Roof Height ~ 1.0 m , Height of subfloor to underside

of floor joists ~ 1.2 m, Floor Plan 11.0 m by 7.0 m
Wind Zone High, EQ Zone 1, Soil Type C (shallow), assume standard

dwelling floor load 1.5 kPa
Try three rows of piles down the length of the building, floor joists span 3.3
metres? OK – divide length by ~1600 → 7 rows of 1500 mm cs and 250
clearance at each end
Bearers – Span 1.5 m - From Table 6.4 for the external bearer, an SG 8
bearer size of 140 x 70 can carry a loaded length of 2.2 over 1.65 m (cf actual
1.85 m loaded length over 1.5 m). For the internal bearer (actual loaded
length 3.3 m) go to 190 x 70 bearer (4.1 m maximum loaded length for a
span 1.65 m)
Piles – Are to be a combination of ordinary, braced and anchor piles
Bracing Requirements at Subfloor level –
Wind – 80 BU/m across ridge x 11m = 880 Bu’s

Table 5.5 90 BU.m along the ridge x 7 m = 630 BU’s
Earthquake - 15 * 0.4 = 6 BU/sq m = 462 BU’s (Table 5.8)
Satisfy demand by
Wind governs – Allow for bracing in the pile lines along the building in
each of the three rows and anchor cantilever piles across the building in each
of three rows (5.25 metres apart). Accept as OK despite Cl 5.5.2.1 (c) -
subfloor bracing lines should not exceed 5.0 apart – refer sketch
Select Fig 6.7 or 6.6 type braced piles each with a rating of 160 BU’s
under wind and 120 BU’s under earthquake
Select Anchor Pile as Fig 6.10 with a rating of 160 BU’s under wind
and 120 BU’s under earthquake
Along the building –

In each of two external rows have one braced system and one anchor
system = 320 BU’s (W) 240 BU’s (E)
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20
In the central row have two braced systems = 320 BU’s (W) 240 BU’s
(E)
TOTAL 960 BU’s (W) and 720 BU’s (E) > 630W and 462E OK
Across the building –

Row M , two anchor piles = 320 BU’s (W) 240 BU’s (E)
Row N and Row O , two braced systems = 320 BU’s (W) 240 BU’s (E)
TOTAL 960 BU’s (W) and 720 BU’s (W) > 880W and 462E OK
Note system is relatively symmetrical and situated such that no brace has
more than two diagonals applied
In the diagram: A = anchor piles, B = braced pile system and O are

ordinary piles
Check angle of the brace – a minimum angle is specified for each type
of braced system and depending on the room or crawlspace under the
building braced piles systems may not be applicable given the lack of room to
fit a brace at a suitable angle
Worst braced bay is 3.3 m long and say 0.9 meters room in height. Angle of
brace is 15o which is > than the 6o or 10o specified for the two types of brace.
Note if this room was not available then anchor piles would be required
throughout
Note also that the minimum requirement of 100 BU or 15 BU/m * 11 m

= 150 BU along the external walls is also satisfied
Floor Joists – (Refer next few pages for Design Rules)
Clear span of floor joists is 3.2 m – (can’t be continuous over 6.6 m and
cantilevered 2 * 200 mm, as single lengths of timber in excess of 6.1 metres
are not readily available)
From Table 7.1 for 1.5 kPa and SG 8

Select – 240 x 45 at 600 mm centres (max span 3.90m) or 190 x 45 @ max
450 cs (max span 3.45 m )
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FLOORS – Flooring –
There are many materials used for structural flooring (termed panel products
or tongue and grooved dressed timber), spanning between floor joists and
often acting as a horizontal diaphragm transferring lateral loads to braced
walls. These are all available by referring to manufacturers literature online
and following the design recommendations for their respective products. Refer
to CECIL for the latest catalogue for “Ecoply” a CarterHoltHarvey Product.
Note carefully that, material used for “Domestic Use” may not be suitable for
“Residential Use” such as in Resthomes or “Commercial Use” such as in
Offices, Schools and Domestic Garages. It is also likely that some panel
products or design charts do not cover the point load requirement for
residential garages (9kN). The key is to refer to the product literature and
NZS3604 Table 1.2 Basic Imposed Loads and if the Category is not covered
there, then refer to AS1170 Loadings Standard Table A3.1 (Note Note(7) for
domestic garages). Then to carefully read the Manufacturer’s literature to
ensure their product meets the required Standard Specification for Structural
Flooring and the strength/stiffness requirements for the floor loadings being
imposed whether gravity imposed or lateral in diaphragm action.
• Particle Board, waferboards, customwood and fibreboards (HDF) –

meeting the requirements of AS/NZS 1860.1 Specifications for
Particleboard Flooring - available in various thicknesses usually in
sizes 2400 by 1200 and nailed, screwed and/or glued to floor joists. All
have varying densities and carrying capacities and Engineers should
be aware of the limitations of each and refer to the manufacturer’s
literature for recommended span, duty and fixings and meeting the
durability requirements of NZ Standards. Generally acceptable in
Residential and Commercial Flooring footway traffic only. Not suitable
to meet the requirements of vehicle traffic.
• Fibre-cement Board often used in potential wet situations such as
showers and ensuites – refer to Manufacturers literature and don’t
forget the substantial weight of these materials. An example is James
Hardie “Hardiepanel Compressed Sheet” which is available in 18 mm
thickness and should be fixed in strict accordance with the
recommendations of the manufacturer
• Plywood meeting the requirements of AS/NZS 2269 Plywood –
Structural. It is essential that Engineers specify correctly the Grade and
Standard which structural ply flooring is required to meet. There are
many imported and “non-standard’ plywoods available which may be
cheaper but may not meet the standard required for buildings in New
Zealand. The most readily available structural plywood material is F8.
Note: Table 7.4 in NZS3604 for selection of plywood thicknesses
applies to Stress grade F11 only and therefore is generally inapplicable
in NZ without additional expense to specify F11 only. Unless you want
something specifically for a higher grade (that is stronger and stiffer
material for a given thickness), and you or your Client is prepared to
pay the extra for it then stick with F8. Plywood is placed with the
outside grain of the board perpendicular to the supporting floor joists
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and is available in varying thicknesses. Usually start with a reasonable

selection of floor joist spacing for a particular duty (whether 1.5 kPa
imposed load in a domestic situation or up to 3 kPa for a commercial
floor) and this governs the selection of ply thicknesses. Refer to the
recommended centres for each thickness in order to make an
engineering judgement and selection
• In determining the span for a floor joist adopt the “clear” span between
support members – see Fig 1.3 and definition of “span”
In all cases select floor joist spacings of 600, 450 or 400 mm to suit an even
spacing for the size of sheet and in most cases use the most logical thickness
for the largest span. The fewer the number of floor joists usually the better
(although the thickness of ply may be slightly higher and the depth of floor
joist greater, often the labour of fixing more floor joists overpowers the extra
money spent on more material
Select flooring thickness and therefore joist spacing before you design
the joists. When selecting ply flooring say for the living spaces of a dwelling
or office building don’t change the ply thickness through the building floor.
The top of the floor level is required to be level throughout the area and to ask
a builder to adjust his floor joists by 3 or 5 mm in height because you have
selected a different thickness of plywood makes little sense
Floor Joists –
For SG 8 floor joists Table 7.1 provides guidance for the size of sawn timber
floor joists for the two floor loads of 1.5 kPa and 2.0 kPa for various centres of
joist and maximum clear span
When selecting floor joist sizes for various spacing, one method of testing
economies is to determine the volume of material per square metre of floor.
Without obtaining actual costs of timber sizes per lineal metre installed, a
gauge of cost difference can be made by calculating the volume of timber per
square metre in common units. For example, a 290 by 45 at 600 mm cs
(volume = 290 by 45 divided by 600 = 21.75) has less volume of timber per
square metre of floor than say a 240 by 45 at 400 mm cs (volume = 240 by 45
divided by 400 = 27.00 – also 400 mm centres attracts higher labour content
with more joists to be placed – eg for 600 mm cs on a job, 72 joists may be
required (in the dwelling) if at 450 mm centres say require 97 joists.
Admittedly there is a saving on the less thickness of flooring but with
approximately the same labour content to fix both the saving is small. These
rules of thumb are only a coarse guide for preliminary work – eventually in the
office some guide may be sought on actual costs in place of various systems
and some Firms have very good databases for establishing costs of
construction and elements
For an example, design a floor joist for a residential property with bearers
running perpendicular to the floor joists at 2.4 m and 3.5 m centres. Clear
spans are 2.3 and 3.4. We would select (at 600 mm cs Table 7.1(a)) 190 x 45
(capacity 3.15) and 240 x 45 (capacity 3.9). Note if you were prepared to
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23
construct the floor with a single piece of timber (length) 6 m (maximum

obtainable 2.4+3.5+0.1), could increase the values in the table by making the
floor joists continuous over the central bearer – 190 x 45 achieves a maximum
span of 3.47 enough to cover the actual longest span of 3.4 – can only do for
combined lengths < 6m
Varying floor joist depths over the entire dwelling has some merit but be
careful. If the architect or the Client wishes to have ceiling heights the same
throughout the dwelling then if the ceiling is attached directly to the underside
of floor joists (commonly) then where the joists are less in depth additional
framing is necessary to drop to the correct ceiling level. This (with high labour
content may be more expensive than providing the increased depth over the
shorter span
Chapter 14 of NZS3604 provides Tables for 3.0 kPa floors – refer below.
Note that residential garage floors under NZS1170 Table 3.1 Note 7, attract a
spread imposed load of 2.5 kPa along with a 9 kN point load. In this instance
NZS3604 does not cover this situation and specific design should be called
for
Layout Criteria and Lateral Restraint – Refer Fig 7.2
When determining layout of floor joists in a multistory building, always look at

the plan for the floor below to decide the layout. This determines where the
supporting beams or walls are. It is not necessary that floor joists run the
same direction over an entire floor, normally floor joists will be run over the
shortest span
In most construction various joist depths are required where some parts of the
floor have 290 by 45 and others only 140 by 45. The plyflooring of course
must be at the same level throughout and this is achieved by blocking the
underside of the smaller joists with an extra bottom plate where appropriate.
In general do not change thickness of ply throughout a single floor unless of
course there is a distinct change in structure for example with a garage
attached to a dwelling
In order for floor joists to perform properly the Standard gives guidance on
what to provide for stiffening between timber floor joists. Beams carry load by
bending over long distances where the top of the beam goes into
compression and the bottom of the beam into tension. However if the top
compression part of the beam is not laterally restrained it can move sideways
and seriously affect the carrying capacity of the beam. Solid blocking with
pieces of timber between provide the stiffening necessary to allow the beam
to be held upright and significantly improves the performance of the joist or
beam. In addition the flooring whether in particle board, t&g flooring or
plywood provides diaphragm action in the distribution of lateral wind & seismic
loads into the structure below. If the floor system is floppy and not strutted the
floor joists provide a weak link. Fig 7.2 provides guidelines and students
should become familiar with the requirements and expect to see this on site –
Cl 7.1.2.3 requires floor joists over 2.5 m span or with a depth more than four
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24
times its width to be blocked or strutted at midspan. Ends of joists also

require restraint
Floor Joists supporting Walls – Refer Cl 7.1.3 and Fig 7.3
Walls in buildings can be non-load bearing (partitions which provide no reason

for being there but to provide a screen between rooms, do not carry any roof
load but may be required to be braced to carry lateral wind or seismic loads)
or load bearing where the upper floors or roof is solidly supported by the wall
and can be braced to carry lateral loads
• Non-load bearing walls which run parallel to the joists and contain
bracing elements must be above a joist (often doubled up) or have
solid blocking between the joists. If the wall does not have bracing
elements there must be a parallel floor joists within 150 mm of the wall
• Load bearing walls running parallel to the floor joists must be supported
by a double joist
• Load bearing walls running perpendicular to the floor joists must be
situated within 200 mm of a bearer (or the floor joists specifically
designed). Refer Fig 7.3 in the standard
Chapter 14 of the Standard – 3 kPa Live Load Structures
The Standard provides for the possibility of designing structures which attract
the higher imposed load of 3 kPa – eg Office Floors, Operating Theatres,
Light Industrial work rooms, laundries, laboratories, classrooms, lecture
theatres (not public) all come under the requirement for 3 kPa imposed
loadings. The complete set of Charts are provided in Chapter 14 for each
element of structure required to be designed
In some commercial premises some of the floor such as in plant rooms, or

filing rooms where, say, proprietary (Lundia shelving) storage is required,
imposed loadings in excess of 3 kPa is required. These areas require
specific design and cannot be considered as coming under NZS3604. There
have been many times when premises have suffered substantial deflection
under imposed loads and found that design has not taken into account the
very high local loads being imposed. Watch out for this!
______________________________________________________________
__
25
WALLS – Section 8
Walls are used in structures to –
• provide vertical support to the structure above

• transmit out of plane horizontal wind and earthquake
loads (spanning vertically) to the levels above & below
• provide support (as a shear wall) to in plane panel
products as bracing
As discussed previously, in-plane horizontal loads are assessed in terms of

the bracing requirements of the Standard and bracing in terms of sheathing
and or structural items such as diagonal timbers and hold down systems are
provided as required.
To ensure the vertical loads are carried adequately attention must be given to
elements such as
• Size and spacing of studs including trimming studs
• Top & Bottom Plates
• Lintels over doors and windows
• Sills for windows
Top Plate
Nogs
Lintel
Trimmer Stud
Sill
Studs
Bottom Plate
ELEVATION TIMBER WALL FRAMING
______________________________________________________________
__
26
Wall Framing –
Load Bearing Walls (External & Internal) –
Size and spacing of studs in walls are dependent on the length (or height) of
the wall, the type of load being carried, the length of the loaded dimension
and its location (ie is it an external wall subject to face loads from wind), the
designation of the wind zone - In each chart is a special case for internal walls
which are not subject to high face loads due to wind forces
Table 8.2 a, b and c (For SG 8) describes the size of studs given whether or
not the stud is in the top storey (light or heavy roof) or the lower of two
storeys, a selected spacing, height of wall, loaded dimension and wind zone
(or an internal wall)
For example – the size of an external load bearing stud in a 3.0 m high room
on the top floor of a building to NZS3604 carrying a concrete tile roof and a
loaded dimension of 2.4 metres, in a high wind zone would be 90 x 70 at 400
or 600 centres
If the wall was internal carrying a loaded dimension of 4.8 metres, a 90 x 45

stud could be selected at 600 mm cs or 90 x 35 at 480/400 mm cs
Non-Load Bearing Walls (External and Internal) –
Size and spacing of studs in non-load bearing walls are dependent on the
height (length of the stud) of the wall, the spacing and the wind zone (for
external walls) - In each chart is a special case for internal walls which are
not subject to high face loads due to wind forces
Table 8.4 (for SG 8) gives guidance on the non-load-bearing stud size and
spacing given the height and wind zone (for external walls) or internal walls
For example the stud size for an internal 3.0 m high non-loadbearing wall for
600 mm cs could be 90 x 35
Notes for Studs – Judicious selection of stud sizes (after design) is important
as the builder will not thank you for having many sizes of studs throughout
any level of floor where the chances of getting things wrong are high and the
advantages of purchasing pack lots can be missed. That is after proper
design it may be useful to provide a single (or two) size/s of stud throughout
the one level which covers all requirements leading to efficiencies in purchase
and few errors likely. On the other hand if there is one small area at say an
entranceway where the walls are unusually high for an effect it is silly to make
all the remaining wall studs the same heavy size demanded by the extra high
wall – it is a matter of engineering judgement
______________________________________________________________
__
27
LINTELS –
Lintels are structural members that span across openings. These members
support floor joists or roofs that are situated above, the loads being similar to
those carried by load bearing walls
The size is dependent on what is being carried, the clear span of the opening
and the loaded dimension of the load being carried. Generally the width of
the lintel beam should match the depth of the wall – ie a 90 mm depth of wall
would generally attract a width of lintel of 90 or two members 45 width each.
Other than sawn timber, there are many innovations in lintel design with flitch
beams (a mixture of steel and timber), plywood ‘cee’ beams, sheet steel
beams, laminated and gang laminated beams, trusses and rolled steel
sections. They have arisen to provide cost effective solutions to long span
lintels. They are not able to be covered in this standard and can be sought
via specific design from a structural engineer or using manufacturers literature
for proprietary systems
The charts in the standard are based on design from a uniformly distributed
load for roof pitches <= 450 – it is important to ensure the structure above
applies the load uniformly and not through the use of one or two discrete
studs (in which case the design tables cannot apply). Generally load
application through studs at 400/600 cs conform – the standard requires that if
load application is greater than 1200 mm cs then the tables may not apply.
Ref Table 8.7 multipliers or SED for roof pitches > 450
There may be occasions when the lintel is carrying very little applied loads
because the floor joists span parallel to the door or window or the roof trusses
or rafters also span parallel to the lintel. In these instances, although the
loads are low there may be situations where the lintel beam is required to
carry a point load from maintenance or called upon to spread face loads due
to wind (at least for the depth of the lintel up to the eaves) back to the vertical
studs. In these instances adopt the minimum loaded dimension in the Chart
(usually 2.0 m) and select the size of lintel to accommodate the clear span
given
Detail –
• Lintels greater than 150 mm depth require a doubling stud for
support which increases the transfer of vertical loads adequately
and provides good stiffness at the sides of the opening
• Lintels that directly support roofs often require additional tie down
during uplift situations
Refer Sketches – Figs 8.2, 8.5
______________________________________________________________
__
28
Selection –
Standard Tables 8.9 through 8.13 for SG 8 provide guidance on the maximum
clear span of lintels given the various sizes, loaded dimension and type of
load being carried (light or heavy roof, floor or not, wall above or not). The
sketches below each table provide an excellent understanding of the location
and expectation from the table. Note particularly whether or not floor joists
span or do not span onto the lintel. The heaviest load generally carried by a
lintel is a floor above and if this is missed some failures can result
Note that these charts assume a maximum of 750 mm overhang – the

overhang dimension therefore should not be added to the loaded dimension –
refer Fig 1.3 Definitions of Spans and Loaded Dimensions - (about page 26a
of the Notes)
The charts assume the loaded dimension is an effective UDL on the lintel
beam for the entire length of the beam. In reality the roof shapes particularly
with a hip roof do not provide a uniform load but a triangular or trapezoidal
load. In addition care must be exercised where unless you investigate
carefully, point loads may also impose onto the lintel (from floor beams etc) in
which case specific design is warranted. For trapezoidal loading (see Board
example) it is suggested you draw to scale the contributing area of the non-
rectangular area and form a judgement on an appropriate “uniform loaded
dimension” which covers the actual situation and enables the lintel NZS3604
charts to be adopted
Example (all using VSG8 dry timber) –
1. Lintel (Clear span 2.1 m) supporting a wall and roof only (roof trusses
span 8.6 metres) – roof is metal (light) and walls are stucco plaster
(medium) – Table 8.10 – Interpolate between figures - select 2 / 240 x
45
2. Lintel (Clear span 3.0 m) supporting heavy roof only (roof structure
spans 12 metres) – Table 8.9 – No go – Specific Design required
3. Lintel (Clear span 1.8 m) supporting a floor, wall and roof (roof trusses
span 7.8 m, floor spans 3.2 m), cladding is light, roof is metal. Table
8.11 – use the maximum of loaded dimension of 1.6 m (Floor) or 3.9 m
(roof) – use 3.9 m for loaded dimension – select 2 / 240 x 45
______________________________________________________________
__
29
Lintel Fixings – Figure 8.12 and Tables 8.19 and 8.14
Within the spans given for lintels in Table 8.14, standard nailed fixings in
accordance with nailing schedule Table 8.19 is permissible. If the lintel span
exceeds those given in Table 8.14 then uplift fixings (7.5 kN transfer) is
required (refer Fig 8.12)
For instance for Example 1 above (2.1 m lintel span carrying 4.3 loaded
dimension) say for Medium Wind Zone , Table 8.14 requires uplift fixings (max
span 1.4 m without fixings and 5.0 m with fixings)
SILLS – Refer Table 8.15
Sills are provided for the bottom support of the window or opening and
logically should be on their flat and the same width as the depth of the wall
studs. Depending on the opening size the thickness of the sill may vary –
refer Table 8.15
For example the sill for Example 2 above (3.0 m window) would be 2 / 90 x 45
TRIMMING STUDS – Refer Fig 8.5 and Table 8.5
The studs at the side of each opening is required to carry the additional axial
load from the lintel beam as well as transfer face loads (say from wind) on the
face of the opening from the sill and lintel through the stud into the top &
bottom support
Table 8.5 provides guidance on selection of trimmer stud thickness (depth

being the same as studs in the wall). Remember for lintels greater than 150
deep a “doubling” stud is required. In addition, for lintels less than 400 mm
from the bottom of the lintel to the ceiling level the doubling stud may be
included in the “thickness of trimming stud”
Table 8.5 gives the trimming stud thickness given the thickness of studs
already in the wall and the width of the opening – eg for a wall with existing
studs 45 thick and an opening of 3.3 m attracting a lintel 2 / 240 x 45 adopt 2 /
90 x 45 trimming studs. Since the lintel depth is > 150 the design requires a
“doubling stud – now, if the distance from the ceiling level to the underside of
the lintel is > 400mm then allow three vertical members (2 trimmers and a
doubling stud); if this distance is < 400 mm then only two vertical members
are needed since the doubling stud can be included within the trimmer stud
requirement
______________________________________________________________
__
30
TOP & BOTTOM PLATES – Tables 8.16 and 8.17
Top & Bottom plates are provided to distribute the vertical and lateral loads
from the roof, floor or wall studs above to the structure below and to tie the
studs together.
Size of top plates are dependent on the
• Position and spacing of the member being supported (roof trusses or

rafters or floor joists)
• Span of the member being supported
• Spacing of the studs below
Table 8.16 a, b and c gives the maximum loaded dimension of the roof
structure which can be carried by a top plate given a light or heavy roof, stud
spacing, spacing of floor joist or roof rafters & trusses, loaded dimension of
the floor. A further requirement is to consider whether or not the rafter/truss is
supported within or outside some 150 mm from the stud (along the top plate)
For the three examples given above for Design of the top plate –
1. Lower of two storeys, Wall & Roof only, Light roof, stud spacing 600
mm, loaded dimension of roof 4.3 m, loaded dimension of floor is zero
– select 90 x 45 as top plate
2. Top storey, heavy roof only, loaded dimension 6 m, 600 stud spacing –
select either 2 / 90 x 45 (roof members can go anywhere along the
plate) or if you can ensure the roof members are within 150 mm of
each stud, 90 X 45 plus a 140 x 35
3. Lower of two storeys, light roof (trusses at 1200 cs) with loaded
dimension 3.9 m, floor (joists at 400 cs) with loaded dimension 1.6 m,
stud spacing 600 mm – select 90 x 45 as top plate
Size of bottom plates are dependent on the
• Spacing of the joists below the bottom plate

• Span of the floor joist members being supported and the span & type of
roof being supported
• Spacing of the studs
Table 8.17 gives the maximum loaded dimension of the roof structure which
can be carried by a bottom plate given a light or heavy roof, stud spacing,
spacing of floor joists, and where applicable, the loaded dimension of the
floor.
______________________________________________________________
__
31
For the three examples given above for Design of the bottom plate –
4. Lower of two storeys, Wall & Roof only, Light roof, stud spacing 600
mm, loaded dimension of roof 4.3 m, say supporting floor joists are at
600 mm cs, loaded dimension of floor is zero – select 90 x 70 as
bottom plate
5. Top storey, heavy roof only, loaded dimension 6 m, 600 stud spacing,
floor joists at 600 mm cs – select either 90 x 70
6. Lower of two storeys, light roof (trusses at 1200 cs) with loaded
dimension 3.9 m, floor (joists at 400 cs) with loaded dimension 1.6 m,
floor below at 400 mm cs, stud spacing 600 mm – select 90 x 70 as top
plate
Top and bottom plates for non-load bearing walls are to be the same depth as
the wall studs and not less than 40 mm thick
Fixings for Timber framed Walls – Table 8.19
Table 8.19 in NZS3604 provides guidance on the standard fixing required for
various elements of Timber Framed Walls
______________________________________________________________
__
32
Example of a Proposed Dwelling – Refer attached Sketches
Wall Design –
Single storey, “Light” gable-end roof, 15˚ pitch, “Light” wall cladding &
Subfloor Cladding
Building Height ~ 5.0 m , Roof Height ~ 1.0 m , Floor Plan 11.0 m by 7.0 m
Wind Zone Medium, EQ Zone 1, assume standard dwelling floor load 1.5 kPa
Additional information – Soil Type C, Stud height 2.4 m, Timber Floors, Use
10 mm GIB throughout
Ground Floor Wall bracing Requirements –
Wind – Table 5.6 – Across ridge - 50*0.7 BU’s / m x 11 = 385 BU’s

Along ridge – 55*0.7 BU’s / m x 7 = 270 BU’s
Seismic – Table 5.8 - 11*0.4 (1C) BU’s / sq m x 11 x 7 = 359 BU’s along

and across the building
Minimum bracing requirements – 3 bracing lines in each direction

• Brace evenly throughout and all external corners
• Any braceline at least 100 BU’s or Total/2*3
• External Wall 15 BU’s/m of external wall = 165 and 105
Dirn Brcg Type Lengt Rating Rating Earthquake Wind Minimum

Line Number h Seismi Wind Demand
c
Along A(E) BL1 1 0.9 100 90 90 81
A(E) GS1 2 1.2 60 70 72 84
Total A 162 165 165
B(I) GS2 3 1.2 85 95 102 114 100
C(E) BL1 4 0.9 100 90 90 81
C(E) GS1 5 1.2 60 70 72 84
Total C 162 165 165
Total 426>339 444>270
Across M(E) GS1 6 0.6 55 50 33 30
M(E) BL1 7 0.7 100 90 70 63
M(E) GS1 8 1.2 60 70 72 84
Total M 175 177 165
N(I) GS2 9 1.2 85 95 102 114 100
O(E) GS1 10 0.9 55 50 50 45
O(E) GS1 11 2.4 60 70 144 168
Total O 194 213 165
Total 471>339 504>385
______________________________________________________________
__
33
Framing Sizes –
Roof Trusses and/or rafters are at 900 mm cs and span across the building
between external walls. All internal walls are non loadbearing. External walls
loaded dimension is 3.5 m, Medium Wind, assume 600 mm cs
• External Walls (Table 8.2) select SG8 90 x 35

• Internal Walls (non load bearing Table 8.4) could be SG8 70 x 45
Both these sizes conform with Gib Bracing Systems recommendations
ADOPT – External walls as all SG8 Timber - 90 x 35 @ 600 mm cs and

Internal Walls as 70 x 45 @ 600 mm cs
Top & Bottom Plates –
Top Plates, External Wall – Table 8.16 – with no restriction on where the
rafter is supported by the top plate select 90 x 45 (to match stud width)
maximum loaded dimension 4.9 m (Light roof, 600 stud spacing, 900 Truss
spacing, loaded dimension 3.5 m, anywhere in span)
Bottom Plates, External wall – Table 8.17 – assume floor joists are at 600 mm
cs, adopt 90 depth to suit wall framing, select 90 x 45 allowable loaded
dimension 6 m (light roof, 600 stud spacing, loaded dimension 3.5 m)
Top & Bottom plates for non load bearing walls are to be the same depth as
the wall studs and no less than 40 mm thick – select 70 x 45 top & bottom
plates
ADOPT – Top & bottom plate of external walls to be SG8 90 x 45. All internal
wall top & bottom plates to be 70 x 45
Lintels -
Consider lintel in dining room say over a ranchslider – span 2.4 m, carrying a
light roof with loaded dimension 3.5 m
Table 8.9, light roof with loaded dimension 3.5 – select 2 / 190 x 45 (VSG8)
for a span of 2.4 m (capacity 2.5 m)
______________________________________________________________
__
34
ROOFS –
Systems to resist vertical loads -
Roof structure consist of rafters or trusses supported by load bearing walls

designed to transfer vertical permanent and imposed loads and horizontal
loads through to the supporting structure braced system
Modern construction generally provides proprietary trusses which can span

large distances obviating the need for internal supporting walls – this provides
the ability to change partitioning at will during the life of the structure without
effecting the roof structure (although bracing may need to be addressed
where preserving the status quo may be needed)
Rafters are used where there is no ceiling space provided or for short span
systems where sawn timber can be used economically. Roof systems using
rafters are shown in the sketches including systems for lateral bracing
Table 10.1 Provides guidance for Rafter Sizes at various spacings and
various weight of roofs under various wind conditions. A column also defines
the fixing type for the rafters including skew nails and wire dogs
Example : Referring to our Roof examples (after Page 20)
1. Say SG8 rafters (instead of trusses) spanning 3.6 m to a central ribbon

board midspan of building at 900 cs– roof is metal (light) in a medium
wind zone – Table 10.1 – select 190 x 45 (2.8m * 1.3 = 3.64) – Fixing
Type E
2. Rafters at 600 cs supporting heavy roof spans 6.0 m in a low wind
zone – Table 10.1 – select 240 x 70 (5.1 * 1.3 = 6.6) with Fixing Type E
Purlins – Tables 10.10 and 10.11
Timber members which span between rafters or trusses often constructed on

their flat and supporting directly the cladding be it metal roofing, ply sarking for
tiles etc. The spacing of purlins may depend on the span capability of the
roofing and checks need to be made with the manufacturer for the location
(wind zone) to ensure this is not exceeded. From Table 10.10 or 10.11, given
the span between support members and the spacing, a selection can be
made of a purlin size on its flat or “on its edge” for various wind zones and
their fixings. Ensure your conclusions or sketches show the purlins (or clearly
call up in their description) in their correct orientation, on their flat or on their
edge.
______________________________________________________________
__
35
Example of Purlin Selection –
Given roof trusses at 1200cs for a dwelling in Wellington (Wind Zone High),
select a purlin size and fixings for the roof using standard corrugated metal
roofing (max span 900 mm)
From Table 10.10 adopt 90 by 45 Purlins on their flat at 900 cs spanning 1200
mm fixing Type U
Systems to resist lateral loads –
Roofs and ceiling systems are required to transfer wind and seismic lateral
forces into the braced wall systems. In the plane of the roof diagonal strip
bracing can be placed, in hipped roofs the hip provides in built diagonal
bracing and sometimes in the roof space diagonal bracing is required
In some instances a diaphragm may be required at the ceiling level – refer
Sketches
Diaphragms at ceiling or floor level transfer horizontal loads to adjacent
bracing elements. Occasionally in a structure with large open areas the
opportunity to have bracing elements at 5 or 6 m intervals is impossible. In
this instance the Standard allows the bracing elements to be further apart
provided a ceiling or floor diaphragm is designed. The standard provides
guidance for the fixing, size and shape of ply or particle board diaphragms to
comply with structural requirements.
______________________________________________________________
__
36
Civil 713 Section 6.2 Non Specific Design Light Steel 20/03/2020
Framing
LSF HOUSE
ANKARA, TURKEY
Introduction to
LSF Housing
and
Slides from Civil 718 are used with
permission from Pouya Pouladi
Source: www.scottsdalesteelframes.com/gallery/residential/steel‐frame‐house‐ankara‐turkey‐1
© 2019 Pouya Pouladi 1
Construction Types in CFS Housing
Stick build constructed entirely or largely on‐site
Source: deliciassalvadorenas.com/xplijf22962/sLYuFx23001/
Panelised Off‐site panel construction in a controlled environment
Modular Off‐site panel construction in a controlled environment
Source: ww.steelframing.org/PDF/quicklinks/SFA_Framing_Guide_07.pdf
Combination Combination of above items
© 2019 Pouya Pouladi 2
Section 6.2 Page 1
Civil 713 Section 6.2 Non Specific Design Light Steel 20/03/2020
Framing
NASH?
NASH NZ started 1989 as a non‐profit industry association
To promote the concept and use of LSF throughout NZ
To be the peak body for the industry
To foster business relationships with decision makers and others
To exchange information about the steel framed housing industry
To develop standards
To advocate with government and others for the industry
Stories
Height
Buildings Plan
Covered by
Roof & Wall
Slope
NASH STANDARD
PART 2 2019 PAGE 9
Gravity Load
Wind Zone
Other
Source: NASH Standard Part Two: 2019, pp. 10 4
Section 6.2 Page 2
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CIVIL AND ENVIRONMENTAL ENGINEERING

Civil 713: 2020

STRUCTURES AND DESIGN 4

These Volume 2 notes contain the

1. Overview of steel member design

Department of Civil & Environmental Engineering

STRUCTURES AND DESIGN 4

Section 1.1: Steel Refresher Notes

Section 1.2: Steel Design Review

Department of Civil & Environmental Engineering

713 Structures and Design 4

By Associate Professor G Charles Clifton,

Structural Steel Properties

The following slides give a very brief review of the

Basic Properties of Steel: 1

Basic Properties of Steel: 2

700Q = quenched and

Structural Steel Mechanical Properties

For New Zealand, Australian, UK/European

See more detailed information in NZS 3404 Table C2.2.1

For I-sections, which fy to use in design?

Two common strengths:

Bolt Modes for Structural Bolts

For property class 8.8 or 10.9

Fully Tensioned Bolts

Design modes for shear transfer through

In the ultimate limit state:

Incomplete Penetration Butt Weld

Complete Penetration Butt Weld

Hold Down Bolts into Concrete

Clause 4.2 of NZS 3404

Sway and Braced

NZS 3404 recognises three forms of

Moment-Rotation Curves for Different

Illustration of connection in simple

Illustrations of connections in semi-rigid

Illustration of connections in rigid

Clause 4.3 of NZS 3404

2 Dimensional and 3 Dimensional

Design Actions in Beams of

Floor by floor analysis of rigidly connected floor

Model used showing variable imposed

Arrangements of Imposed (Live) Loads

Arrangements of Imposed Loads

This applies for

Span Length: beam

= design shear force at end of beam

Span Length: column

= design shear force in column

Clauses 4.4, 4.5 and 4.6 of NZS 3404

NZS 3404 Permits Three Methods

More detail on the first two now follows

Elastic Analysis (Clause 4.5) 1 of 2

• second order effects must be considered by:

• second order effects relate to a loss of elastic stiffness

Elastic Analysis With Redistribution

Example for Propped Cantilever

Elastic distribution Fully redistributed elastic Plastic collapse

Second Order Effects