DM 7.1 - Navfac - Soil Mechanics - 2022

UFC 3-220-10
1 February 2022
UNIFIED FACILITIES CRITERIA (UFC)
SOIL MECHANICS
(DM 7.1)
APPROVED FOR
PUBLIC RELEASE; DISTRIBUTION UNLIMITED
UFC 3-220-10
1 February 2022
SOIL MECHANICS (DM 7.1)
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U.S. ARMY CORPS OF ENGINEERS

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Change No. Date Location
This UFC supersedes UFC 3-220-10N, dated 8 JUNE 2005.

UFC 3-220-10
1 February 2022
FOREWORD
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AUTHORIZED BY:
CHRISTINE T. ALTENDORF, PhD, R. DAVID CURFMAN, P.E., SES

P.E., SES Chief Engineer
Chief, Engineering and Construction Naval Facilities Engineering Command
U.S. Army Corps of Engineers
NANCY J. BALKUS, P.E., SES MICHAEL McANDREW

Deputy Director of Civil Engineers Deputy Assistant Secretary of Defense
DCS/Logistics, Engineering & (Construction) Office of the Secretary of
Force Protection (HAF/A4C) Defense
HQ United States Air Force
UFC 3-220-10
1 February 2022
REVISION SUMMARY
Document: UFC 3-220-10, Soil Mechanics
Superseding: UFC 3-220-10N, Soil Mechanics
Description: “Soil Mechanics” or DM 7.1 (UFC 3-220-10N) has been a valuable legacy
document in geotechnical engineering for 50 years. Revisions to the document
occurred in 1982, 1986, and 2005; but for the most part; the document has remained
substantially unchanged since the original publication in 1971. DM 7.1 has been on the
bookshelf of many civil engineers, it has been used in many graduate and
undergraduate soil mechanics classed attended by generations of geotechnical
engineering students, and charts and correlations from the document have been cited in
numerous textbooks and research papers. Currently, it can be found in electronic
format at a variety of sites on the internet.
The lasting value of DM 7.1 is attributed to its success in distilling geotechnical

engineering design procedures, particularly into graphical examples that are easy to
follow and understand. The manual also contains correlations to estimate engineering
properties of soil and rock that have become ubiquitous in engineering practice.
Although the manual continues to be a part of everyday engineering, changes in the
profession necessitate a substantial update of DM 7.1. The manual was initially written
when the slide rule was the main calculation tool of engineers. Subsequent revisions
predate the widespread use of personal computer software tools that are used by every
practicing engineer. The manual also predates the global use of the internet as a
means to gather pertinent information and to transfer data and documents. In addition,
there have been many new methods of testing, exploration, and analysis that have been
developed since the publication of the original manual.
This current revision was undertaken with an emphasis on retaining the elements that
were responsible for the lasting value of DM 7.1. Graphical examples of engineering
solutions, both old and new, are found throughout the chapters. A new chapter has
been written that focuses on geotechnical engineering correlations. Details about
computer solutions and numerical modeling tools have been added to the manual.
Owing to the rapid changes that occur in geotechnical engineering software tools and
internet addresses, the authors have tried to minimize the number of URLs and the
names of specific software packages in the text. Appendix B contains a listing of
software packages available at the time of publication (2021), along with vendor contact
information, with the intention that this appendix can be updated periodically in the
future.
In accordance with MIL-STD-3007 and UFC 1-300-01, Criteria Format Standard, this
UFC varies in format from traditional UFC format requirements. It was approved for
variation in format as required in UFC 1-300-01.
UFC 3-220-10
1 February 2022
UFC 3-220-10
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This Page Intentionally Left Blank

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TABLE OF CONTENTS
FOREWORD .............................................................................................................................i
TABLE OF CONTENTS ...............................................................................................................i
LIST OF FIGURES .....................................................................................................................x
LIST OF TABLES..................................................................................................................... xxi
CHAPTER 1 IDENTIFICATION AND CLASSIFICATION OF SOIL AND ROCK ...................... 1
1-1 INTRODUCTION. ............................................................................................... 1
1-1.1 Scope. ................................................................................................ 1
1-2 SOIL DEPOSITS................................................................................................. 1
1-2.1 Geologic Origin and Mode of Occurrence. .......................................... 1
1-3 SOIL VISUAL DESCRIPTION, IDENTIFICATION, AND CLASSIFICATION. ...... 5
1-3.1 Definitions. .......................................................................................... 5
1-3.2 Visual Description and Identification (ASTM D2488). .......................... 6
1-3.3 Unified Soil Classification System (ASTM D2487). ............................11
1-3.4 Soil Classification for Highways (AASHTO). ......................................15
1-3.5 Other Classification Systems. ............................................................16
1-3.6 Common Soil and Rock Names. ........................................................17
1-4 ROCK VISUAL DESCRIPTION, AND CLASSIFICATION. .................................23
1-4.1 Definitions. .........................................................................................23
1-4.2 Visual Classification. ..........................................................................24
1-4.3 Classification by Field and Laboratory Measurements. ......................29
1-4.4 Rock Mass Classification Systems.....................................................31
1-5 SPECIAL MATERIALS.......................................................................................35
1-5.1 Expansive Soils. ................................................................................35
1-5.2 Collapsing Soils. ................................................................................39
1-5.3 Frost Susceptibility and Permafrost....................................................41
1-5.4 Limestone and Related Materials. ......................................................44
1-5.5 Coral and Coral Formation. ................................................................46
1-5.6 Quick Clays. ......................................................................................47
1-5.7 Other Materials and Considerations. ..................................................47
1-6 SUGGESTED READING. ..................................................................................51
1-7 NOTATION. .......................................................................................................51
CHAPTER 2 FIELD EXPLORATION, TESTING, AND INSTRUMENTATION .........................53
2-1 INTRODUCTION. ..............................................................................................53
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2-1.1 Scope. ...............................................................................................53
2-1.2 Planning for Field Investigations. .......................................................53
2-2 PUBLISHED REFERENCE MATERIALS. ..........................................................55
2-2.1 Previous Investigations. .....................................................................55
2-2.2 Published Geologic and Hydrogeologic Maps. ...................................56
2-3 REMOTE SENSING DATA METHODS. ............................................................56
2-3.1 Sources. ............................................................................................56
2-3.2 Utilization. ..........................................................................................56
2-4 GEOPHYSICAL METHODS. .............................................................................60
2-4.1 Utilization and Applications. ...............................................................60
2-4.2 Advantages and Limitations. ..............................................................61
2-5 SOIL AND ROCK EXPLORATION METHODS. .................................................61
2-5.1 Drilling and Boring Methods. ..............................................................61
2-5.2 Test Pits and Test Trenches. .............................................................67
2-5.3 Other Exploratory Techniques. ..........................................................67
2-6 SAMPLING. .......................................................................................................68
2-6.1 Soil Sampling. ....................................................................................69
2-6.2 Rock Sampling...................................................................................73
2-6.3 Offshore Sampling. ............................................................................75
2-6.4 Field Logging and Boring Logs. .........................................................76
2-7 PENETRATION RESISTANCE TESTS. ............................................................79
2-7.1 Standard Penetration Test (SPT). ......................................................79
2-7.2 Cone Penetrometer Tests (CPT). ......................................................81
2-7.3 Flat Plate Dilatometer. .......................................................................84
2-7.4 Dynamic Cone Penetrometer. ............................................................87
2-8 GROUNDWATER MEASUREMENTS. ..............................................................89
2-8.1 Types of Standpipe Piezometer. ........................................................89
2-8.2 Multiple or Nested Installations. .........................................................92
2-8.3 Measurement of Groundwater Levels. ...............................................92
2-8.4 Detection of Combustible Gases ........................................................93
2-9 MEASUREMENT OF SOIL AND ROCK PROPERTIES IN SITU. ......................94
2-9.1 Strength and Deformation Properties of Soil. .....................................94
2-9.2 Hydraulic Conductivity of Soil...........................................................101
2-9.3 Engineered Fill and Earthworks. ......................................................103
2-9.4 Rock Properties. ..............................................................................110
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2-10 FIELD INSTRUMENTATION AND MONITORING. ..........................................114
2-10.1 Operating Concepts for Geotechnical Monitoring Instruments. ........115
2-10.2 Linear Deformation Measurements. .................................................116
2-10.3 Angular Displacement Measurements. ............................................121
2-10.4 Pore Pressure and Water Pressure Measurements. ........................121
2-10.5 Earth Pressure Measurements. .......................................................125
2-10.6 Load Measurements. .......................................................................125
2-10.7 Temperature Measurements. ...........................................................126
2-10.8 Vibration Measurements. .................................................................126
2-10.9 Field Applications for Instrumentation. .............................................127
2-11 SUGGESTED READING. ................................................................................129
2-12 NOTATION. .....................................................................................................129
CHAPTER 3 LABORATORY TESTING ................................................................................131
3-1 INTRODUCTION. ............................................................................................131
3-1.1 Scope. .............................................................................................131
3-1.2 Evolution of Laboratory Test Procedures. ........................................131
3-1.3 Laboratory Certification. ...................................................................132
3-2 LABORATORY TESTS ON SOILS. .................................................................133
3-2.1 Sample Selection. ............................................................................133
3-2.2 Index Property Tests. .......................................................................136
3-2.3 Compaction Tests. ...........................................................................142
3-2.4 Strength Tests. ................................................................................143
3-2.5 Dynamic Tests. ................................................................................159
3-2.6 Compressibility Tests. ......................................................................163
3-2.7 Hydraulic Conductivity (Permeability) Tests. ....................................170
3-3 LABORATORY TESTS ON ROCK...................................................................172
3-3.1 Unconfined Compression Test (ASTM D7012). ...............................172
3-3.2 Split Cylinder Test (ASTM D3967). ..................................................173
3-3.3 Rock Direct Shear Test (ASTM D5607). ..........................................173
3-3.4 Point Load Test (ASTM D5731). ......................................................174
3-4 OTHER SOIL AND ROCK TESTS. ..................................................................175
3-6 NOTATION. .....................................................................................................176
CHAPTER 4 DISTRIBUTION OF STRESSES ......................................................................180
4-1 INTRODUCTION. ............................................................................................180
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4-1.1 Scope. .............................................................................................180
4-1.2 State of Stress. ................................................................................180
4-2 STRESS CONDITIONS AT A POINT...............................................................180
4-2.1 Stress Conditions in Soil. .................................................................180
4-2.2 Mohr Circle of Stress. ......................................................................185
4-3 ELASTIC SOLUTIONS FOR STRESSES DUE TO APPLIED LOADS. ............185
4-3.1 Use and Applicability........................................................................185
4-3.2 Semi-Infinite Elastic Conditions........................................................185
4-3.3 Layered or Anisotropic Foundations.................................................196
4-4 SHALLOW PIPES AND CONDUITS. ...............................................................198
4-4.1 General. ...........................................................................................198
4-4.2 Vertical Loads on Rigid Pipe. ...........................................................198
4-4.3 Vertical Loads on Flexible Pipe. .......................................................199
4-4.4 Long Span Metal Culverts. ...............................................................201
4-5 DEEP UNDERGROUND OPENINGS. .............................................................201
4-5.1 General Factors. ..............................................................................201
4-5.2 Openings in Rock. ...........................................................................202
4-5.3 Loads on Underground Openings in Rock. ......................................203
4-5.4 Openings in Soft Ground (Soil). .......................................................206
4-5.5 Pressure on Vertical Shafts..............................................................210
4-6 NUMERICAL SOLUTIONS FOR STRESSES IN SOIL.....................................213
4-6.1 Numerical Analysis Types. ...............................................................213
4-6.2 Linear Elastic Stress Analysis. .........................................................214
4-6.3 Nonlinear Elastic Stress Analysis. ....................................................214
4-6.4 Numerical Modeling Best Practice. ..................................................216
4-6.5 Evaluation of Stress Due to Applied Loads. .....................................217
4-6.6 Evaluation of Stress within Embankments and Slopes. ....................217
4-8 NOTATION. .....................................................................................................218
CHAPTER 5 ANALYSIS OF SETTLEMENT AND VOLUME EXPANSION ...........................222
5-1 INTRODUCTION. ............................................................................................222
5-1.1 Scope. .............................................................................................222
5-1.2 Occurrence of Settlement. ...............................................................222
5-1.3 Occurrence of Heave. ......................................................................223
5-1.4 Applicability. .....................................................................................223
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5-2 MECHANICS OF CONSOLIDATION. ..............................................................224
5-2.1 Consolidation Process. ....................................................................224
5-2.2 Initial Vertical Stress State. ..............................................................225
5-2.3 Stress History. .................................................................................225
5-2.4 Evaluation of Existing Conditions. ....................................................227
5-2.5 Change in Vertical Stress. ...............................................................229
5-3 SETTLEMENT CALCULATIONS. ....................................................................230
5-3.1 Basic Formulation. ...........................................................................230
5-3.2 Soil Layers in Settlement Calculations. ............................................230
5-4 SETTLEMENT OF COARSE-GRAINED SOILS. ..............................................231
5-4.1 Short Term Settlement of Coarse-Grained Soil. ...............................231
5-4.2 Long-Term Settlement of Coarse-Grained Soil. ...............................238
5-5 SETTLEMENT OF FINE-GRAINED SOILS......................................................239
5-5.1 Immediate Settlement of Fine-Grained Soils. ...................................239
5-5.2 Primary Consolidation Settlement of Fine-Grained Soils. .................240
5-5.3 Time Rate of Primary Consolidation.................................................247
5-5.4 Secondary Compression of Fine-Grained Soils................................257
5-5.5 Organic Soils and Peat. ...................................................................258
5-6 DIFFERENTIAL AND TOLERABLE SETTLEMENT. ........................................259
5-6.1 Differential Settlement......................................................................259
5-6.2 Tolerable Settlement. .......................................................................260
5-6.3 Differential Settlement of Mat Foundations. .....................................262
5-7 METHODS OF CONTROLLING SETTLEMENT. .............................................263
5-7.1 Removal or Displacement of Compressible Soils. ............................263
5-7.2 Balancing Load by Excavation. ........................................................265
5-7.3 Preconsolidation by Surcharge. .......................................................265
5-7.4 Vertical Drains. ................................................................................268
5-8 VOLUME EXPANSION. ...................................................................................275
5-8.1 Mechanics of Volume Expansion. ....................................................275
5-8.2 Effects of Volume Expansion. ..........................................................277
5-8.3 Estimates of Heave or Swell Pressure. ............................................277
5-8.4 Design in Expansive Soils. ...............................................................279
5-8.5 Construction Practices in Expansive Soils. ......................................281
5-9 HYDROCOMPRESSION. ................................................................................281
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5-11 NOTATION. .....................................................................................................282
CHAPTER 6 SEEPAGE AND DRAINAGE ............................................................................286
6-1 INTRODUCTION. ............................................................................................286
6-1.1 Scope. .............................................................................................286
6-1.2 Background. ....................................................................................286
6-2 SEEPAGE ANALYSES. ...................................................................................286
6-2.1 Hydraulic Head. ...............................................................................286
6-2.2 Darcy’s Law and One-Dimensional Flow..........................................288
6-2.3 Two-Dimensional Seepage. .............................................................290
6-2.4 Flow Nets.........................................................................................290
6-2.5 Closed-Form Equations. ..................................................................294
6-2.6 Numerical Seepage Analysis. ..........................................................295
6-3 HYDRAUIC CONDUCTIVITY (COEFFICIENT OF PERMEABILTY). ...............300
6-3.1 Laboratory Testing. ..........................................................................300
6-3.2 Field Testing. ...................................................................................300
6-3.3 Empirical Relationships for Hydraulic Conductivity. ..........................301
6-3.4 Anisotropy........................................................................................306
6-4 INTERNAL EROSION. .....................................................................................307
6-4.1 Heave. .............................................................................................307
6-4.2 Erosion and Stoping. .......................................................................308
6-4.3 Internal Instability. ............................................................................310
6-5 SEEPAGE AND INTERNAL EROSION MITIGATION METHODS. ..................310
6-5.1 Problems and General Strategies. ...................................................310
6-5.2 Seepage Barriers. ............................................................................312
6-5.3 Filters and Drains.............................................................................322
6-6 DEWATERING. ...............................................................................................341
6-6.1 Collection and Sump. .......................................................................341
6-6.2 Wellpoint Systems. ..........................................................................341
6-6.3 Extraction Wells. ..............................................................................343
6-8 NOTATION. .....................................................................................................346
CHAPTER 7 SLOPE STABILITY ..........................................................................................349
7-1 INTRODUCTION. ............................................................................................349
7-2 TYPES OF SLOPES AND MODES OF FAILURE. ...........................................349
7-3 DEFINITION OF FACTOR OF SAFETY...........................................................352
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7-4 METHODS OF ANALYSIS OF SOIL SLOPES. ................................................352
7-4.1 Limit Equilibrium Analysis. ...............................................................353
7-4.2 Finite Element Analysis of Slopes. ...................................................355
7-4.3 Limit Analysis. ..................................................................................355
7-5 WATER PRESSURE EFFECTS. .....................................................................358
7-5.1 Incorporating Water Pressures in Computer Analyses. ....................358
7-5.2 Seepage Forces. .............................................................................360
7-6 STRENGTH MODELS AND ANALYSIS CASES..............................................360
7-6.1 End of Construction (Short Term). ...................................................361
7-6.2 Cut Slope in Clay. ............................................................................361
7-6.3 Steady State Seepage in Dams. ......................................................361
7-6.4 Stabilizing Berm for Failed Slope. ....................................................363
7-6.5 Other Analysis Cases. .....................................................................363
7-6.6 Back-Analysis of Slopes. .................................................................364
7-6.7 Evaluation of Slope Stability Results. ...............................................364
7-6.8 Slope Stability Charts.......................................................................365
7-7 SLOPE STABILIZATION..................................................................................367
7-8 REQUIRED FACTOR OF SAFETY FOR SOIL SLOPES. ................................368
7-9 MECHANICALLY STABILIZED EARTH SLOPES. ...........................................369
7-9.1 Applications of MSE. ........................................................................370
7-9.2 Reinforced Slope Materials. .............................................................371
7-9.3 Geosynthetic Reinforcement Strength. ............................................372
7-9.4 Soil-Geosynthetic Interaction. ..........................................................375
7-9.5 Analysis and Design of Reinforced Slopes.......................................375
7-9.6 Required Factor of Safety for MSE Slopes.......................................378
7-10 ROCK SLOPE STABILITY. ..............................................................................378
7-10.1 Modes of Rock Slope Failure. ..........................................................378
7-10.2 Mechanics of a Sliding Block. ..........................................................381
7-10.3 Plane Failure. ..................................................................................381
7-10.4 Plane Failure Analyses. ...................................................................383
7-10.5 Wedge Failure. ................................................................................384
7-10.6 Toppling Failure. ..............................................................................386
7-10.7 Circular Failure. ...............................................................................387
7-10.8 Rock Slope Stabilization and Protection. .........................................387
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7-12 NOTATION. .....................................................................................................393
CHAPTER 8 CORRELATIONS FOR SOIL AND ROCK........................................................396
8-1 INTRODUCTION. ............................................................................................396
8-2 EFFECTIVE STRESS (DRAINED) SHEAR STRENGTH. ................................397
8-2.1 Coarse-Grained Soils.......................................................................397
8-2.2 Fine-Grained Soils. ..........................................................................410
8-3 UNDRAINED SHEAR STRENGTH. .................................................................421
8-3.1 Correlations with Index Properties. ..................................................421
8-3.2 Correlations with Stress History. ......................................................424
8-3.3 Correlations with Cone Penetration Test. .........................................428
8-3.4 Correlations with Standard Penetration Test. ...................................429
8-3.5 Correlations with Dilatometer. ..........................................................431
8-4 CONSOLIDATION PARAMETERS. .................................................................432
8-4.1 Compression and Recompression Indices – Fine-Grained. .............432
8-4.2 Compression and Recompression Indices – Coarse-Grained. .........443
8-4.3 Constrained Modulus. ......................................................................444
8-4.4 Coefficient of Secondary Compression. ...........................................448
8-4.5 Coefficient of Consolidation. ............................................................450
8-5 ELASTIC PARAMETERS. ...............................................................................450
8-5.1 Definitions. .......................................................................................450
8-5.2 Undrained Young’s Modulus of Fine-Grained Soils. .........................452
8-5.3 Drained Young’s Modulus of Coarse-Grained Soils. ........................456
8-6 CALIFORNIA BEARING RATIO (CBR). ...........................................................457
8-6.1 Correlations with Index and Compaction Properties. ........................457
8-6.2 Correlations with Dynamic Cone Penetration. ..................................459
8-7 HYDRAULIC CONDUCTIVITY.........................................................................461
8-7.1 Typical Values. ................................................................................461
8-7.2 Correlations for Coarse-Grained Soils. ............................................461
8-7.3 Correlations for Fine-Grained Soils. .................................................465
8-8 SHEAR WAVE VELOCITY. .............................................................................467
8-8.2 Correlations with Cone Penetration Test. .........................................470
8-10 NOTATION. .....................................................................................................471
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APPENDIX A. REFERENCES ................................................................................................475
APPENDIX B. LIST OF COMPUTER PROGRAMS ................................................................499
APPENDIX C. SYMBOLS USED IN GEOTECHNICAL ENGINEERING..................................513
APPENDIX D. GLOSSARY .....................................................................................................529
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LIST OF FIGURES
Figure 1-1 Typical Angularity of Bulky Grains (after Sowers 1979)........................ 8

Figure 1-2 Plasticity Chart ................................................................................... 12
Figure 1-3 Soil Property Variation with Liquid Limit and Plasticity ....................... 13
Figure 1-4 Rock Strength Characterization (after Broch and Franklin 1972) ....... 31
Figure 1-5 GSI Selection Chart for Jointed Rock (after Marinos et al. 2007) ...... 33
Figure 1-6 GSI Selection Chart for Heterogeneous Rock
(after Marinos et al. 2007) .................................................................. 34
Figure 1-7 Expansive Soils in the United States (Nelson and Miller 1992) .......... 36
Figure 1-8 Soil Expansion Prediction (after Holtz et al. 2011) ............................. 37
Figure 1-9 Collapsibility Based on In situ Dry Density and Liquid Limit
(after Holtz et al. 2011) ....................................................................... 40
Figure 1-10 Design Charts for Predicting Collapse Behavior of Soils
(after Ayadat and Hanna 2007b) ........................................................ 40
Figure 1-11 Maximum Depths (in meters) of Frost Penetration in the
Continental United States (NOAA 1978) ............................................ 42
Figure 1-12 Rates of Heave in Laboratory Freezing Tests on Remolded Soils
(U.S. Department of the Army 1984) .................................................. 42
Figure 1-13 Karst and Potential Karst Areas in Soluble Rocks in the Contiguous
United States (USGS 2014) ............................................................... 45
Figure 2-1 Schematic of Various Drilling Techniques for Soil and Rock
(after NCHRP 2018 and Mayne 2012)................................................ 64
Figure 2-2 Cross Section of Split Barrel Sampler ................................................ 70
Figure 2-3 Cross Section of Shelby Tube Sampler with Ball-check Valve Head.. 70
Figure 2-4 Cross Section of a Stationary or Fixed Piston Sampler ...................... 72
Figure 2-5 Rock Core Samplers (after NCHRP 2018) ......................................... 74
Figure 2-6 Example Geotechnical Boring Log ..................................................... 78
Figure 2-7 Standard Penetration Test (after NCHRP 2018 and Mayne 2012) ..... 80
Figure 2-8 CPT - Example Test Record and Equipment...................................... 84
Figure 2-9 Nine Zone (Normalized) Soil Behavioral Chart for CPT
(after Robertson 2009 and NCHRP 2018) .......................................... 84
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Figure 2-10 Flat Plate Dilatometer Test Schematic
(after NCHRP 2018 and Mayne 2012)................................................ 85
Figure 2-11 Flat Plate Dilatometer and Control Unit (Marchetti et al. 2006) .......... 86
Figure 2-12 Schematic of Dynamic Cone Penetrometer (DCP) Equipment
(after Webster et al. 1992) .................................................................. 88
Figure 2-13 Open Piezometers .............................................................................. 90
Figure 2-14 Schematic of Pressuremeter Test
(after NCHR 2018 and Mayne 2012) .................................................. 96
Figure 2-15 Typical Result and Characteristic Pressures from Pressuremeter
Test (after FHWA 2002) ..................................................................... 97
Figure 2-16 Example Result from Self-boring Pressuremeter Test in Clay
(after Windle and Wroth 1977) ........................................................... 98
Figure 2-17 Schematic of Vane Shear Test (after Mayne 2012) .......................... 100
Figure 2-18 Schematic of Equipment and Process to Perform a Sand Cone
Test (after Dunn 2017) ..................................................................... 105
Figure 2-19 Schematic of Equipment to Perform Water Balloon Test
(after Dunn 2017) ............................................................................. 105
Figure 2-20 Schematic of Drive Cylinder (after ASTM D2937) ............................ 106
Figure 2-21 Schematic of Nuclear Gauge in Direct Transmission Mode
(after NRC 1996) .............................................................................. 107
Figure 2-22 Example Plate Load Test Result on Intact Limestone
(after NCHRP 2017) ......................................................................... 111
Figure 2-23 Double Packer Set-up to Conduct Five-step Lugeon test
(after Clayton et al. 1995) ................................................................. 114
Figure 2-24 Electrical Crack Gauge and Reference Pins (after Dunnicliff 1993) . 118
Figure 2-25 Surface Settlement (a) Plate or (b) Platform (after Dunnicliff 1993) . 119
Figure 2-26 Liquid Level System to Continuously Profile Settlements
(after Dunnicliff 1993) ....................................................................... 120
Figure 2-27 Borehole Extensometer (after Dunnicliff 1993) ................................. 120
Figure 2-28 Slope Inclinometers – (a) Manual System, (b) Measurement Principle,
and (c) In-Place Inclinometer System (after Dunnicliff 1993)............ 122
Figure 2-29 Dual-tube Hydraulic Piezometer in Embankment Dam
Figure 2-30 Example of Electrical Diaphragm Piezometer Transducer
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Figure 2-31 Estimated Hydrodynamic Lag Time for Various Piezometers
and Wells (after Dunnicliff 1993) ...................................................... 125
Figure 3-1 Basic Elements for a Consolidated Drained Direct Shear Test
Along with Example Data Collected for One Test Specimen ............ 146
Figure 3-2 Basic Elements of a Consolidated Drained Triaxial Test Along
with Example Data Collected for One Test Specimen ...................... 148
Figure 3-3 Basic Elements of a Consolidated Undrained Triaxial Test Along
with Example Data Collected for One Test Specimen ...................... 149
Figure 3-4 Basic Elements of a Ring Shear Test Along with Sample Data ........ 151
Figure 3-5 Undrained Shear Strength Envelopes for Saturated and Partially
Saturated Soils ................................................................................. 152
Figure 3-6 Example Distribution of Undrained Strength Versus Depth
Relationship for a Hypothetical Saturated Clay ................................ 153
Figure 3-7 Basic elements of a UU Test Apparatus with Sample Data for a
Single Test ....................................................................................... 155
Figure 3-8 Basic Elements of the Direct Simple Shear Test (ASTM D6528) ..... 157
Figure 3-9 Laboratory Miniature Vane Shear Apparatus ................................... 158
Figure 3-10 Fall Cone Apparatus ......................................................................... 160
Figure 3-11 Loading Function and Stresses Applied for a Cycle of Loading
in a Cyclic Triaxial Test for a Cyclic Stress Ratio of 0.2.................... 162
Figure 3-12 Basic Information Obtained from a Consolidation Test..................... 165
Figure 3-13 Fixed-Ring and Floating-Ring Consolidometers ............................... 166
Figure 3-14 Basic Elements of a Constant Rate of Strain Consolidation Test ..... 168
Figure 3-15 Volume Change of Soil as a Function of Stress at Inundation.......... 169
Figure 3-16 Specimen Container for Rock Direct Shear Test
(after ASTM D5607) ......................................................................... 173
Figure 3-17 Rock Direct Shear Apparatus for High Normal and Shear Loads ..... 174
Figure 3-18 Point Load Apparatus for Rock Index Testing .................................. 175
Figure 4-1 Calculation of Vertical Stresses for Hydrostatic Conditions .............. 182
Figure 4-2 Variation in Contact Pressure – a) Rigid Foundation and
b) Completely Flexible Foundation ................................................... 184
Figure 4-3 Mohr Circle Relationships................................................................. 186
Figure 4-4 Vertical Stress Contours from Strip and Square Loaded Areas –
Boussinesq ....................................................................................... 190
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UFC 3-220-10
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Figure 4-5 Influence Factors for a Rectangular Loaded Area – Boussinesq...... 191
Figure 4-6 Influence Factors for a Circular Loaded Area – Boussinesq
(after Ahlvin and Ulery 1962, Poulos and Davis 1974) ..................... 192
Figure 4-7 Influence Factors for Embankment Loading – Boussinesq
(after Poulos and Davis 1974) .......................................................... 193
Figure 4-8 Use of Superposition to Determine Change in Vertical Stress ......... 194
Figure 4-9 Stress Distribution Examples............................................................ 195
Figure 4-10 Influence Factors for a Rectangular Loaded Area – Westergaard .... 197
Figure 4-11 Loading Mechanisms for Soft Ground Tunneling ............................. 209
Figure 4-12 Radial Stress at the Sides of a Vertical Shaft in Sand
(based on Cheng et al. 2007) ........................................................... 212
Figure 4-13 Parameters Used in Duncan-Chang Model ...................................... 216
Figure 5-1 Initial Vertical Stresses for a) Hydrostatic and b) Artesian Pore
Water Pressure Conditions............................................................... 225
Figure 5-2 Vertical Stress History Examples...................................................... 226
Figure 5-3 Vertical Stress Profile Cases – Transient ......................................... 227
Figure 5-4 Example Evaluation of Existing Conditions ...................................... 229
Figure 5-5 Three Possible Methods to Define Layers for Homogeneous
Conditions ........................................................................................ 231
Figure 5-6 Elastic Influence Factors for ν = 0.5 for (a) µ1 (after Giroud 1972)
and (b) µ0 (after Burland 1970) ........................................................ 233
Figure 5-7 Influence Diagram and Modulus Correlation for Schmertmann
CPT Method (Schmertmann 1970, Schmertmann et al. 1978) ......... 236
Figure 5-8 Comparison of Settlement Calculation Methods for Coarse-Grained
Soils based on SPT Blow Count (after Tan and Duncan 1991) ........ 238
Figure 5-9 Creep Factors for Settlement of Coarse-Grained Soils
( t0 = 0.1 year) (after Schmertmann 1970, Terzaghi et al. 1996) ...... 239
Figure 5-10 Correlation of Normalized Undrained Modulus and
Overconsolidation Ratio (after Duncan and Buchignani 1987) ......... 240
Figure 5-11 Consolidation Behavior based on (a) Void Ratio and
(b) Vertical Strain .............................................................................. 241
Figure 5-12 Common Compression Index Correlations ....................................... 244
Figure 5-13 Primary Consolidation Example ....................................................... 245
Figure 5-14 Vertical Movements during a Typical Construction Process ............. 246
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UFC 3-220-10
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Figure 5-15 Correction Factor for Overconsolidated Clays and Loads of
Limited Lateral Extent (after Leonards 1976) ................................... 247
Figure 5-16 Degree of Consolidation for Instantaneous Uniform Loading and
One-Dimensional Flow ..................................................................... 248
Figure 5-17 Degree of Compression and Excess Pore Pressure ........................ 249
Figure 5-18 Effect of Load Geometry on Time Rate of Consolidation
(after Davis and Poulos 1972) .......................................................... 251
Figure 5-19 Effect of Anisotropy on Time Rate of Consolidation
(after Davis and Poulos 1972) .......................................................... 251
Figure 5-20 Degree of Consolidation for Gradual Load Application for Vertical
Drainage (after Olson 1977) ............................................................. 252
Figure 5-21 Determination of Coefficient of Consolidation from Laboratory Data 254
Figure 5-22 Determination of cv from Lab and Field Data ................................... 255
Figure 5-23 Multi-layer Consolidation Example ................................................... 256
Figure 5-24 Calculation of Secondary Compression............................................ 258
Figure 5-25 Components of Settlement (after Duncan and Buchignani 1987,
Ricceri and Soranzo 1985) ............................................................... 259
Figure 5-26 Allowable Deflection Ratios Related to Structural Proportions
(after Burland and Wroth 1974, Wahls 1981) ................................... 262
Figure 5-27 Surcharge Load and Consolidation Required to Eliminate
Settlement under Final Load ............................................................ 267
Figure 5-28 Surcharge Loading Example ............................................................ 268
Figure 5-29 Vertical Drains – (a) Triangular Pattern, (b) Rectangular Pattern,
and (c) Equivalent Cylinder for Theoretical Solutions ....................... 269
Figure 5-30 Degree of Radial Consolidation ........................................................ 272
Figure 5-31 Radial Consolidation with Gradual Loading (after Olson 1977) ........ 273
Figure 5-32 Design Chart for Radial Drainage ..................................................... 274
Figure 5-33 Radial Consolidation Example .......................................................... 275
Figure 6-1 Example of the Components of Hydraulic Head ............................... 287
Figure 6-2 One-Dimensional Flow through Soil ................................................. 288
Figure 6-3 Flow Net for Seepage Through an Isotropic Soil Layer Beneath
an Impermeable Dam ....................................................................... 291
Figure 6-4 Deflection of Flow at a Boundary with Changed Permeability .......... 293
Figure 6-5 Flow Net Example Calculations ........................................................ 295
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Figure 6-6 Variation of Hydraulic Conductivity with Soil Type for Various Unit
Systems (after Freeze and Cherry 1979) ......................................... 301
Figure 6-7 Variation of Hydraulic Conductivity with Fines Content
(after California Department of Water Resources 2013)................... 306
Figure 6-8 Heave – (a) Effective Stress and (b) Total Stress ............................ 308
Figure 6-9 Erosion and Stoping Mechanisms .................................................... 309
Figure 6-10 Internal Instability – (a) Suffusion and (b) Suffosion ......................... 310
Figure 6-11 Required Depth of Penetration of Cutoff Wall-Supported
Excavations in Homogenous Isotropic Sand (after Marsland 1953) . 317
Figure 6-12 Corrections to Required Depth of Penetration of Cutoff Wall-
Supported Excavations for Stratified Sand (after Marsland 1953) .... 318
Supported Excavations in Sand Containing Fine-Grained Layers
(after Marsland 1953) ....................................................................... 319
Figure 6-14 Seepage Blankets and Berms .......................................................... 320
Figure 6-15 Example Base Soil and Filter Gradations ......................................... 323
Figure 6-16 Example Filter Design ...................................................................... 326
Figure 6-17 Subsurface Drain Constructed of Filter Fabric, Drainage Rock,
and a Slotted Pipe ............................................................................ 327
Figure 6-18 Linear Coefficient of Uniformity (after Giroud 2010) ......................... 328
Figure 6-19 Use of Subsurface Interceptor Drains and Blanket Drains for
Roadway Drainage ........................................................................... 331
Figure 6-20 Subsurface Drain with a Two-Stage Filter and a Slotted Pipe .......... 332
Figure 6-21 Drainage of an Aggregate Base Course (after Barber 1959)............ 333
Figure 6-22 Retaining Wall Drainage Alternatives ............................................... 334
Figure 6-23 Seepage into Drainage Trenches Used for Draining Ponded
Areas (after Kirkham 1950 and 1960)............................................... 335
Figure 6-24 Embankment Fill Subdrains.............................................................. 336
Figure 6-25 Toe, Blanket, and Chimney Drains ................................................... 337
Figure 6-26 Outlet Filter Collars ........................................................................... 338
Figure 6-27 Relief Trench Used to Relieve Pressure from Beneath a Blanket
Layer with Low Hydraulic Conductivity (not to scale)........................ 339
Figure 6-28 Relief Well Used to Relieve Pressure from Beneath a Blanket
Layer with Low Hydraulic Conductivity (not to scale)........................ 340
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Figure 6-29 Typical Relief Well Construction ....................................................... 340
Figure 6-30 Calculation of Relief Well Discharge and Spacing
(after USACE 1952).......................................................................... 342
Figure 6-31 Staged Installation of Wellpoints to Lower the Groundwater Table
for a Deep Excavation ...................................................................... 343
Figure 6-32 Drawdown and Pumping Quantities for Single Extraction Wells
and Groups of Extraction Wells (after USACE 1952) ....................... 345
Figure 7-1 Failure Conditions for Different Cross-sections through Natural
Slopes .............................................................................................. 351
Figure 7-2 Failure Conditions in Embankment Foundations and Cut Slopes..... 351
Figure 7-3 Examples of Limit Equilibrium Analysis ............................................ 354
Figure 7-4 Example Slope Stability Analysis using Bishop’s Simplified
Method for an Effective Stress Analysis ........................................... 356
Figure 7-5 Examples of Internal and External Water Pressures in
Slope Stability Analyses ................................................................... 358
Figure 7-6 Approximate Flow Net for Seepage into a Drain Showing the
Difference between the Piezometric Surface and the
Phreatic Surface ............................................................................... 359
Figure 7-7 Analysis Cases for (a) End of Construction for Embankment on
Clay, (b) Cut Slope in Clay, and (c) Levee or Dam in a Condition
of Steady State Seepage.................................................................. 362
Figure 7-8 Stabilizing Berm Used to Increase the Factor of Safety of a
Failed Slope ..................................................................................... 363
Figure 7-9 Chart Solution for Infinite Slope Analysis
(after Duncan et al. 2014) ................................................................. 367
Figure 7-10 Methods of Stabilizing Slopes .......................................................... 368
Figure 7-11 Typical Cross-Section of an MSE Slope ........................................... 370
Figure 7-12 Difference in Usable Land for Walls and Slopes .............................. 371
Figure 7-13 Rock Discontinuity Conditions (after FHWA 1998) ........................... 379
Figure 7-14 Weathering and Weak Rock Conditions (after FHWA 1998) ............ 380
Figure 7-15 Rock Slope with Sliding Block .......................................................... 381
Figure 7-16 Definition of Sloped Surface Orientation Terms ............................... 382
Figure 7-17 Geometry for Plane Failure (after Hoek and Bray 1981) .................. 383
Figure 7-18 Rock Slopes with Tension Cracks (after Hoek and Bray 1981) ........ 384
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UFC 3-220-10
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Figure 7-19 Rock Slope with Sliding Wedge (after Hoek and Bray 1981)............ 385
Figure 7-20 View of Wedge Geometry (after Hoek and Bray 1981)..................... 386
Figure 7-21 Rock Slope Subject to Toppling Failure............................................ 387
Figure 7-22 Rock Slope Stabilization and Protection Measures
(after FHWA 1998) ........................................................................... 388
Figure 8-1 Approximate Relationship between the Effective Stress Friction
Angle and Dry Unit Weight for Various Relative Densities and
Types of Soil..................................................................................... 398
Figure 8-2 Relationship between Effective Stress Friction Angle of Coarse-
Grained Soils and SPT N 60 Value .................................................... 400
Figure 8-3 Relationship between Peak Effective Stress Friction Angle,
Overburden Pressure, and SPT Blow Count for Sands (top)
after Parry (1977) and (bottom) after DeMello (1971) and
Schmertmann (1975) ........................................................................ 401
Figure 8-4 Variation of Effective Stress Friction Angle with N1,60
(after Peck at al. 1974, and Hatanaka and Uchida 1996) ................. 402
Figure 8-5 Relationship between Effective Stress Friction Angle and Cone
Tip Resistance (after Kerisel 1961, Kahl et al. 1968, Melzer 1968,
Muhs and Weiss 1971, and Meyerhof 1976) .................................... 404
Figure 8-6 Estimation of φ ' from a Cone Resistance Profile
Figure 8-7 Relationship between Bearing Capacity Number N q and Peak
Effective Stress Friction Angle from Large Calibration Tests
Figure 8-8 Variation of Peak Effective Stress Friction Angle with σ 'v and Cone
Resistance for Normally Consolidated, Uncemented, Quartz Sands
(after Robertson and Campanella 1983) .......................................... 406
Figure 8-9 Estimation of Relative Density for Normally Consolidated Sands
from Cone Penetration Resistance (after Schmertmann 1978) ........ 407
Figure 8-10 Relationship between Friction Angle and Relative Density based on
Triaxial Compression Tests on North Sea Sands
(after Schmertmann 1975, and Lunne and Kleven 1982) ................ 408
Figure 8-11 Correction for Effects of Overconsolidation on Cone Penetration
Tip Resistance in Sand (after Lunne and Christoffersen 1985,
and Duncan et al. 1989) ................................................................... 408
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UFC 3-220-10
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Figure 8-12 Range of Effective Stress Friction Angle for Clean Sands based
on the Horizontal Stress Index from the Dilatometer Test
(after Marchetti 1997, and Ricceri et al. 2002) .................................. 410
Figure 8-13 Relationship between the Effective Stress Friction Angle of Fine-
Grained Soil and Plasticity Index
(after Gibson 1953, Carter and Bentley 2016) .................................. 411
Figure 8-14 Correlation between φ 'FS and PI based on Triaxial Tests on NC Clays
(after Kenney 1959, Bjerrum and Simons 1960, Ladd et al. 1977) ... 412
Figure 8-15 Relationship between φ 'FS and PI (after Terzaghi et al. 1996) ........ 412
Figure 8-16 Variation of the Fully Softened Friction Angle with Plasticity Index
(after Tiwari and Ajmera 2011) ......................................................... 413
Figure 8-17 Fully Softened Friction Angle based on Mineral Composition
(after Tiwari and Ajmera 2011) ......................................................... 413
Figure 8-18 Correlation between Power Function Parameters aFS and bFS and
Plasticity Index (after Castellanos et al. 2021) .................................. 415
CF × PI (after Castellanos et al. 2021)............................................. 415
Figure 8-20 Correlation between the Residual Friction Angle and Clay-sized
Fraction (after Skempton 1964, 1985) .............................................. 417
Figure 8-21 Residual Friction Angle vs. Plasticity Index – (top) Data Collected
by Voight (1973), and (bottom) Measurements by Bovis (1985) ...... 418
Figure 8-22 Drained Residual Secant Friction Angle as a Function of LL
and CF (after Stark and Hussain 2013) ........................................... 419
Figure 8-23 Residual Shear Strength Power Function Parameters Related to
Plasticity Index (after Castellanos et al. 2021) .................................. 420
Plasticity Index and Clay Fraction (after Castellanos et al. 2021).... 420
Figure 8-25 Relation between Liquidity Index and Undrained Shear Strength of
Remolded Clays (after Skempton and Northey 1952) ...................... 421
Figure 8-26 Relationship between Remolded Undrained Shear Strength and
Liquidity Index (after Terzaghi et al. 1996) ....................................... 422
Figure 8-27 Correlation between Undrained Strength Ratio and Plasticity Index –
Field Vane (after Robertson and Campanella 1984) ........................ 423
Laboratory Testing (after Ladd and DeGroot 2004) .......................... 423
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UFC 3-220-10
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Figure 8-29 Variation of the Undrained Strength Ratio with Liquid Limit
(after Larson 1980) ........................................................................... 424
Figure 8-30 Normalized Undrained Strength Ratio vs. OCR
(after Schmertmann 1978)................................................................ 426
Figure 8-31 Correlation between Undrained Shear Strength, SPT N Value,
and Plasticity Index for Overconsolidated Clays
(after Stroud and Butler 1975) .......................................................... 431
Figure 8-32 Relationship between Undrained Shear Strength and SPT N
(after Terzaghi and Peck 1967, Hara et al. 1974, and Sowers 1979)431
Figure 8-33 Range of Compression Index based on Liquid Limit Predicted by
Correlations ...................................................................................... 437
Figure 8-34 Range of Compression Index based on Initial Void Ratio Predicted
by Correlations ................................................................................. 437
Figure 8-35 Range of Compression Index based on Natural Water Content
Predicted by Correlations ................................................................. 438
Figure 8-36 Correlations for Recompression Index based on Liquid Limit........... 438
Figure 8-37 Range of Recompression Index based on Initial Void Ratio
Figure 8-38 Range of Recompression Index based on Natural Water Content
Figure 8-39 Sensitivity ( St ), In situ Void Ratio, and Compression Index
Relationship (after Leroueil et al. 1983) ............................................ 440
Figure 8-40 Correlation between Modified Compression Index and Water
Content (after Lambe and Whitman 1969) ...................................... 440
Figure 8-41 Sedimentation and Intrinsic Compression Lines
(after Burland 1990).......................................................................... 442
Figure 8-42 Example NC Compression Curves based on a) Intrinsic
Compression Line and b) Secondary Compression Line.................. 442
Figure 8-43 Relationship between Modulus Number and Void Ratio for NC
Soils (after Janbu 1963) .................................................................. 445
Figure 8-44 Modulus Number for NC Clays (after Janbu 1985) ......................... 446
Figure 8-45 Modulus Number for NC Silts and Sands (after Janbu 1985) .......... 446
Figure 8-46 Variation of Empirical Coefficient f used for Calculating
Constrained Modulus with PI (after Stroud 1974) ........................... 447
Figure 8-47 Correlation between Normalized Constrained Modulus and Normalized
qt from CPTu for Clays (after Kulhawy and Mayne 1990) ................ 448
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UFC 3-220-10
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Figure 8-48 Correlation between Modified Secondary Compression Index and
Natural Water Content for Normally Consolidated Clays
(after Mesri 1973, Holtz and Kovacs 1981) ...................................... 449
Figure 8-49 Secondary Compression Index for Silts and Clays ........................... 450
Figure 8-50 Approximate Relationship between Coefficient of Consolidation
and Liquid Limit ............................................................................... 450
Figure 8-51 Correlation of Undrained Modulus Normalized by Undrained
Shear Strength to Overconsolidation Ratio
(after Duncan and Buchignani 1987) ................................................ 453
Figure 8-52 Correlation between PMT Modulus for Clays and SPT N
(after Ohya et al. 1982)..................................................................... 454
Figure 8-53 Undrained Modulus for Deep Foundations in Compression
(after Poulos and Davis 1980) .......................................................... 455
Figure 8-54 Undrained Modulus for (left) Drilled Shafts in Compression and
Uplift and (right) Spread Foundations in Uplift
(after Callanan and Kulhawy 1985) .................................................. 455
Figure 8-55 Correlations for Drained Young’s Modulus of Granular Soils ........... 457
Figure 8-56 Range of CBR based on DCP Predicted by Correlations ................ 460
Figure 8-57 Correlation for CBR in Terms of SPT N Value (Livneh 1989) ......... 461
Figure 8-58 Horizontal Hydraulic Conductivity based on D10
(after USACE 1993).......................................................................... 462
Figure 8-59 Hydraulic Conductivity based on D5 (after Kenney et al. 1984) ....... 463
Figure 8-60 Hydraulic Conductivity of Sands and Sand-Gravel Mixtures as a
Function of D5 , D10 , Cu , and e ........................................................ 464
Figure 8-61 Range of Shear Wave Velocities based on SPT N Value
xx
UFC 3-220-10
1 February 2022
LIST OF TABLES
Table 1-1 Principal Soil Deposits in Terms of Origin ............................................ 2

Table 1-2 Principal Soil Deposits by Mode of Occurrence .................................... 3
Table 1-3 Classification of Fine-grained Soils .................................................... 10
Table 1-4 Other Soil Classification Systems ....................................................... 16
Table 1-5 AASHTO Soil Classification System ................................................... 18
Table 1-6 Simplified Rock Classification - Common Igneous Rocks .................. 24
Table 1-7 Simplified Rock Classification - Common Metamorphic Rocks .......... 25
Table 1-8 Simplified Rock Classification – Common Sedimentary Rocks .......... 25
Table 1-9 Rock Color Descriptors (Geological Society of London 1977) ............ 26
Table 1-10 Grain-Size Descriptors for Rock ......................................................... 26
Table 1-11 Criteria for Defining Rock Grain Size (after FHWA 2017) ................... 27
Table 1-12 Weathering Classification ................................................................... 27
Table 1-13 Discontinuity Spacing ......................................................................... 28
Table 1-14 Hardness Classification of Intact Rock (Hough 1969) ........................ 28
Table 1-15 Criteria and Descriptions for Relative Rock Strength
(after FHWA 2017) ............................................................................. 29
Table 1-16 Engineering Classification for In Situ Rock Quality
(Merritt and Coon 1970) ..................................................................... 29
Table 1-17 Other Rock Classification Systems..................................................... 32
Table 1-18 Swelling Potential (Dakshanamurthy and Raman 1973) .................... 36
Table 1-19 Expansion Potential from Classification Test Data (Holtz et al. 2011) 36
Table 1-20 Classification of Potential Expansion of Soils using EI
(ASTM D4829) ................................................................................... 38
Table 1-21 U.S. Army Corps of Engineers Frost Design Soil Classification ......... 43
Table 1-22 Selection of Geophysical Method (after ASTM D6429) ...................... 46
Table 1-23 Corrosive Soil Environments (FHWA 2009) ....................................... 48
Table 2-1 Items that can be Evaluated During Field Reconnaissance
(NCHRP 2018 and FHWA 2002) ........................................................ 54
Table 2-2 Sources of Readily Available Subsurface Information
(after NCHRP 2018, FHWA 2002, and FHWA 2016) ......................... 57
Table 2-3 Historic Remote Sensing Data Sources ............................................. 58
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UFC 3-220-10
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Table 2-4 Current Remote Sensing Data Sources ............................................. 59
Table 2-5 Surface Geophysical Methods and Investigation Objectives
(after NCHRP 2018, Fenning and Hasan 1995, USACE 1995a,
Sirles 2006, FHWA 2006, and Anderson et al. 2008) ......................... 60
Table 2-6 Methods of Advancing an Exploration Hole in Soil
(NCHRP 2018 and Day 1999) ............................................................ 62
Table 2-7 Rock Core Drilling Methods (NCHRP 2018 and Day 1999)................ 63
Table 2-8 Soil and Rock Investigation Equipment and Their Applications
(NCHRP 2018 and Australian Drilling Industry Training Committee
2015) .................................................................................................. 64
Table 2-9 Selecting Number, Locations, and Depths of Investigation
(after NCHRP 2018, FHWA 2002, FHWA 2016, NYDOT 2013,
and SCDOT 2010) .............................................................................. 65
Table 2-10 Use and Limitations of Test Pits and Test Trenches
(after NCHRP 2018) ........................................................................... 68
Table 2-11 Samplers to Collect Disturbed Soil Samples ...................................... 69
Table 2-12 Samplers Used to Collect Intact Soil Samples ................................... 71
Table 2-13 Standard Size of Rock Casing, Drill Rods, Core Barrels, and
Coreholes (after ASTM D2113) .......................................................... 73
Table 2-14 Common Samplers for Rock Cores (after NCHRP 2018) ................... 74
Table 2-15 Common Underwater Samplers (after NCHRP 2018) ........................ 75
Table 2-16 Factors Affecting the Standard Penetration Test and SPT results
(after Kulhawy and Mayne 1990)........................................................ 81
Table 2-17 Types of Standpipe Piezometers ........................................................ 91
Table 2-18 In situ Testing Methods Used in Soil for Strength and Deformation
(after FHWA 2002) ............................................................................. 95
Table 2-19 Summary of In situ Test Procedures for Measuring Hydraulic
Conductivity of Soil Deposits ............................................................ 102
Table 2-20 Comparison of Non-nuclear Technologies for Assessing Soil
Density (Berney et al. 2013, 2016) ................................................... 108
Table 2-21 Gravimetric Testing Methods for Moisture Content
(after Berney et al. 2013) .................................................................. 108
Table 2-22 Indirect Testing Methods to Assess As-compacted Moisture
Content (after Berney et al. 2012) .................................................... 109
Table 2-23 Bias, Accuracy, and Precision of Test Methods for the As-Compacted
Measurement of Moisture Content (after Berney et al. 2012, 2013) . 110
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UFC 3-220-10
1 February 2022
Table 2-24 Interpretation of Lugeon Test Results (after Tunbridge 2017) .......... 114
Table 2-25 Types of Geotechnical Monitoring Instruments ................................ 115
Table 2-26 Methods of Determining Linear Deformation .................................... 117
Table 2-27 Angular Displacement Instruments ................................................... 121
Table 2-28 Piezometer Types............................................................................. 122
Table 2-29 Instruments for Measuring Load ....................................................... 125
Table 2-30 Example Questions for Instrumentation Decisions ........................... 128
Table 3-1 Amount of Soil Needed for Common ASTM Tests ........................... 134
Table 3-2 Summary of Phase Relationship Calculations .................................. 139
Table 3-3 Index Property Tests and Engineering Parameters Obtained .......... 141
Table 3-4 Laboratory Strength Tests with ASTM Standards ............................ 144
Table 3-5 Dynamic Tests for Soils .................................................................... 160
Table 3-6 Tests for Volume Change with ASTM Standards ............................. 164
Table 3-7 Potential Expansion for EI Values................................................... 169
Table 3-8 Laboratory Rock Strength Tests with ASTM Standards ................... 172
Table 4-1 Lateral Earth Pressure Coefficients .................................................. 183
Table 4-2 Equations for the Calculation of Change in Vertical Stress
Below Various Loading Conditions ................................................... 187
Table 4-3 Recommended Values for Trench Load Coefficient
(after Moser 1990) ............................................................................ 199
Table 4-4 Impact Factors for Live Loading of Buried Pipe
(from American Lifelines Alliance 2001) ........................................... 199
Table 4-5 Live Load Pressures from Various Vehicle Loading
(after American Lifelines Alliance 2001) ........................................... 200
Table 4-6 Approximate Overburden Rock Load Carried by Roof Support ........ 204
Table 4-7 Approximate Relationship Between Rock Quality Indices (after Deere
et al. 1970, Barton et al. 1974, Bieniawski 1990, Hemphill 2012)..... 205
Table 4-8 Types of Ground Behavior ................................................................ 206
Table 4-9 Ground Behavior for Clayey Fine-Grained Soils and Silty Sand
(after FHWA 2009) ........................................................................... 207
Table 4-10 Ground Behavior for Coarse-Grained Soils (after FHWA 2009) ....... 208
Table 4-11 Simplified Tunnel Support Loads based on Ground Behavior
(FHWA 2009) ................................................................................... 210
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UFC 3-220-10
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Table 4-12 Soft Ground Tunnel Support Loads for H > 1.5( B + H t ) .................... 211
Table 4-13 Summary of Model Parameters for Duncan-Chang Model ............... 215
Table 5-1 Settlement Calculation Methods for Different Soil Types
(after Coduto et al. 2011, Salgado 2008).......................................... 223
Table 5-2 Approximate Modulus Values for Coarse-Grained Soil
(after Bowles 1996, Duncan and Mokwa 2001) ................................ 234
Table 5-3 Estimates of Es based on SPT N 60 values....................................... 235
Table 5-4 Empirical Equations for Settlement of Coarse-Grained Soils ........... 237
Table 5-5 Correlations for Compression Indices............................................... 243
Table 5-6 Typical Values of Cα Cc (after Terzaghi et al. 1996)........................ 257
Table 5-7 Angular Distortion Limits for Various Structures
(after Skempton and MacDonald 1956, Polshin and Tokar 1957,
Duncan and Buchignani 1987, and Day 1990) ................................. 261
Table 5-8 Relative Mat Stiffness and Behavior
(after Brown 1969, Frazer and Wardle 1976) ................................... 263
Table 5-9 Methods to Reduce, Accelerate, or Prevent Excess Settlement
(after FHWA 2017) ........................................................................... 264
Table 5-10 Empirical Correlations to 1D Heave and Swell Pressure and Required
Input Parameters (after Rao et al. 2011, Vanapalli and Lu 2012)..... 278
Table 5-11 Foundation Design Approaches in Expansive Soil
(after Bowles 1996) .......................................................................... 280
Table 6-1 Common Finite Element Analysis Boundary Conditions................... 297
Table 6-2 Examples of Boundary Condition Usage .......................................... 298
Table 6-3 Typical Ranges of Horizontal Hydraulic Conductivity for Natural
Soil and Unfractured Rock Deposits (after USBR 2014) .................. 301
Table 6-4 Typical Range of Vertical Hydraulic Conductivity for Compacted
Soil in Embankments (after USBR 2014) ......................................... 302
Table 6-5 Estimating Hydraulic Conductivity based on Effective Grain Size .... 304
Table 6-6 Typical Values of Anisotropy in Natural Soils (after USBR 2014) ..... 306
Table 6-7 Typical Values of Anisotropy in Engineered Fill (after USBR 2014) . 306
Table 6-8 Seepage Severity Categories
(after Duncan et al. 2011, USACE 1956).......................................... 311
Table 6-9 Construction Methods for Vertical Seepage Barriers (Cutoff Walls) . 313
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UFC 3-220-10
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Table 6-10 Backfill Material Description and Characteristics for Vertical
Seepage Barriers (Cutoff Walls) ....................................................... 315
Table 6-11 Base Soil Categories for Mineral Filter Design (After FEMA 2011) .. 324
Table 6-12 Restraint Criteria for Mineral Filter Design (after FEMA 2011) ......... 324
Table 6-13 Flow Criteria for Mineral Filter Design (after FEMA 2011) ................ 325
Table 6-14 Segregation Criteria for Mineral Filter Design (after FEMA 2011) .... 325
Table 6-15 Geotextile Opening Size Criteria for Soils with Less than
10% Fines (after Luettich et al. 1992) ............................................... 329
Table 7-1 Strength Models for Different Soil Types and Drainage Conditions.. 353
Table 7-2 Factors of Safety for New Earth and Rockfill Dams (USACE 2003) . 368
Table 7-3 Factor of Safety for Dams using Spencer’s Method for Dams
(USBR 2011) .................................................................................... 368
Table 7-4 Recommendations for Reinforced Fill Soil in MSE Slopes
Based on Geometry.......................................................................... 371
Table 7-5 Summary of Applications and Materials for RSS .............................. 372
Table 7-6 Methods of Incorporating Geosynthetic Reinforcement Strength
in Factor of Safety Equation ............................................................. 376
Table 7-7 Steps for Designing an MSE Slope .................................................. 377
Table 7-8 Recommended Limits of Electrochemical Properties for
Reinforced Fill with Steel Reinforcement (after FHWA 2009b) ......... 390
Table 8-1 Typical Values of the Effective Stress Friction Angle for Coarse-
grained Soils (Carter and Bentley 2016)........................................... 397
Table 8-2 Typical Values of the Effective Stress Friction Angle for
Compacted Coarse-grained Soils (Carter and Bentley 2016)........... 397
Table 8-3 Relationship between SPT N Value, Relative Density and
Effective Stress Friction Angle (Meyerhof 1956) .............................. 398
Table 8-4 Relationship between SPT N Value, Relative Density, and
Angle of Internal Resistance (after Mitchell 1981) ............................ 399
Table 8-5 Relationship between Relative Density, Cone Tip Resistance, and
Effective Stress Friction Angle (after Bergdahl et al. 1993, and
Ameratunga et al. 2016) ................................................................... 402
Table 8-6 Coefficients for Stark and Hussain (2013) Residual Friction Angle
Correlation ........................................................................................ 419
Table 8-7 Typical Normally Consolidated Undrained Strength Ratios .............. 425
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Table 8-8 Approximate Relation of Undrained Strength Ratio and OCR
(after Schmertmann 1978)................................................................ 425
Table 8-9 Typical Values of m ......................................................................... 427
Table 8-10 Approximate Undrained Shear Strength for Cohesive Soils
Based on SPT N ............................................................................. 430
Table 8-11 Undrained Shear Strength Correlations to Dilatometer .................... 431
Table 8-12 Typical Values for Cc for Undisturbed Clays .................................... 432
Table 8-13 Compression Index Correlations. ..................................................... 433
Table 8-14 Recompression Index Correlations................................................... 436
Table 8-15 Modified Compression Indices for Saturated, Normally
Consolidated Sands (after Burmister 1962, Coduto et al. 2011) ...... 443
Table 8-16 Compressibility Data for Six Sands (Been et al. 1987) ..................... 443
Table 8-17 Typical Values of Cα Cc for Natural Soils
(after Mesri and Godlewski 1977) ..................................................... 449
Table 8-18 Relationships between Common Elastic Parameters ....................... 452
Table 8-19 Typical Range of Undrained Young’s Modulus for Clays.................. 453
Table 8-20 Typical Ranges of Drained Young’s Modulus for
Coarse-Grained Soils. ...................................................................... 456
Table 8-21 Correlations for Drained Young’s Modulus of
Coarse-Grained Soils using SPT N Values..................................... 456
Table 8-22 CBR Correlations with Grain Size, Atterberg Limits,
and Unit Weight. ............................................................................... 458
Table 8-23 CBR Correlations to Index and Compaction Properties
(after Singh et al. 2011) .................................................................... 458
Table 8-24 CBR Correlations with DCP .............................................................. 459
Table 8-25 Typical Ranges of Hydraulic Conductivity based on Soil Type
(after Terzaghi et al. 1996) ............................................................... 462
Table 8-26 Shear Wave Velocity Correlated to SPT N Value and Depth .......... 468
Table 8-27 Shear Wave Velocity Correlated to SPT N Value ........................... 468
Table 8-28 Correlations for Shear Wave Velocity with CPT results .................... 470
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IDENTIFICATION AND CLASSIFICATION OF SOIL AND ROCK
1-1 INTRODUCTION.
1-1.1 Scope.
This chapter presents criteria for soil and rock identification and classification based on
internationally accepted standards as well as information on their physical engineering
properties. Common soils and rocks are discussed as well as special materials, such
as expansive and collapsing soils, permafrost, limestone and related materials, coral
and coral formations, quick clays, and other materials (i.e. man-made fills, chemically
reactive and lateritic soils, calcareous sands, and submarine soils).
1-2 SOIL DEPOSITS.
1-2.1 Geologic Origin and Mode of Occurrence.
Soils are masses of solid particles along with the materials within the voids between the
particles. The solid particles typically are a mixture of sediments or other accumulated,
unconsolidated 1 material produced by the chemical and physical disintegration of rocks.
Soils can contain organic materials. From a geologic standpoint, soils can be classified
in terms of origin (e.g., transported, pyroclastic, residual, and organic), and mode of
occurrence (e.g., aeolian, alluvial, colluvial, glacial, and marine). A geologic description
can assist in correlating experiences between several sites, and in a general sense, can
indicate the pattern of strata to be expected prior to making a field investigation (test
borings, etc.).
Soils with similar origin and mode of occurrence are expected to have comparable, if
not similar, engineering properties. For quantitative foundation analysis, a geological
description is inadequate and a more specific classification and testing is required. A
study of references on local geology should precede a major subsurface exploration
program as this will help with planning the exploration and also with identifying possible
challenges for the project. Also, information on known projects near the site should be
obtained, if available, to give specific details about the soils and conditions that will likely
be encountered. Table 1-1 describes the principal soil deposits grouped in terms of
origin, and Table 1-2 describes the principal soil deposits by mode of occurrence.
1In this context, unconsolidated means that the particles have not lithified into rock. It does not imply a
particular state of consolidation as described in Chapter 5.
1
Table 1-1 Principal Soil Deposits in Terms of Origin
Major Division Principal Soil Deposits Pertinent Engineering Characteristics
Peats: Somewhat fibrous aggregate of decayed and decaying

Organic: Accumulation of highly vegetation matter having a dark color and odor of decay
organic material formed in place by Very compressible; entirely unsuitable for
the growth and subsequent decay supporting building foundations
of plant life Mucks: Peat deposits which have advanced in stage of decomposition
to such extent that the botanical character is no longer evident
Ejecta: Loose deposits of volcanic ash, lapilli, bombs, etc.

Typically, shard-like particles of silt size with larger
Pyroclastic: Material ejected from
volcanic debris; weathering and redeposition
volcanoes and transported by
produce high plasticity, compressible clay; unusual
gravity, wind and air
Pumice: Highly porous volcanic rock that is frequently associated with and difficult foundation conditions
lava flows and mud flows, or may be mixed with nonvolcanic sediments
Residual sands and fragments of gravel-sized material formed by

dissolution and leaching of cementing material, leaving behind the Generally favorable foundation conditions
more resistant particles, commonly quartz
Residual: Material formed by Residual clays formed by the decomposition of silicate rocks,
disintegration of underlying parent Variable properties requiring detailed investigation;
disintegration of shales, and solution of carbonates in limestone; with
rock or partially indurated material deposits present favorable foundation conditions
few exceptions, becomes more compact, rockier and less weathered
except in humid and tropical climates, where depth
with increasing depth; at intermediate stage may reflect composition,
and rate of weathering are very great
structure and stratification of parent rock
Transported soils: See Table 1-2
2
Table 1-2 Principal Soil Deposits by Mode of Occurrence

Relatively uniform deposits characterized by ability to
Loess: A calcareous unstratified deposit of silts or sandy or clayey silt traversed
Aeolian: Material stand in vertical cuts; collapsible structure; deep
by a network of tubes formed by root fibers now decayed
transported and weathering or saturation can modify characteristics
deposited by wind. Dune sands: Mounds, ridges, and hills of uniform fine sand characteristically Very uniform grain size; may exist in relatively loose
exhibiting rounded grains condition
Floodplain: Low-lying stream or river deposits that are subject to inundation by
floodwaters
Generally favorable foundation conditions; however,
Point bar: Alternating deposits of arcuate ridges and swales (lows) formed on
detailed investigations are necessary to locate
the inside or convex bank of mitigating river bends; ridge deposits consist
discontinuities; flow slides may be a problem along
primarily of silt and sand, swales are clay filled
riverbanks; soils are quite pervious
Channel fill: Deposits laid down in abandoned meander loops isolated when Fine-grained soils are usually compressible; portions
rivers shorten their courses; composed primarily of clay; however, silty and may be very heterogeneous; silty soils generally
sandy soils are found at the upstream and downstream ends present favorable foundation conditions
Backswamp: The prolonged accumulation of floodwater sediments in flood
Relatively uniform in a horizontal direction; clays are
basins bordering a river; materials are generally clays but tend to become siltier
Alluvial: Materials usually subjected to seasonal volume changes
near riverbank
transported and Terrace: Relatively narrow, flat-surfaced, river-flanking remnants of floodplain Usually drained and oxidized; generally favorable
deposited by running deposits formed by entrenchment of rivers and associated processes foundation conditions
water.
Estuarine: Mixed deposits of marine and alluvial origin laid down in widened
Generally fine grained and compressible; many local
channels at mouths of rivers and influenced by tide of body of water into which
variations in soil conditions
they are deposited
Lacustrine: Material deposited within lakes (other than those associated with
glaciation) by waves, currents, and organo-chemical processes; deposits consist Usually very uniform in horizontal direction; fine-
of unstratified organic clay or clay in central portions of the lake and typically grained soils generally compressible
grade to stratified silts and sands in peripheral zones
Deltaic: Deposits formed at the mouths of rivers, which result in extension of the Generally fine-grained and compressible; many local
shoreline variations in soil condition
Piedmont: Alluvial deposits at foot of hills or mountains; extensive plains or
Generally favorable foundation conditions
alluvial fans
3
Table 1-2 (cont.) Principal Soil Deposits by Mode of Occurrence
Talus: Deposits created by gradual accumulation of unsorted rock fragments Previous movement indicates possible future
and debris at base of cliffs difficulties; generally unstable foundation conditions
Colluvial: Material
transported and Hillwash: Fine colluvium consisting of clayey sand, sand silt, or clay
deposited by gravity
Landslide deposits: Considerable masses of soil or rock that have slipped down,
more or less as units, from their former position on steep slopes
Glacial till: An accumulation of debris, deposited beneath, at the side (lateral Consists of material from boulder and gravel to clay;
moraines), or at the lower limit of a glacier (terminal moraine); material lowered deposits are unstratified; present generally favorable
to ground surface in an irregular sheet by a melting glacier is known as a ground foundation conditions but rapid changes in conditions
moraine. are common.
Glacial: Material
transported and Glacio-fluvial deposits: Coarse and fine-grained material deposited by streams
deposited by glaciers, of meltwater from glaciers; material deposited on ground surface beyond Many local variations; generally present favorable
or by meltwater from terminal edge of a glacier is known as an outwash plain; gravel ridges known as foundation conditions
the glacier. kames and eskers; depressions known as kettles can be filled with peat
Glacio-lacustrine deposits: Material deposited within lakes by meltwater from
glaciers; consisting of clay in central portions of lake and alternate layers of silty Very uniform in the horizontal direction
clay or silt and clay (varved clay) in peripheral zones
Marine: Material Shore deposits: Deposits of sands and/or gravels formed by the transporting,
transported and Relatively uniform and of moderate to high density
destructive, and sorting action of waves on the shoreline
deposited by ocean
waves and currents in
shores and offshore Generally, very uniform, compressible and usually very
Marine clays: Organic and inorganic deposits of fine-grained material
areas. sensitive to remolding
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1-3 SOIL VISUAL DESCRIPTION, IDENTIFICATION, AND

CLASSIFICATION.
Standardized procedures for visual description, identification, and formal classification

of a soil specimen are presented in this section. These procedures follow the
corresponding ASTM standard available for this purpose. Visual description entails
describing the characteristics of the soil that can be perceived with the senses (e.g.
vision, touch, and smell). The identification of the soil refers to knowing the soil type
without having to use specialized equipment to do so. The visual description and
identification of soils are normally done in the field and the procedures are based on
ASTM D2488. The classification of the soils involves using specialized equipment and
tests to classify the soil using a standard classification system.
1-3.1 Definitions.
The definitions used in this chapter agree with the Unified Soil Classification system
presented in ASTM D2487.
Boulders: Rock particles will not pass a 12-inch square opening.
Clay: Soil particles passing a No. 200 (75-μm) sieve that exhibit plasticity (putty-like
properties) within a range of water contents, and considerable strength when air dried.
For classification of clayey soils, refer to Section 1-3.3.
Coarse-grained soils: Soils that contain 50% or more particles retained on a No. 200 (75
μm) sieve.
Cobbles: Rock particles that pass through a 12-inch square opening sieve but are
retained on a 3-inch square opening sieve.
Fine-grained soils: Soils that contain 50% or more particles passing a No. 200 (75 μm)
sieve.
Gravel: Soil particles that pass through a 3-inch square opening sieve but are retained
on a No. 4 (4.75 mm) sieve. Gravels can be divided into: (1) coarse gravels, gravel
particles that are retained on a ¾-inch square opening sieve, and (2) fine gravels, gravel
particles that pass through a ¾-inch square opening sieve.
Sand: Soil particles that pass through a No. 4 (4.75 mm) sieve and are retained on a
No. 200 (75 μm) sieve. Sands can be divided into: (1) coarse sands, sand particles that
are retained on a No. 10 (2.00 mm) sieve, (2) medium sands, sand particles that pass
through a No. 10 (2.00 mm) sieve and are retained on a No. 40 (425 μm) sieve, and (3)
fine sands, sand particles that pass through a No. 40 (425 μm) sieve.
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Silt: Nonplastic or very slightly plastic soil particles passing a No. 200 (75-μm) sieve
that exhibit little or no strength when air dried. For classification of silty soils, refer to
Section 1-3.3.
1-3.2 Visual Description and Identification (ASTM D2488).
Visual description of soil samples is commonly performed in the field during the drilling
process and consists of a visual description of the soil accompanied by an identification
of the type of soil. This should be done by an engineer or a qualified person and should
include as much information as possible regarding the observed conditions of the soil in
situ. The visual description of the soil, along with the drilling logs, can provide very
useful qualitative information to the engineer if done correctly. One of the most widely
used standards for this purpose is ASTM D2488, which uses visual examination and
simple manual tests to describe and identify soils.
1-3.2.1 Visual Description.
The descriptors for soils consist of properties and qualitative information of the soil that
can be perceived with our senses. This information can be very valuable to the
engineer. Below are some guidelines on what should be observed based on ASTM
D2488.
1-3.2.1.1 Descriptors for All Soils.
Color: Use the color or colors that best describes the sample. Color is an important
property that can help in identifying organic soils. Within a given locality, it may also be
useful in identifying materials of similar geologic origin. Layers or patches of different
colors should also be noted. The color described should be that of a moist sample. If
the color represents a dry condition, this should be stated in the report. A Munsell color
chart is a useful tool to help describing the color.
HCl reaction: Diluted hydrochloric acid (HCl) (one part of HCl to three parts of distilled
water) can be used to identify the presence of calcium carbonate. The HCl reaction
should be described as: (1) none, for no visible reaction, (2) weak, for some reaction
with bubbles forming slowly, or (3) strong, for violent reaction with bubbles forming
immediately.
Moisture condition: The moisture condition of the soil should be described as follows:
(1) dry, for soils with absence of moisture, dusty, or dry to the touch, (2) moist, for damp
soils with no visible water, or (3) wet, for soils with visible free water.
Odor: The odor of the soils should be described if the soil is organic or has an unusual
odor.
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Others: Additional comments like the presence of roots or root holes, difficulty in drilling
the hole, caving of the trench or hole, or the presence of mica should be included. In
addition, a local or commercial name, or a geologic interpretation of the soil could be
added to help identifying the soil.
1-3.2.1.2 Descriptors for Fine-Grained Soils.
Consistency: The consistency of intact fine-grained soils should be described as: (1)
very soft, if the thumb will penetrate soil more than 1 inch (25 mm); (2) soft, if the thumb
will penetrate soil about 1 inch (25 mm); (3) firm, if the thumb will indent soil about 1⁄4
inch (6 mm); (4) hard, if the thumb will not indent soil but readily indented with
thumbnail; or (5) very hard, if the thumbnail will not indent soil.
Structure: The structure for intact soils should be described using the following terms:
1) Stratified: Use for soils with layers of different material or color of at least ¼ inch
in thickness. The layer thickness should be noted.
2) Laminated: Use for soils with layers of different material or color of less than ¼
inch in thickness. The layer thickness should be noted.
3) Fissured: Use for soils that break along predetermined planes with little
resistance.
4) Slickensided: Apply to fissured soils that show polished, glossy, or sometimes
striated fracture planes.
5) Blocky: Describes soils that can be broken down into small angular lumps which
are hard to break down further.
6) Lensed: Use for soils with inclusions of small pockets of different soils scattered
through the mass of the clay. The lens thickness should be noted.
7) Homogenous: Use for soils with the same color and appearance throughout.
1-3.2.1.3 Descriptors for Coarse-Grained Soils.
Angularity: Describe the angularity of coarse-grained soils as: (1) angular, if the
particles have sharp edges and relatively plane sides with unpolished surfaces, (2)
subangular, if the particles are angular but with rounded edges, (3) subrounded, if
particles have nearly plane sides but well-rounded corners and edges, or (4) rounded, if
particles have smoothly curved sides and no edges. Figure 1-1 shows examples of
these four terms.
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Figure 1-1 Typical Angularity of Bulky Grains (after Sowers 1979)
Cementation: Describe the cementation of intact coarse-grained soils as: (1) weak, if
the soil crumbles or breaks with handling or little pressure, (2) moderate, if the soil
crumbles or breaks with considerable finger pressure, or (3) strong, if the soil will not
crumble or break with finger pressure.
Hardness: Describe the hardness of coarse-grained soils as hard if the particles do not
crack, fracture, or crumble when struck by a hammer, or state what happens to the
particles when hit by a hammer.
Maximum particle size: Describe the maximum particle size. For sands, describe it as
coarse, medium, or fine. For gravels, the maximum particle size is the smallest sieve
opening that the particle will pass. For cobbles and boulders, the maximum particle size
is the maximum dimension of the largest particle.
Range of particle size: Describe the range of particle sizes within each component. For
example, about 15% of coarse gravel and about 45% of fine to coarse sand.
Shape: Describe the shape as: (1) flat, for particles with width/thickness > 3, (2)
elongated, for particles with length/width > 3, or (3) flat and elongated, for particles that
meet both criteria.
1-3.2.2 Identification.
The identification method presented in this section follows ASTM D2488. The
identification should be performed on a sample that excludes cobbles and boulders.
These large particles should be manually removed from disturbed samples and ignored
for intact samples. The percentage of cobbles and boulders from the total samples
should be estimated by volume and noted. Estimate the percentage, by dry mass, of
gravel, sand and fines. The percentages should be estimated to the closest 5% and all
the percentages should add to 100%. If one type of soil is encountered but the amount
is less than 5% the term trace should be used to indicate its presence. A component
described as trace should not be included in the 100%.
1-3.2.2.1 Identification of Fine-Grained Soils.
The identification of fine-grained soils is based upon the results of the dry strength,
dilatancy, toughness, and plasticity tests.
Dry strength test: This test should be performed using a 0.5-inch diameter ball of soil.
The ball needs to be air-dried or dried by artificial means at a temperature not
exceeding 140°F. After drying, the ball is crushed between the fingers and the strength
is classified as:
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1 February 2022
1) None, if the dry specimen crumbles into powder with the mere pressure of
handling,
2) Low, if the dry specimen crumbles into powder with some finger pressure,
3) Medium, if the dry specimen breaks into pieces or crumbles with considerable
finger pressure,
4) High, if the dry specimen cannot be broken with finger pressure but will break
into pieces between thumb and a hard surface, or
5) Very high, if the dry specimen cannot be broken between the thumb and a hard
surface.
Dilatancy: This test is performed using a 0.5-inch diameter ball molded to a soft but not
sticky consistency. The ball is smoothed in the palm of one hand using the blade of a
knife or a small spatula. The hand is then shaken horizontally and vigorously struck
against the other hand several times. The reaction of water appearing on the surface
should be noted. The soil is then squeezed by closing the hand or pinched between the
fingers and the reaction of water is noted. The dilatancy is classified as:
1) None, if no visible change in the specimen was observed,

2) Slow, if water appears slowly on the surface of the specimen during shaking and
does not disappear or disappear slowly upon squeezing, or
3) Rapid, if water appears quickly on the surface of the specimen during shaking
and disappears quickly upon squeezing.
Toughness: This test is performed after the dilatancy test is completed and using the
same specimen. The test specimen is rolled by hand on a smooth surface into a thread
of about 1/8 inches in diameter. The sample is folded, mixed again, and rerolled until
the threads break at a diameter of about 1/8 inches, which means the soil is near the
plastic limit. The toughness of the soil is classified as:
1) Low, if only slight pressure is required to roll the thread near the plastic limit and
the thread and the lump are weak and soft,
2) Medium, if medium pressure is required to roll the thread to near the plastic limit
and the thread and the lump have medium stiffness, or
3) High, if considerable pressure is required to roll the thread to near the plastic limit
and the thread and the lump have very high stiffness.
Plasticity: The plasticity of the soil is classified based on observations made during the
toughness test as:
1) Nonplastic, if a 1/8-in-diameter thread cannot be rolled at any water content,

2) Low, if the thread can barely be rolled and the lump cannot be formed when drier
than the plastic limit,
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3) Medium, if the thread is easy to roll and not much time is required to reach the
plastic limit, if it cannot be rolled after reaching the plastic limit, and the lump
crumbles when drier than the plastic limit, or
4) High, if it takes considerable time rolling and kneading to reach the plastic limit, if
the thread can be rerolled several times after reaching the plastic limit, and the
lump can be formed without crumbling when drier than the plastic limit.
After these tests are performed, classify inorganic-fine-grained soils using the
information in Table 1-2. If the soil contains enough organic matter, identify the soil as
organic soil, OL/OH. Normally organic soils have a brown to black color and some
organic odor. Normally organic soils will not have a high toughness or plasticity and the
threads for the toughness test will be spongy.
Table 1-3 Classification of Fine-grained Soils
Soil Symbol Dry Strength Dilatancy Toughness and Plasticity
ML None to low Slow to rapid Low or thread cannot be formed
CL Medium to high None to slow Medium
MH Low to medium None to slow Low to medium
CH High to very high None High
For fine-grained soils with an estimated percentage of sand, gravel or both the term
“with sand” or “with gravel” depending on which one is more predominant should be
added to the group name. If the percentage of sand and gravel is the same, use the
term “with sand.” For fine-grained soils with an estimated percentage of sand, gravel or
both above 30%, the words “sandy” or “gravelly” should be added to the group name
depending on which one is more predominant. If the percentage of sand and gravel is
the same, use the word “sandy.”
1-3.2.2.2 Identification of Coarse-Grained Soils.
For coarse-grained soils, the identification is only based on visual observations.

Classify the soil as gravel or sand depending on which soil type is more predominant.
The soil is considered a clean gravel or a clean sand if the percentage of particles that
pass the #200 (75 μm) sieve is less than 5%. If the soil has a wide range of particle
sizes and considerable amount of the intermediate particle sizes, the soil is considered
to be a well-graded gravel or sand (GW or SW, respectively). If not, the soil is
considered a poorly-graded gravel or sand (GP or SP, respectively).
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For soils with 10% fines, a dual classification should be used. The first set of symbols
consist of the clean gravel or sand symbols (GW, GP, SW, or SP) followed by the gravel
or sand with fines symbols (GC, GM, SC, or SM). The group name should consist of
the name of the first set of symbols followed by the words “with clay” or “with silt” to
identify the fines.
If the soil has 15% or more fine-grained particles, the soil shall be identified as clayey
gravel (GC) or clayey sand (SC), if the fines are clay as determined in the previous
section, or silty gravel (GM) or silty sand (SM) if the fines are silty.
For gravels or sands with an estimated 15% or more of other coarse-grained particles,
the words “with gravel” or “with sand” should be added. If the sample contains cobbles,
boulders, or both the words “with cobbles,” “with boulders,” or “with cobbles and
boulders” should be added to the group name.
1-3.2.3 Examples.
Below are a few examples of visual descriptions and identifications (ASTM D2488):
1) Poorly-Graded Gravel with Sand (GW): About 80% medium to coarse, hard,
angular gravel; about 20% fine to coarse, hard, subangular sand; trace of fines;
maximum size, 70 mm, gray, moist; no reaction with HCl.
2) Silty Sand with Gravel (SM): About 65% predominantly medium to fine sand;
about 20% silty fines with low plasticity, low dry strength, low dilatancy, and low
toughness, about 15% fine, hard, rounded gravel, a few gravel-size particles
fractured with hammer blow; maximum size, 1.5 inch (38 mm); weak reaction
with HCl.
3) Organic Soil (OL/OH): About 100% fines with low plasticity, slow dilatancy, low
dry strength, and low toughness; wet, black, organic odor; strong reaction with
HCl.
4) Well-Graded Gravel with Clay, Sand, Cobbles and Boulders (GW-GC): About
70% medium to coarse, hard, rounded to subangular gravel; about 20% fine,
hard, rounded to subangular sand; about 10% clay low plasticity fines; moist,
dark grey; no reaction with HCl; original field sample had about 5% (by volume)
hard, rounded cobbles and a trace of hard, rounded boulders, with a maximum
dimension of 18 inches (450 mm).
1-3.3 Unified Soil Classification System (ASTM D2487).
The unified soil classification system (USCS) is the most common classification system
used for soils in the engineering community. This section is based on the USCS as
presented in ASTM D2487. This classification system consists of three major soil
divisions: coarse-grained soils, fine-grained soils, and highly organic soils, which are
further subdivided into 15 soil groups. To use this soil classification system the grain-
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size distribution (ASTM D6913) of the minus 3-inch (75-mm) material, and the liquid
limit and plasticity index (ASTM D4318) of the minus No. 40 (425-μm) sieve material
should be known. The various groups used in this classification system have been
divided to correlate in a general way with the engineering behavior of soils.
The grain-size distribution is needed for soils with 10% or more coarse-grained particles
and it can be estimated for soils with less than 10% coarse-grained particles. The liquid
limit and plasticity index are required for soils with 15% or more fines and the plasticity
can be estimated for soils with 5% to less than 15% fines as described in Section 1-
3.2.2.1. For soils with less than 5% fines, the plasticity is not needed.
1-3.3.1 Classification of Fine-Grained Soils.
Using the liquid limit and plasticity index, classify inorganic soils as lean clay (CL), fat
clay (CH), silt (ML), elastic silt (MH), or silty clay (CL-ML) using Figure 1-2. For dark
soils with organic odor, two liquid limit tests should be performed on the soil. One test is
performed before drying, and a second test is completed after oven drying the soil at
110 ± 5°C. The soil is considered an organic silt or clay if the liquid limit of the oven-
dried material is less than 75% of that of the material before oven drying. Classify the
organic soil as organic silt or clay OL or OH depending on where the liquid limit and
plasticity index of the non-oven-dried material plot in Figure 1-2.
Figure 1-2 Plasticity Chart
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If the fine-grained soil contains 30% or more retained in the No. 200 (75 μm) sieve the
words “gravelly” or “sandy” should be added to the group name based on the type of
particle that is predominant in the coarse-grained portion. For soils with equal
percentage of sand and gravel, use “sandy.” If the coarse-grained portion is less than
30% but greater or equal than 15%, the words “with gravel” or “with sand” should be
added to the group name depending whichever is predominant. For soils with equal
percentage of sand and gravel, use “with sand.”
Some properties of fine-grained soils are usually related to the plasticity characteristics
of the soil. Figure 1-3 describes how the liquid limit and plasticity index affect the
compressibility, permeability, toughness at the plastic limit, and the dry strength of fine-
grained soils.
Figure 1-3 Soil Property Variation with Liquid Limit and Plasticity
1-3.3.2 Classification of Coarse-Grained Soils.
Coarse-grained soils that contain more than 50% of the coarse-grained fraction retained
on the No. 4 (4.75-mm) sieve should be classified as gravel, and as sand otherwise.
Using the information on the grain-size distribution curve, calculate the following to
define whether the soil is well-graded or poorly-graded:
D60
Cu = (1-1)
D10
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and
D302
Cc = (1-2)
D60 × D10
where:
D60 , D30 , and D10 = particle-size diameters corresponding to 60%, 30%, and 10%,
respectively, passing on the cumulative particle-size distribution curve,
Cu = coefficient of uniformity, and
Cc = coefficient of curvature.
Coarse-grained soils are classified as well-graded if Cu is greater than or equal to 4.0
for gravels or greater than 6.0 for sands, and Cc is at least 1.0 but not more than 3.0.
Otherwise, the soil is poorly-graded.
Coarse-grained soils with less than 5% passing the No.200 (75-μm) sieve are
considered clean and are classified as well-graded gravel (GW), well-graded sand
(SW), poorly-graded gravel (GP), or poorly-graded sand (SP).
For coarse-grained soils with more than 12% fines, the classification of the fines needs
to be determined using the plasticity chart presented in Figure 1-2. If insufficient fines
are available to run plasticity tests, the classification of the fines shall be completed as
described in Section 1-3.2.2.1. Classify the soil as silty gravel or sand (GM or SM,
respectively) if the fines are silt or clayey gravel or sand (GC or SC, respectively) if the
fines classify as clay. If the fines plot as silty clay (CL-ML) classify the soil as a silty,
clayey gravel (GC-GM) or a silty, clayey sand (SC-SM).
Coarse-grained soils with a fine content between 5% and 12% require the use of a dual
classification. The first group symbol corresponds to that for a gravel or sand having
less than 5% fines (GW, GP, SW, or SP), and the second symbol correspond to a
gravel or sand having more than 12% fines (GC, GM, SC, or SM). The group name is
formed by the name of the first group symbol following the words “with clay” or “with silt”
depending on the characteristics of the fines. If the fines plot as a silty clay, CL-ML, the
second group symbol would be either GC or SC and the words “with silty clay” will be
used in the name.
If the soil is mainly sand or gravel but contains 15% or more of the other coarse-grained
constituent, the words “with gravel” or “with sand” shall be added to the group name.
Soils with cobbles and boulders should have the words “with cobbles,” “with boulders,”
or “with cobbles and boulders” added to the group name.
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1-3.3.3 Examples.
Below are a few examples of visual descriptions and identifications accompanied by the
proper USCS classification (ASTM D2487):
1) Well-Graded Gravel with Sand (GW): 71% fine to coarse, hard, angular gravel;
25% fine to coarse, hard, angular sand; 4% fines; Cc = 2.7, Cu = 12.4.
2) Silty Sand with Gravel (SM): 62% predominantly medium sand; 22% silty fines,
LL = 32, PI = 6; 16% fine, hard, rounded gravel; no reaction with HCl.
3) Poorly-Graded Gravel with Silt, Sand, Cobbles and Boulders (GP-GM): 75%
medium to coarse, hard, rounded to subangular gravel; 19% fine to medium,
hard, rounded to subangular sand; 6% silty (estimated) fines; moist, brown; no
reaction with HCl; original field sample had 7% hard, subrounded cobbles and
2% hard, subrounded boulders with a maximum dimension of 18 inches.
1-3.4 Soil Classification for Highways (AASHTO).
The American Association of State Highway and Transportation Officials (AASHTO)

developed their soil classification system, which is mainly used for highway design and
construction purposes. This system classifies the soils in 12 divisions based on the
grain-size distribution, the liquid limit, and the plasticity index, using only the soil
particles that pass through a 3-inch sieve. This section is based on the classification
system as detailed in ASTM D3282.
An important distinction between this classification system and the USCS is the
threshold used between the different types of soils. Coarse-grained or granular
materials are considered to be any soil that has 35% or less passing a No. 200 (75 μm)
sieve. Gravel is any material passing a 3-inch sieve and retained on a No. 10 (2.00
mm) sieve. Coarse sand is considered any soil that passes a No. 10 (2.00 mm) sieve
and is retained on a No. 40 (0.425 mm) sieve. Fine sand is any material passing a No.
40 (0.425 mm) sieve and retained on a No. 200 (75 μm) sieve. Silts and clays are
anything passing a No. 200 (75 μm) sieve, silts being materials with plasticity indices of
10 or less and clays being materials with plasticity indices above 10.
Soils are classified using Table 1-5 below from left to right. Highly organic soils (peat or
muck) may be classified in Group A-8. Classification of organic soils is based on visual
inspection and is not dependent on the percentage passing the 75-µm (No. 200) sieve,
liquid limit, or plasticity index. Organic material is composed primarily of partially
decayed organic matter, generally has a fibrous texture, a dark brown or black color,
and an odor of decay. These organic materials are unsuitable for use in embankments
and subgrades. They are highly compressible and have low strength.
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The classification obtained with the table above might be modified by adding a group-
index value that will be shown in parenthesis after the group symbol. The group index
is calculated using the empirical equation shown below:
GI =( F − 35)[0.2 + 0.005( LL − 40)] + [0.01( F − 15)( PI − 10)] (1-3)
where:
GI = group index,
F = percentage passing a No. 200 (75 μm) sieve (only considering the particles
passing a 3-inch sieve),
LL = liquid limit of the soil, and
PI = plasticity index of the soil.
The group index should be reported as zero if calculated to be negative, if the soil is
nonplastic, and when the liquid limit cannot be determined. For soils in the A-2-6 and A-
2-7 subgroups, the group index should be calculated using the second part of the
equation only (the part that contains the PI ).
1-3.5 Other Classification Systems.
Different regions in the United States and countries around the world have their own soil
classification systems. Below is a list containing the name of the country or region in
the United States and the reference to the standard used. This list is not intended to be
exhaustive but to show some examples. For the countries who are member of the
European Union, all the local standards are superseded by the ISO standards which
have the same numbers as European Norms (EN). Each country is allowed to further
refine the ISO/EN standards by adding appendices as long as the appendices do not
contradict the main standard. Each local standard will have the same number as the
ISO standard but will have the country designation at the beginning (e.g. BS for British
standards). When working on different projects in different parts of the United States or
the world the engineer should investigate the standards and norms that are used in that
particular area.
Table 1-4 Other Soil Classification Systems

Country / Region of USA / Agency Reference / Name
Australia AS 1726
Canada Canadian System of Soil Classification
International Organization for Standardization (ISO) ISO 14688
Occupational Safety and Health Administration 1926 Subpart P App A
(OSHA)
New Orleans USACE New Orleans District Internal Document
USDA USDA Soil Taxonomy
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1-3.6 Common Soil and Rock Names.
In practice, some areas have specific names for different types of soils. Some of these
names can be considered colloquialisms within the geotechnical engineering
community. Below is a list with definitions of the most common soil names used in
practice.
Adobe: Refers to sandy clays and silts of medium plasticity usually found in the semiarid
regions of the southwestern United States. These soils were commonly used to make
sun-dried bricks. The name is also applied to some high plasticity clays with high clay
content and high swell and shrink potential usually found in the western part of the
United States.
Baby poop: Refers to a very soft clay located just above limestone in karst. Frequently
orange and formed by dissolution.
Back-packing: Refers to any material (commonly granular) that is used to fill the empty
space between the lagging of a wall system and the rock surface.
Bank-run sand and gravel: Refers to the raw material excavated from a borrow pit, but
not sorted or separated into specific grades.
Beachrock: See reefrock.
17
Table 1-5 AASHTO Soil Classification System
Instructions: Work from left to right checking each column. The classification is the first one that matches all the criteria in the column.
Coarse-grained (granular) Materials Fine-grained (Silt-Clay) Materials Highly
General Classification
(35% or less passing No. 200 sieve) (more than 35% passing No. 200 sieve) Organic
A-1 A-3 A-2 A-4 A-5 A-6 A-7b A-8
Group Classification
A-1-a A-1-b A-2-4 A-2-5 A-2-6 A-2-7 A-7-5 A-7-6
Sieve analysis
Percent passing:
#10 (2 mm) ≤ 50
#40 (0.4 mm) ≤ 30 ≤ 50 ≥ 51
#200 (0.075 mm) ≤ 15 ≤ 25 ≤ 10 ≤ 35 ≤ 35 ≤ 35 ≤ 35 ≥ 36 ≥ 36 ≥ 36 ≥ 36 ≥ 36
Characteristics of
fraction passing #40
Liquid Limit ≤ 40 ≥ 41 ≤ 40 ≥ 41 ≤ 40 ≥ 41 ≤ 40 ≥ 41 ≥ 41
Plasticity Index ≤6 ≤6 NPa ≤ 10 ≤ 10 ≥ 11 ≥ 11 ≤ 10 ≤ 10 ≥ 11 ≥ 11 ≥ 11

Usual types of significant Stone
constituent materials Peat or
fragments; Fine sand Silty or clayey gravel and sand Silty soils Clayey soils
muck
gravel and sand
General rating as subgrade Excellent to good Fair to poor Unsuitable
Notes:
aNP indicates that the soil is "non-plastic." NP soils have
LL = PL , LL < PL , or PL that cannot be determined.
bUse the following criteria to divide A-7: A-7-5 has PI ≤ ( LL − 30 ) and A-7-6 has PI > ( LL − 30 ) .
Group Index, GI GI =( F −35) 0.2+0.005( LL −40 )  + 0.01 ( F −15) ( PI −10 )
Calculate the group index as:

where: F = %fines = P#200, LL = liquid limit, and PI = plasticity index.
If a negative value is calculated for GI , then report GI = 0 .
Note: Use only the second term in the GI equation for A-2-6 and A-2-7 soils.
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Bentonite: Refers to a high plasticity clay consisting of mostly montmorillonite, resulting
from the weathering of volcanic ash mainly in the presence of water. It is normally hard
when dry but swells considerably when wet. This clay is commonly used with water as
drilling mud and as liner in landfills.
Black cotton soil: Refers to a black expansive soil commonly encountered in India. The
name comes because this soil is common in areas where the main crop is cotton.
Blow sand: Term normally used for wind-driven or drifted sands.
Blue Marl: Name given to a bluish-green clay from the Miocene that can be found along
the fall line from Richmond into Maryland. This soil is considered to be acidic, usually
with a pH less than 4.0, which can affect water quality and prevent plant or aquatic life.
Bog: Refers to a wetland covered with peat with a high water table that accumulates
dead plants, such as sphagnum. It is generally nutrient poor and acidic.
Boney ground: Ground containing significant amounts of large gravel, cobbles and
boulders.
Boulder clay: Geological term used to designate clays formed from glacial drift that have
not been subjected to the sorting action of water and therefore contains particles from
boulders to clay sizes. Boulder clays are also called tills.
Breaker run: Crushed rock with large particles refers to large broken stone obtained as
part of quarrying or mining activities.
Buckshot: Term applied to clays of the southern and southwestern United States that
cracks into small, hard, relatively uniform-sized lumps on drying. The lumps are usually
the size of buckshot and the soil is very sticky when wet.
Bull’s liver: Name given to an inorganic silt or silty sand usually encountered in the New
York City area. The name Bull’s liver comes from its red color and jelly-like behavior
when it is subjected to vibration.
Bull’s Tallow or Bull Tallow clay: Refers to a tan or gray high plasticity clay typically
found in relatively thin layers directly above partially weathered rock or rock in the
Charlotte, NC area. This clay normally has high shrink and swelling potential.
Caliche: Refers to a sedimentary rock from arid and semiarid climate in which soil
particles, such as gravel, sand, clay, and silt, are cemented and coated by carbonate
(often calcium or magnesium carbonate). The level of cementation varies significantly
within a deposit. The soil has light coloration often exhibits light colored concretions of
various sizes depending on the level of development of the soil profile. The consistency
of caliche varies from soft rock to firm soil.
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Chip: Name given to crushed angular rock fragments smaller than a few centimeters.
Coffee grounds: Soil formed from freshwater marshes that has been dry for decades
and has decomposed to the point that is black and inert with no plasticity. It is black
and granular even when wet.
Colluvium: Loose soil deposited at the bottom of a slope.
Coquina: Soft, porous sedimentary rock, mainly limestone, composed largely of shells,
coral, and fossils cemented together with particles averaging 0.079 inch (2 mm) or
greater in size.
Desert varnish: Also called patina, rock varnish or rock rust. Consists of a thin, dark red
to black mineral coating found on pebbles and rocks surfaces in arid regions.
Diatomaceous earth: Soft, siliceous sedimentary rock that usually crumbles into powder.
When crumbled, the particles are silty and contain large amounts of diatoms, the
siliceous skeletons of minute marine or freshwater organisms.
Dispersive clays: These clays contain a high percentage of dissolved sodium in the pore
water. When these soils are exposed to water, the clay particles deflocculate (i.e.,
separate) making these soils very susceptible to erosion.
Fibric peat: Peat in which the original plant fibers are slightly decomposed and contain
67% or more of fibers.
Fill: Any man-made soil deposit. It can range from soils that are free of organic matter
and that are carefully compacted to heterogeneous accumulations of rubbish and
debris.
Fuller's earths: Soils having the ability to absorb fats or dyes. These soils have the
capability to decolorize oil or other liquids without chemical treatment. They are usually
high plasticity sedimentary clays.
Glacial till: See boulder clay.
Glassified sand: Term used to name the ground surface after a big forest fire.
Goonies: Name given to the cobbles found floating in a soil matrix.
Grove sand: See sugar sand.
Gumbo: Refers to a fine-grained, high plasticity clay of the Mississippi Valley according
to Sowers (1979). It has a sticky, greasy feel and forms large shrinkage cracks on
drying.
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Gyp or gip soil: Refers to gypsum soil (or soil containing gypsum) or caliche soil.
Hardpan: Normally refers to a soil layers that has become hard as rock due to
cementing minerals, does not become plastic when mixed with water and is relatively
impervious. The name has also been applied to any hard or overconsolidated layer that
is hard to excavate. Because of this ambiguity, Sower (1979) recommends that
engineers should avoid this term because many lawsuits have centered about the
meaning of it. The name implies a condition of the soil rather than a type of soil.
Humus: Refers to a brown or black material formed by the partial decomposition of

vegetable or animal matter. It is the organic portion of soil.
Kaolin: Refers to a white or pink clay of low plasticity. It is composed largely of minerals
of the kaolinite family.
Laterites: Refers to residual soils rich in iron formed in hot and humid climates (tropical
regions). The cementing action of iron oxides and hydrated aluminum oxides makes
dry laterites extremely hard. The high iron oxide content makes nearly all laterites to be
rusty-red. These are usually developed after significant weathering of the parent rock.
Ledge: Name used for bedrock in Vermont, and sometimes in New Hampshire.
Loam: Refers to a low plasticity sandy silt or silty sand mixed with organic matter that is
well suited to tilling. Mainly applies to the uppermost soil layer and should not be used
to describe deep deposits of parent materials. Major soil type in the USDA system.
Marl: Refers to a calcium carbonate or lime-rich sedimentary rock. It is mainly

composed of a mixture of sand, silt, or clay. Marls are often light to dark gray or
greenish in color and sometimes contain colloidal organic matter.
Montmorillonite: A group of very fine clay minerals with extreme swelling and shrinking
properties. Normally results from volcanic or hydrothermal activities. Bentonite is a
form of montmorillonite.
Muskeg: Refers to peat found in North America. According to Sowers (1979) the bogs
in which the peat forms are often termed muskegs.
Peat: Refers to a fibrous, partially decomposed and highly compressible organic soil.
Peats are dark brown or black.
Pit-run sand and gravel: See bank run.
Pluff mud: Refers to an odoriferous and very soft mud usually encountered in South
Carolina.
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Recycled concrete aggregate (RCA): Recycled road or structural concrete. The
concrete is usually processed and screened. The processing consists of crushing the
concrete into smaller pieces. Any leftover steel is removed using a magnet. This type
of material is very popular as a replacement for natural stone aggregates.
Recycled or reclaimed asphalt pavement (RAP): Term used to describe the removed
asphalt layer. When properly processed, it consists of high-quality and well-graded
aggregates coated by asphalt cement.
Recycled or reclaimed asphalt shingles (RAS): Recycled shingles that are used as
aggregate for hot mix asphalt. Depending on the quality, this can reduce the cost of the
new asphalt mix and the amount of fine aggregate used in the mix.
Recycled pavement material (RPM): Pulverized mixture of asphalt and base course
material usually forming a broadly-graded material.
Reefrock: Cemented coralline deposits.
Riprap: Boulder-size material normally placed to strengthen structures against scour,

wave action, and ice erosion.
Riverjack: Name usually given to alluvial cobbles and boulders.
Rock dirt combination (RDC): Local term used in the Harrisonburg, VA area to describe
material from a quarry consisting of a mixture of overburden soil and rock.
Rock flour: Fine-grained soil with silt-sized particles formed by the grinding of bedrock
by glaciers.
Shale: Refers to a fine-grained sedimentary rock made of silt and clay particles. This
rock usually breaks along thin laminates and can slake when subjected to wet-dry
cycles.
Shot rock: Refers to the material from a rock quarry that has not been sorted. It
includes everything that can be picked up (from fine sand to small boulders) after a
quarry blast. It is also a name given to riprap, although riprap is typically sorted and
graded.
Slickensided clay: Name given to a clay that has experienced repeated or enough
displacement along a fissure or a failure plane causing the surface to be smooth and
shiny.
Stone: Gravel size-particles manufactured by crushing rock.
Sugar sand: Local name for a type of fine sandy soil found in New Jersey. In Kansas,
the term refers to a type of granular calcite found in Ness and Hodgeman counties. In
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addition, the term may refer to a fine sand usually found in Florida that does not hold
water or nutrients very well. It is normally windblown medium and/or fine sand, poorly-
graded, nonplastic. It often contains nonplastic silt.
Till: See boulder clay.
Tire derived aggregate (TDA): Refers to a lightweight construction material obtained by

shredding or chipping scrap tires. The particle size usually ranges from 0.5 inches to 12
inches. TDA has been used in a wide range of projects, including lightweight
embankment fill, landslide repair or stabilization, landfills, retaining wall backfill, roads,
vibration mitigation, among others.
Topsoil: Upper and outermost layer of soil that support plant life. Usually contains
considerable organic matter.
Trap: Includes any dark-colored, fine-grained, non-granitic intrusive rock. The most
common trap rock is basalt, but also includes peridotite, diabase, and gabbro.
Tuff: Refers to a soft porous rock made from consolidated volcanic ash.
Varved clays: Sedimentary deposits consisting of alternate thin layers of silt and clay.
According to Sowers (1979), each pair of silt and clay layers is from 1/8 inch to 1/2 inch
thick. These soils result from deposition in lakes during periods of alternately high and
low water in the in flowing streams and are often formed in glacial lakes.
1-4 ROCK VISUAL DESCRIPTION, AND CLASSIFICATION.
1-4.1 Definitions.
Azimuth: Angle of a feature measured from North at 0° in a spherical coordinate

system.
Bedding: Planes of dissimilar materials caused by deposition normally encountered in

sedimentary rocks.
Dip: Angle that the surface of the rock forms with a horizontal plane.
Flow banding: Refers to the layering that is normally seen in rocks formed from magma.
Foliation: Refers to the laminated structure of the minerals in a rock created by the
deformation.
Igneous rocks: Rocks formed from the cooling and solidification of magma.
Lamination: Sequence of fine layers in a small scale (usually less than one centimeter in
thickness) normally observed in sedimentary rocks.
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Metamorphic rocks: Rocks formed from the transformation by heat, pressure, or both of
existing rocks. This transformation can alter the physical and chemical properties of the
rock.
Rock: Natural solid mineral or aggregate of minerals which is normally classified by the
way it was formed.
Rock mass: A large body containing rock in intact and weathered conditions
accompanied by structural discontinuities like fault, joints, etc., which can be
interbedded with soil material.
Sedimentary rocks: Rocks formed by the accumulation and cementation of smaller

particles.
Strike: Is the line representing the intersection of the rock surface with a horizontal
plane.
1-4.2 Visual Classification.
Rock samples and exposures can be visually classified by weathering, discontinuities,

color and grain size, hardness, and geological origin.
1-4.2.1 Geological Name and Origin.
The first step in visually classifying a rock is to identify the type of rock (e.g., igneous,
metamorphic, or sedimentary). Then, the geologic name and local name (if any) is
identified based on characteristics, such as texture and mineralogy. Igneous rocks are
normally classified by their mineralogical composition, texture, and color as can be seen
in Table 1-6.
Table 1-6 Simplified Rock Classification - Common Igneous Rocks

Color Light Intermediate Dark
Quartz and
Augite,
Feldspar Feldspar and Augite and
Principal Mineral Feldspar Hornblende,
Few other Hornblende Feldspar
Olivine
minerals
Coarse, Irregular Syenite Diorite Gabbro
Pegmatite
Crystalline Pegmatite Pegmatite Pegmatite
Coarse and Diorite Gabbro Peridotite
Granite Syenite
Medium Crystalline Dolerite Dolerite Dolerite
Fine Crystalline Aplite Diabase
Texture
Aphanitic Felsite Basalt

Glassy Volcanic Glass Obsidian
Porous (Gas
Pumice Scoria or vesicular basalt
Openings)
Fragmental Tuff (fine), Breccia (coarse), cinders (variable)
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Metamorphic rocks are normally classified by their texture and structure, as can be seen
in Table 1-7. Sedimentary rocks are normally classified by whether they are derived
from clastic sediments or chemical precipitates/organisms, as can be seen in Table 1-8.
Subordinate constituents in rock samples, such as seams or bands of other type of
minerals, should also be identified (e.g., dolomitic limestone, calcareous sandstone,
sandy limestone, mica schist).
Table 1-7 Simplified Rock Classification - Common Metamorphic Rocks

Structure
Texture
Foliated Massive
Coarse Crystalline Gneiss Metaquartzite
Sericite Marble
Mica Quartzite
Medium Crystalline Schist
Talc Serpentine
Chlorite Soapstone
Phyllite Hornfels
Fine to Microscopic
Slate Anthracite Coal
Table 1-8 Simplified Rock Classification – Common Sedimentary Rocks

Group Grain Size Composition Name
Rounded pebbles in medium-grained matrix Conglomerate
Mostly coarse grains
Angular coarse-grained fragments, often quite variable Breccia
Less than 10% of other minerals Siliceous sandstone
Appreciable quantity of clay
Argillaceous sandstone
minerals
More than 50% Medium coarse Appreciable quantity of calcite Calcareous sandstone
medium grains grains
Over 25% feldspar Arkose
Clastic
25%-50% feldspar and darker

Graywacke
minerals
Siltstone (if laminated,
Fine to very fine quartz grains with clay minerals
Shale)
<10% other minerals Shale
More than 50% fine
Appreciable calcite Calcareous Shale
grains Microscopic clay
Appreciable carbon /
minerals Carbonaceous Shale
carbonaceous material
Appreciable iron oxide cement Ferruginous Shale
Fossiliferous
Variable Calcite and fossils
Limestone
Organic
Medium to Dolomite Limestone or

Calcite and appreciable dolomite
microscopic Dolomite
Variable Carbonaceous material Bituminous coal
Calcite Limestone
Chemical
Dolomite Dolomite
Microscopic Quartz Chert, Flint, etc.
Iron compounds with Quartz Iron Formation
Halite Rock Salt
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Gypsum Rock Gypsum
1-4.2.2 Color and Grain Size.
Rock can be described with respect to basic colors on a rock color chart. The most
common chart used for this purpose in the United States is the Munsell rock color chart
which includes 115 color chips and works with both wet and dry specimens. Another
commonly used system is the one published by the Geological Society of London
(1977). This system is based on three descriptors as can be seen in Table 1-9.
Table 1-9 Rock Color Descriptors (Geological Society of London 1977)

1st Descriptor 2nd 3rd Descriptor
Descriptor
White
Yellow
Yellowish
Buff
Buff
Orange
Orangish
Brown
Brownish
Pink
Light Pinkish
Red
Dark Purplish
Blue
Orange
Green
Olive
Purple
Greenish
Olive
Greyish
Grey
Black
Grain size for rock refers to the sizes of the small grains that comprise the rock.
Because of the nature of some rocks, a 10X hand lens can be used, if necessary, to
examine rock sample. Various grain-size criteria have been established, and no single
criteria is standard or used most often. An example of grain-size descriptors for
different types of rock are found in Table 1-10. Another criterion presented by FHWA
(2017) is also included here in Table 1-11 which is similar to that presented by the
Geological Society of London (1977).
Table 1-10 Grain-Size Descriptors for Rock

Igneous and Metamorphic Sedimentary Rocks
Description Grain Diameter Description Grain Diameter
Coarse-grained > 5 mm Coarse-grained > 2 mm
Medium-grained 1 to 5 mm Medium-grained 0.06 to 2 mm
0.002 to 0.06
Fine-grained < 1 mm Fine-grained
mm
Aphantic Too small to be perceived by eye Very fine-grained < 0.002 mm
Glassy No grain form distinguishable
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Table 1-11 Criteria for Defining Rock Grain Size (after FHWA 2017)
Grain Size Description Criteria
< 0.003 in. Cannot be distinguished by unaided eye. Few to no mineral grains
Very Fine-Grained
(< 0.075 mm) are visible with a hand lens.
0.003 – 0.02 in. Few crystal boundaries are visible; grains can be distinguished with
(0.075 – 0.425 Fine-Grained difficulty by the unaided eye but can be somewhat distinguished by
mm) hand lens.
0.02 – 0.8 in. Most crystal boundaries are visible; grains distinguishable by eye
Medium-Grained
(0.425 – 2 mm) and with hand lens.
0.8 – 2 in. Crystal boundaries are visible; grains distinguishable with naked
Coarse-Grained
(2 – 4.75 mm) eye.
2 in. Crystal boundaries are clearly visible; grains are distinguishable
Very Coarse-Grained
(> 4.75 mm) with the naked eye.
1-4.2.3 Weathering.
Weathering is the mechanical or chemical deterioration of rock properties by the

exposure to water, temperature changes, among other factors. Rock can be described
as fresh, slightly weathered, etc. in accordance with Table 1-12 as indicated by the
International Society of Rock Mechanics (ISRM). As the degree of weathering
increases, usually the strength, stiffness, and quality of the rock decrease.
Table 1-12 Weathering Classification

Weathering
Grade Symbol Diagnostic Features
Grade1
No visible sign of decomposition or discoloration; rings under
Fresh F I
hammer impact
Slightly Slight discoloration inwards from open fractures, otherwise similar to
WS II
Weathered F
Discoloration throughout; weaker minerals such as feldspar
Moderately
WM III decomposed; strength somewhat less than fresh rock but cores
Weathered
cannot be broken by hand or scraped by knife; texture preserved
Most minerals somewhat decomposed; specimens can be broken
Highly
WH IV by hand with effort or shaved with knife; core stones present in rock
Weathered
mass; texture becoming indistinct but fabric preserved
Completely Minerals decomposed to soil but fabric and structure preserved
WC V
Weathered (saprolite); specimens easily crumbled or penetrated
Advanced state of decomposition resulting in plastic soils; rock
Residual Soil RS V
fabric and structure completely destroyed; large volume change.
1 After FHWA (2017).
1-4.2.4 Discontinuities.
The spacing of discontinuities in the rock can be described as close, wide, etc., in
accordance with Table 1-13. The structural features of a rock mass can be described
as thickly bedded or thinly bedded, in accordance with Table 1-13. Depending on
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project requirements, the form of joint should be identified as stepped, smooth,
undulating, planar, etc. In addition, the dip (in degrees), surface condition (rough,
smooth, slickensided), opening size (giving width), and filling (none, sand, clay, breccia,
etc.) should also be recorded.
Table 1-13 Discontinuity Spacing

Description for Joints, Faults,
Type of Feature Description Spacing
or Other Fractures
Very thickly (bedded, Very widely (fractured or
> 6 feet
foliated, or banded) jointed)
Thickly 2 to 6 feet Widely
Macrostructural: Bedding, 8 to 24
Medium Medium
Foliation, or Flow Banding inches
2.5 to 8
Thinly Closely
inches
0.75 to 2.5
Very Thinly Very Closely
inches
Intensely (laminated, 0.25 to 0.75
Microstructural: Lamination, Extremely close
foliated, or cleavage) inch
Foliation, or Cleavage
Very Intensely < 0.25 inch
1-4.2.5 Hardness.
The hardness of rock can be estimated by field tests using a geologic hammer or knife
and, in the laboratory, using the point load test in accordance with Table 1-14. The
corresponding range of strength values for intact rock is also provided. A more recent
grading system presented by the ISRM is presented in Table 1-15.
Table 1-14 Hardness Classification of Intact Rock (Hough 1969)

Approximate
compressive
Class Hardness Field Test
strength
(kg/cm2 or tsf)
Extremely Many blows with geologic hammer required to break intact
I >2000
hard specimen
Handheld specimen breaks with hammer end of pick under more
II Very hard 1000 - 2000
than one blow
Cannot be scraped or peeled with knife, hand held specimen
III Hard 500 - 1000
can be broken with a single moderate blow with pick
Can just be scraped or peeled with knife. Indentations of 1 mm to
IV Soft 250 - 500
3 mm show in specimen with moderate blow with pick
Material crumbles under moderate blow with sharp end of pick
V Very soft and can be peeled with a knife, but is too hard to hand trim for 10 – 250
triaxial test specimen
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Table 1-15 Criteria and Descriptions for Relative Rock Strength
(after FHWA 2017)
Approximate
Grade Description Field Identification Compressive Strength
(kg/cm2 or tsf)
Extremely Weak Specimen can be indented by thumbnail
R0 2.5 – 10.8
Rock
Specimen crumbles under sharp blow with point of
R1 Very Weak Rock geological hammer and can be peeled with a pocket 10.8 – 52.2
knife
Shallow cuts or scrapes can be made in a specimen
R2 Weak Rock with a pocket knife; a firm blow with a geological 52.2 – 252
hammer creates shallow indents
Specimen cannot be scraped or cut with a pocket
R3 Medium Strong Rock knife; specimen can be fractured with a single firm 252 – 522
blow with a geological hammer point
Specimen requires more than one firm blow of the
R4 Strong Rock 522 – 1,044
point of a geological hammer to fracture
Specimen requires many firm blows from the hammer
R5 Very Strong Rock 1,044 – 2,610
end of a geological hammer to fracture
Extremely Strong Specimen can only be chipped with firm blows from
R6 > 2,610
Rock the hammer end of a geological hammer
1-4.3 Classification by Field and Laboratory Measurements.
1-4.3.1 Rock Quality Designation.
The Rock Quality Designation ( RQD ) is only for NX size core samples and is computed
by summing the lengths of all pieces of core equal to or longer than 4 inches and
dividing by the total length of the coring run. The resultant is multiplied by 100 to get
RQD in percent. It is necessary to distinguish between natural fractures and those
caused by the drilling or recovery operations. The fresh, irregular breaks should be
ignored and the pieces counted as intact lengths. Depending on the engineering
requirements of the project, breaks induced along highly anisotropic planes, such as
foliation or bedding, may be counted as natural fractures. A qualitative relationship
between RQD , velocity index, and rock mass quality is presented in Table 1-16. The
velocity index is defined as the square of the ratio of the in situ to laboratory or intact
compressional wave velocities.
Table 1-16 Engineering Classification for In situ Rock Quality

(Merritt and Coon 1970)
RQD Velocity Index Rock Mass Quality
90-100 0.80-1.00 Excellent
75-90 0.60-0.80 Good
50-75 0.40-0.60 Fair
25-50 0.20-0.40 Poor
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0-25 0.00-0.20 Very Poor
1-4.3.2 Classification by Strength.
The uniaxial compressive strength and modulus ratio can be used to classify rock using
the results of ASTM D7012. The strength of intact sample can be used with Figure 1-4
to assign a classification as weak, strong, etc.
The point load strength can also be used to classify rock as indicated in Figure 1-4.
Point load strength tests, described in ASTM D5731, are sometimes performed in the
field for larger projects where rippability and rock strength are critical design factors.
This simple field test can be performed on core samples and irregular rock specimens.
The point load strength index, I s (50) , is defined as:
P
I s (50)= F ⋅ (1-4)
d2
De
F= (1-5)
50
where:
F = size correction factor,
P = the applied force at failure,
d = the distance between the loaded points, and
De = equivalent core diameter.
This index is related to the direct tensile strength of the rock by a proportionality
constant F depending on the size of sample. Useful relationships of point load tensile
strength index to other parameters such as specific gravity, seismic velocity, elastic
modulus, and compressive strength are readily available in the literature.
1-4.3.3 Classification by Durability.
Short-term weathering of rocks, particularly shales, and mudstones, can have a

considerable effect on their engineering performance. The weatherability of these
materials is extremely variable, and rocks that are likely to degrade on exposure should
be further characterized by use of tests for durability under standard drying and wetting
cycles. The slake durability test is a standardized procedure, described in ASTM
D4644, used for this purpose. For example, if wetting and drying cycles reduces the
grain size of shale, then rapid slaking and erosion in the field is probable when the rock
is exposed. Another method used for this purpose is the jar slake test described by
Santi (1998).
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1-4.4 Rock Mass Classification Systems.
Various classification systems have been developed for classifying rock mass for
engineering projects. Three of the main systems are described in this section. The
reader is encouraged to check for the classification system used in the region of
practice and that most applies to the project in question.
Figure 1-4 Rock Strength Characterization (after Broch and Franklin 1972)
The classification of the rock mass using some of these systems is useful to estimate
physical and engineering properties using published values and charts. Also, some
design methodologies rely on the classification of the rock mass.
1-4.4.1 Q System.
Barton et al. (1974) defines the value of Q in terms of RQD , the number of joint sets,
the joint properties, and a stress reduction factor. Extensive tables are provided to
guide the engineer in the selection of appropriate values. The roof pressure and
support requirements for tunnels can be estimated from the value of Q , as well as some
of the joint properties.
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1-4.4.2 Rock Mass Rating System.
The rock mass rating system ( RMR ), also known as the Geomechanics classification,
by Bieniawski (1973) classifies rock based on the uniaxial compressive strength, RQD ,
the spacing and properties of the joints, and groundwater conditions. While not solely
intended for tunneling applications, the RMR can be related to stand-up time,
unsupported active span length, and roof pressure.
1-4.4.3 Geological Strength Index.
The Geological Strength Index ( GSI ) has become a commonly used approach to
describe rock mass quality in a qualitative manner (Marinos et al. 2007). Because it is
closely linked to rock strength, GSI is most useful as a tool to help estimate rock
properties for stability analysis. The GSI can be quantified as indicated in Figure 1-5;
however, a range of values should always be used. Such values of GSI may be used
as input for empirical equations for the shear strength of a rock mass.
The GSI is most useful for rock masses with many discontinuities that cannot be
effectively modeled in direct fashion. According to Marinos et al. (2007), GSI should
not be used for (1) rocks with clear discontinuities and well-defined dominant structure,
(2) excavated faces in strong, hard rock with discontinuities spaced at similar
dimensions to the tunnel or slope, or (3) low strength “young” rocks such as marls,
claystones, siltstones, and weak sandstones. Marinos et al. (2007) developed a
modified GSI system for heterogeneous rocks, such as layered shales and sandstones,
as shown in Figure 1-6. The application of GSI to these rocks should account for their
tendency to behave differently at depth compared to near the ground surface.
1-4.4.4 Other Classification Systems.
Some other classification systems have been proposed depending on the region,
purpose, and needs. Some of these systems are summarized in Table 1-17.
Table 1-17 Other Rock Classification Systems

Rock Mass
Classification Main Uses Reference
System
Rock Structure Tunnel support and excavation and other ground support work in
Skinner (1988)
Rating mining and construction
Foundations, methods of excavation, slope stability, uses of earth
Unified Rock Williamson and
materials, blasting characteristics of earth materials, and
Classification Kuhn (1988)
transmission of groundwater
Shallow excavation, particularly with regard to hydraulic erodibility
Rock Material Field
in earth spillways, excavatability, construction quality of rock, fluid NRCS (2002)
Classification
transmission, and rock-mass stability
Conventional (cyclical, such as drill and-blast) and continuous
New Austrian
(tunnel-boring machine or TBM) tunneling; this is a tunneling Lauffer (1997)
Tunneling Method
procedure in which design is extended into the construction phase
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by continued monitoring of rock displacement; support
requirements are revised to achieve stability
bedded coal-measure rocks, in particular with regard to their
Coal Mine Roof Molinda and Mark
structural competence as influenced by discontinuities in the rock
Rating (1994)
mass
Figure 1-5 GSI Selection Chart for Jointed Rock (after Marinos et al. 2007)
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Figure 1-6 GSI Selection Chart for Heterogeneous Rock

(after Marinos et al. 2007)
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1-5 SPECIAL MATERIALS.
1-5.1 Expansive Soils.
1-5.1.1 Characteristics.
Expansive soils are distinguished by their potential for excessive volume increase upon
access to moisture. The swelling potential and the magnitude of the swelling pressure
are controlled by the clay minerals contained in the soil, the structure and fabric of the
soil, overburden pressure, and other physical-chemical aspects of the soil (Holtz et al.
2011). These soils usually contain montmorillonite and vermiculite clay minerals.
Expansive soils are characterized by a very high dry strength and high plasticity, are
often shiny when cut with a knife, and are very weak when wet (Holtz et al. 2011).
These soils usually form deep cracks during the dry season and expand closing the
gaps creating a homogenous appearance during the wet season.
Even though expansive soils can be encountered at great depth, they are mainly a
problem at shallow depths were the effect of variations in water content is greater
(FHWA 2017). The zone affected by seasonal variation in water content is also called
the active zone for expansive soils. This is very important when designing foundations,
roads, etc.
According to Holtz et al. (2011) expansive soils can be found around the world. In the
United States, the regions with the greatest occurrence of highly expansive soils are
North and South Dakota, Montana, eastern Wyoming, eastern Colorado, the Four
Corners Region, California, and east Texas. Figure 1-7 illustrates the distribution of
expansive soils throughout the United States.
1-5.1.2 Identification and Classification.
Expansive soils can be identified in various ways, and their swelling potential can be
nominally predicted. Expansive soils can sometimes be identified during visits to a
project site by looking for cracks in nearby structures or desiccation cracks in the soil
surface. Another method is identifying the clay minerals in the soil. If the soil has highly
expansive clay minerals (e.g. montmorillonite), that is a good indication that the soil
could be expansive. Some of the methods that can be used to identify clay minerals are
x-ray diffraction, differential thermal analyses, cation exchange capacity, and electron
microscopy.
Soil plasticity is often used to identify expansive soils. As the plasticity index or liquid
limit of the soils increases, the potential for swelling upon contact with water also tend to
increase. Dakshanamurthy and Raman (1973) presented the method shown in Table
1-18 to infer the swelling potential based on the liquid limit.
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The information presented in Table 1-19 and Figure 1-8 provides another method of
assessing the potential for volume change of a given soil.
Figure 1-7 Expansive Soils in the United States (Nelson and Miller 1992)
Table 1-18 Swelling Potential (Dakshanamurthy and Raman 1973)

Liquid Limit Classification
0 to 20 Non-Swelling
20 to 35 Low-Swelling
35 to 50 Medium-Swelling
50 to 70 High Swelling
70 to 90 Very High Swelling
> 90 Extra High Swelling
Table 1-19 Expansion Potential from Classification Test Data (Holtz et al. 2011)
Probable Expansion as a Percent of Total Colloidal
Degree of Plasticity Shrinkage
Volume Change (Dry to Saturated Content
Expansion Index Limit
Condition)1 (% < 1μm)
Very high >30 > 28 >35 <11
High 20-30 20-31 25-41 7-12
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Medium 10-20 13-23 15-28 10-16
Low <10 <15 <18 >15
1Under a surcharge of 1 psi.
Figure 1-8 Soil Expansion Prediction (after Holtz et al. 2011)
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The International Building Code 2018 (ICC 2018) considers a soil to be expansive if
these four criteria are met:
1) Plasticity index equal or greater than 15 as determined using ASTM D4318,

2) More than 10% of soil particles passing a No. 200 (75 μm) sieve, as determined
using ASTM D6913 or D1140,
3) More than 10% of soil particles are less than 5 μm in size, as determined using
ASTM D7928, and
4) Expansion index is greater than 20, as determined using ASTM D4829
(described below).
If the soil shows compliance with Item 4, it is not necessary to show compliance with
Items 1 through 3.
Two laboratory tests have standardized procedures to measure the swelling potential of
soils. In the Expansion Index test (ASTM D4829), the soil is compacted in a rigid mold
at a water content and unit weight that gives a degree of saturation of 50% ± 2%. A
vertical confining pressure of 1 psi is applied to the specimen before the specimen is
submerged in distilled water, and the deformation of the specimen is recorded for 24
hours or until the rate of deformation is below 0.0002 inch/hour, whichever occur first
with a minimum recording time of 3 hours. This test is used to obtain the expansion
index of the soil, defined as follows:
∆H
EI
= ×1000 (1-6)
Hi
where,
EI = expansion index,
∆H = change in height during the test, and
H i = initial height of the test specimen.
According to ASTM D4829, the expansion index can be used to estimate the swelling
potential of soils as described in Table 1-20.
Table 1-20 Classification of Potential Expansion of Soils using EI

(ASTM D4829)
Expansion Index, EI Potential Expansion
0-20 Very low
21-50 Low
51-90 Medium
91-130 High
>130 Very High
The one-dimensional swell or collapse test (ASTM D4546) can also be used to measure
expansion potential. This test method allows intact samples and samples compacted at
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different water contents and compactive effort to be tested. In addition, this test allows
different loading conditions, wetting and drying schedules, and reading intervals to be
used.
1-5.2 Collapsing Soils.
Collapsing soils are distinguished by their potential to undergo large decrease in volume
upon increase in moisture content without an increase in external loading. When dry,
these soils are stable and able to support significant structural loads. Examples of soils
exhibiting this behavior are loess, weakly cemented sands and silts where cementing
agent is soluble (e.g., soluble gypsum, halite, etc.), and certain granite residual soils. A
common feature of collapsible soils is loose bulky grains held together by capillary
stresses. Collapsible soils are also characterized by loss of strength when wetted, low
density, moisture sensitivity, and the presence of gypsum or anhydrite. Deposits of
collapsible soils are usually associated with regions of moisture deficiency (arid or semi-
arid regions). According to FHWA (2017), the following conditions are necessary for
collapse to occur:
1) an open, and partially saturated and unstable fabric,

2) enough total stress to make the soil structure metastable,
3) existence of a bonding agent or negative pore pressures to create a metastable
structure, and
4) addition of water to destroy the metastable structure.
The collapse of soils supporting structures can cause significant damage as a result of
total and differential settlement. The magnitude of the collapse depends on factors,
such as the soil composition, dry density, water content, confining stress, and the agent
causing the metastable structure. For this reason, it is important to identify collapsible
soils during the subsurface investigation so they can be remediated or considered in the
design phase.
One method to identify the potential of soils to collapse is presented by Holtz et al.
(2011) using data from the USBR and is shown in Figure 1-9. From this figure, the
potential for collapse increases with decreasing liquid limit and in situ dry density.
Ayadat and Hanna (2007b) presented a detailed study on the potential of collapse of
soil. In this study, they investigated the effect of the uniformity coefficient, water
content, and dry unit weight on the collapse potential. Figure 1-10 was presented as a
method to assess the potential for a soil to be collapsible along with a detailed method
to estimate the strain caused by collapsing for different soils. Ayadat and Hanna
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(2007a) also presented a method to assess the potential for a soil to be collapsible
using the fall cone apparatus (Section 3-2.4.2.6).
Figure 1-9 Collapsibility Based on In situ Dry Density and Liquid Limit
(after Holtz et al. 2011)
Figure 1-10 Design Charts for Predicting Collapse Behavior of Soils

(after Ayadat and Hanna 2007b)
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A method for quantifying the collapse potential of soils is presented in ASTM D4546.
This test method allows intact samples and samples compacted at different water
contents and compactive effort to be tested in a one-dimensional apparatus. In this
test, the specimen is loaded to a desired normal stress using any loading sequence,
water is added, and the vertical displacement is monitored.
1-5.3 Frost Susceptibility and Permafrost.
In non-frost susceptible soil, a typical volume increase due to ground freezing is about
4%. This volume increase is caused by the increase in water volume as it freezes. In
soils susceptible to frost, soil heave is much greater as water flows to colder zones
forming ice lenses. The formation of ice lenses typically is not uniform, meaning that
the increase in volume is not evenly distributed throughout a site and can cause distress
to structures. During warmer weather, the soil and ice lenses will tend to thaw from the
top down. Water can become trapped in the soil near the surface, leading to an
increase in water content and softening of the soil. The associated loss of support can
be even more detrimental to structures than the frost heave itself. This is specially a
problem for pavement structures and structures supported on shallow foundations, as
well as utilities, if not buried well below the depth of freezing.
Permafrost refers to a thick top later of soil that stays frozen throughout the year.
Permafrost particularly occurs in the Northern Hemisphere, including Canada and
Alaska. Construction in permafrost is very challenging and requires special
considerations during design and construction.
Problematic frost action requires both frost penetration into the ground and frost
susceptible soils. According to Holtz et al. (2011), if the frost penetration during the
worst part of the winter is less than about 0.30 m, frost heave should not be of concern
to structures. The maximum depths of frost penetration in the United States are shown
in Figure 1-11. These depths are for extremely cold winters without much snow cover.
Snow cover, especially early in the winter, will decrease the frost depth significantly.
Silts are the most susceptible to frost heave, but most soils with some fines content
have also some susceptibility to freezing. This includes soils classifying as SM, ML,
GM, SC, GC, and CL. Holtz et al. (2011) presented the information shown in Table 1-
21 summarizing a design classification system for frost susceptible soils. This system
uses the soil classification system and the percent of soil finer than 0.02 mm (8x10-4
inches) to assess the susceptibility to freezing.
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Figure 1-11 Maximum Depths (in meters) of Frost Penetration in the Continental
United States (NOAA 1978)
Figure 1-12 Rates of Heave in Laboratory Freezing Tests on Remolded Soils

(U.S. Department of the Army 1984)
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Table 1-21 U.S. Army Corps of Engineers Frost Design Soil Classification
Frost Percent Finer than Typical USCS
Frost Group Soil Type
Susceptibility 0.02 mm Classification
Gravels, crushed
Not frost-susceptible 0-1.5 GW, GP
Negligible to low stone and rock
(NFS)
Sands 0-3 SW, SP
Gravels, crushed
Possibly frost- 1.5-3 GW, GP
Possibly stone and rock
susceptible (PFS)
Sands 3-10 SW, SP
GW, GP, GW-GM,
S1 Very low to medium Gravelly soils 3-6
GP-GM
SW, SP, SW-SM,
S2 Very low to medium Sandy soils 3-6
SP-SM
GM, GW-GM, GP-
F1 Very low to high Gravelly soils 6-10
GM
GM, GM-GC, GW-
Gravelly soils 10-20
GM, GP-GM
F2 Medium to high
SM, SW-SM, SP-
Sands 6-15
SM
Gravelly soils >20 GM, GC
Medium to very high Sands except very
F3 >15 SM, SC
fine silty sands
Low Clays, PI >12 CL, CH
All silts ML, MH
Low to very high
Very fine silty sands >15 SM
Low to high Clays, PI <12 CL, CL-ML
F4 CL and ML; CL, ML,
Varved clays and
Very low to very and SM; CL, CH,
other fine-grained
high and ML; CL, CH, ML
banded sediments
and SM.
Figure 1-12 relates the rate of frost heave to the percent of particles finer than 0.02 mm
(8x10-4 inches) based on USCS classification. This figure also includes the
susceptibility classification for each type of soil. The information presented in this figure
is based on laboratory testing by the U.S. Department of the Army (1984). According to
Holtz et al. (2011), these rates are higher than those expected in the field, and soils with
a laboratory rate of frost heave of up to 1 mm/day (0.04 inches/day) can be used under
pavements, unless severe conditions are expected, but some frost heave should be
expected.
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1-5.4 Limestone and Related Materials.
Limestone, dolomite, gypsum, and anhydrite are characterized by their solubility and
thus the potential for the presence and/or development of cavities. Limestones are
defined as those rocks composed of more than 50% carbonate minerals of which 50%
or more consist of calcite and/or aragonite. Some near-shore carbonate sediments
(also called limestone, marl, and chalk) could fit this description. Such sediments are
noted for erratic degrees of induration, and thus variability in load supporting capacity
and uncertainty in their long-term performance under sustained loads. The most
significant limestone feature is its solubility. An extremely soluble limestone can contain
many solution caves, channels, or other open, water, or clay-filled features. These
features are often referred to as karst geology or topography.
Karst features that present important engineering challenges include vertical and
horizontal fissures and joints, pinnacles, and sinkholes. Fissures and joints may contain
very weak soil and also provide conduits for the flow of water, which are particularly
problematic for water retaining structures. Pinnacles are spires or spines of rock left
behind by the dissolution process and result in very uneven foundation support.
Sinkholes are the result of soil erosion into karst voids or the sudden collapse of voids.
Structures and pavements can be catastrophically damaged by sinkholes.
The identification of karstic areas should start by desk studies and site visits to look for
surface expressions of solution features. Sinkholes are the surface expression of rock
dissolution and can be used to infer that karstic rocks are found in the area. Aerial
photos, local geology maps, LIDAR data, etc. are also a useful source of information to
identify features that are caused by karstic rocks. A map of the karst and potential karst
areas in the United States presented by USGS (2014) is shown here in Figure 1-13.
A subsurface investigation program is very important in karstic areas to better

characterize the possible caverns, sinkholes, pinnacles, etc. Drilling is a very powerful
tool for this purpose and should be done more extensively in this type of terrain (FHWA
2017). Geophysical techniques, including shallow seismic refraction, resistivity, and
gravimetry, are often found to be valuable supplements. The suggested methods by
ASTM D6429 to estimate the depth to bedrock and the occurrence of sinkholes and
voids is presented in Table 1-22.
44
Figure 1-13 Karst and Potential Karst Areas in Soluble Rocks in the Contiguous United States (USGS 2014)
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Table 1-22 Selection of Geophysical Method (after ASTM D6429)

Seismic Electrical Electromagnetic
Application Refraction Ground
DC Freq. Time Very Low
or Method or MASW Penetrating Gravity
Resistivity Domain Domain Frequencies
Reflection Radar
Depth to
A A B B B B A B
rock
Voids and
B A B B B A A
sinkholes
Notes: “A” means preferred method and “B” alternate method based on the 2020 version of the standard.
1-5.5 Coral and Coral Formation.
Living coral and coralline debris are generally found in tropical regions where the water
temperature exceeds 20°C. Coral is a term commonly used for the group of animals
which secrete an outer skeleton composed of calcium carbonate, and which generally
grow in colonies. The term coral reef is often applied to large concentrations of such
colonies which form extensive submerged tracts around tropical coasts and islands. In
general, coralline soils deposited after the breakdown of the reef, typically by wave
action, are thin (a few meters thick) and form a veneer upon cemented materials
(limestones, sandstones, etc.).
Coralline deposits are generally poor foundation materials in their natural state because
of their variability and susceptibility to solution by percolating waters, and their generally
brittle nature. Coralline materials are often used for compacted fill for roads and light
structures. Under loads, compaction occurs as the brittle carbonate grains fracture and
consolidate. They can provide a firm support for mats or spread footings bearing light
loads, but it is necessary to thoroughly compact the material before using it as a
supporting surface. Heavy structures in coral areas are generally supported on pile
foundations because of the erratic induration. Predrilling frequently is required.
Because of extreme variability in engineering properties of natural coral formations, it is

not prudent to make preliminary engineering decisions on the basis of "typical
properties." Unconfined compression strengths of intact specimens may range from 50
tsf to 300 tsf.
Because the granular coralline and algal materials are derived from organisms which
vary in size from microscopic shells to large coral heads several meters in diameter, the
fragments are broadly graded and range in size from boulders to fine-grained muds.
Similarly, the shape of these materials varies from sharp, irregular fragments to well-
rounded particles. Coralline deposits are generally referred to as "biogenic materials"
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by geologists. When cemented, they may be termed "reefrock," or "beachrock," or other

names which imply an origin through cementation of particles into a hard, coherent
material.
1-5.6 Quick Clays.
Quick clays are clays from marine origin that are characterized by their very high
sensitivity or strength reduction upon disturbance. Quick clays are formed when the
formation water containing salts is replaced with fresh water. Disturbance of these
clays can be caused by construction activities or seismic ground shaking. In their
undisturbed state, they are relatively strong. Following disturbance, they become very
weak and possibly liquid. Because of their brittle nature, collapse occurs at relatively
small strains. This type of clay is normally found in Norway, Canada, Sweden, Finland,
Russia, the United States and other locations around the world. The Leda clay and
Champlain Sea clay in Canada are examples of quick materials.
Quick clays are readily recognized by measured sensitivities greater than about 15 and
by the distinctive, strain-softening shape of their stress-strain curves from strength or
compressibility tests. The sensitivity of clays is defined as the ratio of the undrained
shear strength in the undisturbed state to that in the disturbed state. The in situ liquidity
index of quick clays is typically above one, which means the water content is in excess
of the liquid limit.
1-5.7 Other Materials and Considerations.
1-5.7.1 Man-made Fills.
Man-made fills can be divided into engineered fills and uncontrolled fills. Engineered
fills are fills that were properly compacted to a specified density within a specified range
of water contents. These fills are normally strong, have low compressibility, and are
very favorable for building foundations. More detail on engineered fill can be found in
NAVFAC DM 7.2 (NAVFAC 1982).
Uncontrolled fills are very problematic because these fills were placed under conditions
that were not controlled and/or the materials that compose these fills were not
controlled. These fills can be made from uncompacted soils and may contain
deleterious building debris, old pavement, or concrete. Because of the variability of the
materials and uncontrolled placement conditions, the engineering properties are very
unpredictable and should be analyzed on a case-by-case basis.
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1-5.7.2 Chemically Reactive Soils.
For foundation construction, the main concerns related to chemically reactive soils are
usually are corrosion and gas generation. Corrosive soils can be problematic when
dealing with foundations, especially steel foundation systems. Potential for corrosion
must be considered in the design of foundations compared to the design life of the
structure. Protection systems can be used to reduce the corrosion rate. For concrete
foundations, increasing the cover thickness over the steel, the use of additive treated
concrete, or a specialized cement for this purpose can help mitigate the effect of
corrosion on the reinforcement.
Corrosion potential is determined in terms of pH, resistivity, stray current activity,

groundwater position, chemical analysis, etc. Based on this information, a compatible
foundation treatment, (e.g., sulfate resistant concrete, lacquers, creosote, cathodic
protection, etc.) can be prescribed. According to AASHTO (2017), a soil is considered
have high corrosion potential if: (1) the resistivity is less than 2,000 ohm-cm, (2) the pH
less than 5.5, (3) pH between 5.5 and 8.5 for soils with high organic content, (4) the
sulfate concentration is greater than 1,000 ppm, (5) is subjected to mine or industrial
drainage, or (6) the chloride concentration is greater than 500 ppm.
The location of the water table also influences the corrosion rate. Decker et al. (2008)
observed higher corrosion rate on steel piles in the section located above the water
table in the fluctuation zone.
FHWA (2009) presents an extensive study on the corrosion potential of soils focused on
MSE walls. Table 1-23 indicates regions in the United States with soils with high
potential for corrosion.
Table 1-23 Corrosive Soil Environments (FHWA 2009)

Environment Prevalence Characteristics
Pyritic, pH < 4.5, SO4 (1000 to 9000 ppm), Cl- (200 to
Acid-Sulfate Soils Appalachian Regions
600 ppm)
Sodic Soils Western States pH > 9, high in salts including SO4 and Cl-
FL, TX, NM and Western
Calcareous Soils High in carbonates, alkaline but pH
States
FL (Everglades), GA, NC, MI, Contain organic material in excess of 1% facilitating
Organic Soils
WI, MN microbial induced corrosion
Eastern, Southern and
Coastal Atmospheric salts and salt laden soils in marine
Western Seaboard States and
Environments environments
Utah
Road Deicing Salts Northern States Deicing liquid contain salts that can infiltrate into soils
Slag, cinders, fly ash, mine Either acidic or alkaline and may have high sulfate and
Industrial Fills
tailings chloride content
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For gas concentration, organic matter content and field testing for gas are usually
performed. If gas generation is expected, some form of venting system is designed.
The potential presence of noxious or explosive gases should be considered during the
construction excavations and tunneling.
1-5.7.3 Lateritic Soils.
Lateritic soils are found in tropical climates throughout the world. These soils are rich in
iron and aluminum. Because of the high iron content, most of these soils have a rusty-
red color. Extensive weathering of the parent rock normally develops these soils.
These soils can be problematic because of their loss in strength with time, high void
ratio and permeability, aggregate deterioration, and shrinkage cracks. These soils tend
to have shear strength characteristics between those of sands and silts. They are
prone to cause landslides, have highly variable moisture content, and provide erratic
conditions for foundations.
1-5.7.4 Calcareous Sands.
Calcareous sands are composed mainly of the skeletal remains of marine organism
having high carbonate content. These sands have significant intra-particle voids
created by shells that have not broken yet and by the cavities in the corals. These
sands are also characterized by having angular particles. The engineering properties of
these sands vary over a wide range and are controlled by the cementation and the
structure of the sand. More information on calcareous sands can be found in an ASTM
Special Technical Publication on the topic (ASTM International 1981).
1-5.7.5 Submarine Soils.
Ocean environments contains the following main topographic features: (1) the
continental margin including the continental shelf fringing the coast and the continental
slope; (2) the continental rise; and (3) the abyssal plains with local seamounts and
trenches (Randolph and Gourvenec 2011). The distribution of marine sediments along
those geomorphological regions varies with thickest sediment deposits being mostly
near continents and thinnest on recently formed mid-oceanic ridges. Continental
margins contain almost 75% of marine sediments, while only representing 20% of the
seabed area. The continental rise is also considered a depositional feature with
sediment thickness reaching locally up to 1.6 km (Randolph and Gourvenec 2011).
Marine sediments are either terrigenous (i.e. transported from land to the ocean), or
pelagic (i.e. settled through the water column). Coastal and nearshore zones are
dominated by terrigenous sediments. Terrigenous sediments are often granular silicate
minerals formed from erosion (lithogenous). Pelagic sediments are often finer and
derived from insoluble remains of marine organisms. Poulos (1988) presented samples
from abyssal plain and hill environments and found that most samples from abyssal
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plains and hills characterized as clayey silt to clay. Ocean sediment mapping also
revealed that deep ocean floors are widely covered with calcareous ooze that classified
as mostly silty clay (Poulos 1988).
Marine sediments that were deposited slowly and remained undisturbed from physical,
chemical, benthic biogenic, and/or anthropogenic processes are commonly normally
consolidated. Overconsolidated sediments can result from glaciation, recent sediment
erosion, or submarine landslides. Underconsolidated marine sediments can follow from
rapid sedimentation events and recent sediment dynamics, as well as from benthic
biogenic processes, amongst others. More information on the stress states of marine
sediments can be found in Randolph and Gourvenec (2011).
The following key differences between marine and terrestrial sediments can be listed:
• Environmental conditions cover a wider range of pressures (depending on water

depth) and temperature and can affect the engineering behavior of marine
sediments, particularly in deep ocean environments.
• Oceans feature saline water. Local salinity may affect the engineering behavior,
particularly of clays.
• Hydrodynamic conditions (i.e., waves, tides, currents) vary on spatial and
temporal scales, particularly in nearshore environments and on the continental
shelf. Hydrodynamic forcing can exert stresses onto the seabed and change
pore pressures. It also drives sediment dynamics, potentially leading to complex
sediment dynamics including erosion, transport, and deposition, and resulting in
geomorphodynamics including the formation, destruction, and migration of
bedforms on scales of centimeters to hundreds of meters. These processes
affect sediment composition, texture, and thus, engineering properties.
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1-6 SUGGESTED READING.
Topic Reference
Holtz, R. D., Kovacs, W. D., and Sheehan, T. C. (2011). An Introduction to
Geotechnical Engineering. Pearson.
Soil Classification FHWA. (2017). Geotechnical Engineering Circular No. 5 - Geotechnical Site
Characterization. U.S. Department of Transportation - Federal Highway
Administration, Washington, DC.
Hoek, E. (2007). Practical Rock Engineering.
Rock Description and
ASTM. (1988). Rock Classification Systems for Engineering Purposes -
Classification
STP984. ASTM International.
Holtz, R. D., Kovacs, W. D., and Sheehan, T. C. (2011). An Introduction to
Geotechnical Engineering. Pearson.
Expansive and Collapsing Soils FHWA. (2017). Geotechnical Engineering Circular No. 5 - Geotechnical Site
U.S. Department of the Army. 1984. Engineering and Design - Pavement
Frost Susceptibility Criteria for Seasonal Frost Conditions - Mobilization Construction - EM 1110-
3-138.
FHWA. (2017). Geotechnical Engineering Circular No. 5 - Geotechnical Site
Veress, M. (2020). “Karst Types and Their Karstification.” Journal of Earth
Limestone / Karst
Science, 31(3), 621–634.
Waltham, A. C., and Fookes, P. G. (2003). “Engineering Classification of Karst
Ground Conditions.” Quarterly Journal of Engineering Geology and
Hydrogeology, 36, 101–118.
Wesley, L. D. (2010). Geotechnical Engineering in Residual Soils. John Wiley
Lateritic Soils
& Sons, Ltd.
Randolph, M., and Gourvenec, S. (2011). Offshore Geotechnical Engineering.
Submarine Soils
CRC Press.
1-7 NOTATION.
Symbol Description
Cu Coefficient of uniformity
Cc Coefficient of curvature
d Distance between the loaded points in rock point load test
De Equivalent core diameter
D10 Particle size diameter corresponding to 10% passing
D30 Particle-size diameter corresponding to 30% passing
D60 Particle-size diameter corresponding to 60% passing
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Symbol Description
EI Expansion index
Percentage passing a No. 200 (75 μm) sieve (only considering the particles passing a 3-inch
F sieve)
F Size correction factor for rock point load test
GI Group index
GSI Geological strength index
Hi Initial height of the test specimen in the expansion index test
LL Liquid limit of the soil
P Applied force at failure
PI Plasticity index of the soil
PL Plastic limit of the soil
RMR Rock mass rating
RQD Rock quality designation
∆H Change in height during the expansion index test
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FIELD EXPLORATION, TESTING, AND INSTRUMENTATION
2-1 INTRODUCTION.
2-1.1 Scope.
This chapter contains information on exploration methods including use of geologic

maps, air photos, and remote sensing; geophysical methods; test borings and test pits,
and penetration resistance tests. Also presented is information on methods of drilling
and sampling, obtaining groundwater measurements, measuring in situ properties of
soil and rock, selecting field instrumentation and geotechnical performance monitoring
equipment.
2-1.2 Planning for Field Investigations.
The initial phase of field investigations should commence with a thoughtful assessment
of the data needs for the specific project, which will help define the objectives of the
subsequent field investigation. Prior to mobilizing to the field, readily available
information should be located that is relatively inexpensive and often invaluable. In
cases where the new project is adjacent to an existing project (e.g., highway widening,
lateral expansion of an existing levee, etc.), the initial research should focus on
information and data that have previously been collected and/or compiled for the
project. For a new project, the initial effort should include a detailed review of geological
conditions at the site and within the region where the site is located. This should then
be followed by a “desk top” study, utilizing sources of available data, including historical
and current aerial photography, remote sensing imagery, and (whenever possible) a
field reconnaissance. The collective information obtained from these activities should
be used as a guide in planning the project-specific field exploration.
To the extent possible as dictated by project data needs, individual test borings should
be supplemented by lower cost exploration techniques that include test pits, test probes,
and geophysical surveys. This is particularly true for remote sites, sites exhibiting wide
variability in subsurface conditions, projects occupying a large footprint, linear projects
(e.g., roadways, pipelines, etc.), and projects in the offshore environment where
mobilizations and test borings can be exceptionally expensive.
Project explorations generally have three distinct phases: (1) reconnaissance/feasibility

exploration; (2) preliminary exploration; and (3) detailed/final exploration. These phases
usually have different objectives. A fourth phase of exploration that involves additional
sampling and/or in situ testing may be desired and/or required during or after
construction to confirm conditions. Frequently (and most common for relatively small
projects), these three phases are combined into a single exploration effort.
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Reconnaissance includes a review of available topographic, geologic, and

hydrogeologic information; aerial photographs; data from previous investigations and
projects; and a site visit. Geophysical methods may prove to be helpful in many cases,
particularly for large projects where subsurface conditions are variable and for linear
projects (e.g., levees, highways, etc.). Reconnaissance/feasibility exploration frequently
reveals difficulties which may be expected in later exploration phases and assists in
determining the type, number and locations of borings required. Examples of
information that can be obtained from field reconnaissance activities are presented in
Table 2-1.
Table 2-1 Items that can be Evaluated During Field Reconnaissance

(NCHRP 2018 and FHWA 2002)
Item Things to Note Comments

Rank access using one of the following
criteria: (1) easy, (2) accessible by four-wheel Evaluating access helps determine the types
Access
drive, (3) dozer and grading required, and (4) of equipment that will be required.
inaccessible.
Existing overhead lines, marked gas lines,
Utilities information helps select appropriate
Utilities manholes, sewer outfalls, and power
in situ testing, drilling, and sampling locations.
substations.
Evaluating surface soils can reveal evidence

Presence of fill, debris, pollutants, slope
Surface soils of abandoned landfills, historic landslides,
instabilities, heave, subsidence, and scour
contamination, subsidence, and flooding.
Visual soil and rock classifications, loose

Shallow Subsurface materials can provide evidence
cobbles, boulders, rock outcrops, rock joint
subsurface for subsidence, landslide activity, unstable
patterns, faults, discontinuities, weathering,
materials soil and rock, and stratigraphy.
planes of weakness, talus, karst features
Surface drainage information provides

Surface indications of the depth to groundwater level,
Swamps, ponds, lakes, streams, and rivers
drainage hydraulic conductivity of the underlying
materials, and potential for flooding.
Subsurface drainage information provides

Subsurface Major aquifers, water wells, and pumping
indication of groundwater level, natural
drainage from deep wells
springs, and potential artesian conditions.
Evaluating terrain helps with selecting
Rank terrain in terms of (1) level ground, (2)
appropriate exploration and construction
Terrain sloping, (3) hummocky, (4) rolling hills, and
equipment, assessing the need for slope
(5) mountainous.
stability investigations, and site access.
Past investigations can provide information
Past Existing test pits, boreholes, coreholes, and
regarding stratigraphy, types of soil and rock,
investigations past blasting operations
and groundwater levels.
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Preliminary exploration may include borings and/or penetration soundings to identify

specific features (e.g., top of rock, etc.) and/or to recover samples. The collected
samples are generally used for index testing only. Penetration sounding test results are
often used to help identify the location of strata or formations where detailed exploration
activities will be advanced.
The detailed investigation phase typically includes subsurface borings, disturbed and
intact sampling for laboratory testing, standard penetration resistances, cone
penetration test soundings, and other in situ measurements. At critical sites it may also
include test pits, piezometer installation and measurements, pumping tests, etc.
Following completion of this phase and the associated testing, the site conditions and
soil/rock properties should be sufficiently known to design the project.
Monitoring of the site or structure is recommended throughout the construction and the
post-construction phases. Performance monitoring instrumentation (e.g., piezometers
and/or settlement plates to assess consolidation during staged loading) may need to be
installed. In some cases, further evaluation of foundation conditions may be required
during the construction phase. This is particularly true when foundation conditions have
the potential to vary widely across the project site (e.g., when using deep foundations
for project sites underlain by karst).
2-2 PUBLISHED REFERENCE MATERIALS.
When starting an investigation, the first step is to identify sources of readily available
and pertinent information. In general, this information comes from two sources: (1)
previous investigations; and (2) published literature in the public domain.
2-2.1 Previous Investigations.
For studies in developed areas, subsurface conditions and selected foundation

recommendations may be available from previous work for surrounding projects. Earlier
site-specific data may be “dated” and while the underlying geology is unchanged, the
site-specific information may have been superseded by recent activities. For example,
industrialized waterfront areas near major cities may undergo cycles of expansion and
reconstruction, causing subsurface conditions to change. Often old foundations and
wharf structures remain buried in place. Records of former construction may contain
information on borings, field tests, groundwater conditions, and potential or actual
sources of construction difficulties. Note that explorations from state departments of
transportation (DOT), the United States Geological Survey (USGS), and United States
Environmental Protection Agency (USEPA) may be publicly available.
Review of data from previous work should receive the greatest attention of any phase in
a reconnaissance investigation because it is likely very relevant. Additionally, this
information generally comes at relatively little cost and allows the project team to quickly
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become familiar with the project location and noted problems related to geology and
construction.
2-2.2 Published Geologic and Hydrogeologic Maps.
Data on the physical geology and topography of the United States (and foreign
countries) are available in maps and reports by government agencies, universities, and
professional societies. An example of documents and sources of available information
is provided in Table 2-2. While providing excellent regional and general information, the
information from these sources may not be entirely “site-specific.” However, this
information often can be used to identify specific data gaps that need to be addressed
during subsequent phases.
2-3 REMOTE SENSING DATA METHODS.
2-3.1 Sources.
Aerial photographs are a common type of remote sensing, including older printed
images (scales from 1:12,000 to 1:80,000) and reasonably high-resolution digital
images for most of the United States (scale of 1:1000 or better). Some regions possess
a wealth of “historic” imagery that may extend before the current site was developed.
Photos are useful for topographic and/or geologic mapping in addition to identifying
drainage patterns, locations of existing structures, vegetation, access routes and site
locations for planned explorations. Remote sensing also refers to non-photographic
data gathering satellites, from which data, such as vegetation development, water
sources, etc., are available. Table 2-3 summarizes sources and types of remote
sensing data that have been historically (i.e., pre-2019) used by geotechnical engineers.
The technologies identified in Table 2-3 generally require the purchase of images from
the entities that generated the images.
Table 2-4 provides a summary of more recent remote sensing technologies. Data from
some of these are free to the user and are often available on the internet. Data from
remote sensing technology can be incorporated into developing augmented reality (AR)
platforms, which provide an interactive experience where objects are projected into a
perceived real-world environment. This requires computer-generated information
presented in a geospatial environment through the use of special lenses and headsets.
2-3.2 Utilization.
Remote sensing represents a well-adopted resource by geotechnical engineers. The

emergence of internet-based mapping tools coupled with the cross-section profiling
capabilities using geographic information system (GIS) tools currently exceed the
capabilities and functionality of earlier tools.
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Table 2-2 Sources of Readily Available Subsurface Information

(after NCHRP 2018, FHWA 2002, and FHWA 2016)
Types of Sources of
Type of Available Information Comments
Documents Information
Maps can be used to evaluate
USGS and state
Topographic Site topography, physical features, access issues for field equipment
geological survey
maps and index map of site area and identify areas susceptible to
agencies
slope instability.
Available information is for
National Resource AASHTO and USCS classifications,
shallow depths (6 ft. or less) and
Conservation moisture contents, Atterberg limits,
is useful for identifying near-
Soil survey Service, Web Soil organic contents, chemical properties
surface problematic soils (e.g.,
reports Survey, and local (e.g., pH), permeability of soils,
soils susceptible to swelling and
soil conservation climate, stratigraphy, and groundwater
shrinkage) or identifying potential
agencies level
borrow sources.
Geologic Soil and rock formations (rock types,

maps and fracture, orientation and approximate These documents can be used to
USGS and state
reports, age), groundwater flow patterns, and identify areas susceptible to
geological survey
including bedrock contours that provide sinkholes, landslides,
agencies
sinkhole and approximate estimates of rock depths, subsidence, and other hazards.
karst maps and potential geologic hazards
Internet mapping
Man-made structures, geologic and Photographs can track site
sites, National
hydrogeologic information, current changes over time to identify
Aerial Agriculture
and past land use, borrow sources, potential problematic past land
photographs Imagery Program
and potential geologic and man-made use activities or geologic events,
(NAIP), and aerial
hazards including landslides.
survey companies
USGS, state Well maps and logs can be useful
Hydrological Hydrogeological features (e.g.,
natural resources to evaluate the need for
and well maps springs), groundwater hazards,
and soil survey construction dewatering and
and well logs stratigraphy, and groundwater depths
agencies permanent groundwater control.
Utility companies Very useful to identify locations

and local for in situ testing, drilling, and
Utility maps Locations of buried utilities
government sampling, useful to map
agencies equipment access routes
FEMA, USACE,
Flood This information can be used to
USGS, State and 100- and 500-year floodplains, data
insurance ensure that the site isn’t in the
local government for evaluating scour potential
maps 100- and 500-year floodplains.
agencies
Library of
Sanborn fire Congress, state
Environmental hazards and historical Sanborn maps are available for
insurance and university
land use urban areas.
maps libraries, and
Sanborn Company
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Types of Sources of
Type of Available Information Comments
Documents Information
Agencies: United States Geological Service (USGS), American Association of State Highway and Transportation
Officials (AASHTO), Unified Soil Classification System (USCS) , Federal Emergency Management Agency
(FEMA), United States Army Corps of Engineers (USACE)
Table 2-3 Historic Remote Sensing Data Sources
Type Description and General Use Availability
USGS, NCIC,
Available in 9-inch frames with overlap for stereoscopic viewing. Valuable because of
NCRS, USFS,
high resolution and available scales could range from 1:12,000 (or larger) to 1:80,000.
BLM, TVA
Satellite imagery with repetitive coverage every 18 days in four spectral bands:
Aerial photography
• BAND 4: emphasizes movement of sediment-laden water and delineates areas of

shallow water and useful in differentiating lithology
• BAND 6: emphasizes cultural features, such as metropolitan areas
• BAND 7: emphasizes vegetation, the boundary between land and water,
EROS
landforms and useful in structural interpretation of geology;
• BAND 8: provides the best penetration of atmospheric haze, the best band for
detecting faults, lineaments, mega-joint patterns or other structural features, and
also emphasizes vegetation, the boundary between land and water, and
landforms.
High-quality photography of Earth’s surface useful for regional planning, environmental
Skylab
studies, and geologic analyses. Images cover an area of 100 x 100 miles or 70 x 70
EROS
miles depending on the camera used. Images are from 1973-74 and do not provide
full coverage.
Black and white, color, or false-color infrared aerial photography produced from NASA
NASA
Earth Resources Aircraft Program with scales ranging from 1:120,000 to 1:60,000.
EROS
Coverage not available for all areas. Useful for planning, environmental and site-
oriented studies, and fault/lineament evaluation (color IR).
NCIC, Goodyear
Aerospace
Side-looking airborne radar (SLAR) is a valuable complement to photos for regional
SLAR
Corporation and
studies especially applicable in areas of persistent cloud cover. Scales range from
Motorola,
1:2,000,000 to 1:250,000. Best imagery for identifying regional faults/lineaments.
Westinghouse
Electric Corp.,
Obtained as
Thermal infrared (IR) imagery can be useful where temperature contrasts are
needed by aerial
Thermal IR
significant. Useful for special projects or as a complement to other remote sensing

survey firms.
data Useful in fault detection in covered alluvial areas, geothermal exploration, location
Images may be
of seepage, location of near surface peat deposits, covered meander scars, and heat
available from
loss studies.
an HCMM.
Agencies: United States Geological Service (USGS), National Information Center (NCIC), National Resources
Conservation Service (NCRS), U.S. Forest Service (USFS), U.S. Bureau of Land Management (BLM), Tennessee
Valley Authority (TVA), Earth Resources Observation System (EROS), Heat Capacity Mapping Mission (HCMM)
by National Space Science Data Center Goddard Space Flight Center
For project sites where limited information is available, aerial images greatly aid in
planning and layout of an appropriate boring program and currently be considered a
minimal requirement for projects. For large engineering studies, including highway and
airfield work, a three-dimensional (3D) visualization may be beneficial. Individual users
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can develop digital terrain model (DTM) files using data from UAVs data, and
commercial companies can economically develop local 3D topography with the use of
UAVs.
Table 2-4 Current Remote Sensing Data Sources

Type Description and General Use Availability
Aerial Photography
Recent and historical aerial maps (including approximate

topography) for most of the United States. Generally, very good Various internet map tools are
resolution at <1:1000 scale. Excellent to see regional and site- available, with some databases
specific topography, roads, drainage features. In many areas, it is updated quarterly. Most images are
possible to get a relatively recent “street view” 3D image to depict generally less than 3 years old.
observations from the ground surface.
Light Detection and Ranging (LIDAR) uses a pulsed laser light Usually provided by commercial
LIDAR
whose signal is reflected back to a sensor to record distance. The vendors as a specialized commodity
signal source is usually positioned on a moving vehicle and due to high equipment and
recovered data can be used to generate 3D images of terrain. processing costs.
Provided by commercial vendors

Synthetic Aperture Radar (SAR) is an advanced form of SLAR that with specialized electronic
uses radio waves from a moving platform. Data can be used for equipment for data capture and
SAR
high resolution 2D and 3D images, with the larger aperture (or processing, Images may be
larger antennae) providing higher resolution. available to the general public at
reasonable cost in the future.
Unmanned Aerial Vehicles (UAVs) or drones are increasingly

Equipment is readily available at low
useful for project aerial imagery. UAVs can carry digital and
cost for individual users.
UAV
infrared cameras and other sensors. High resolution is possible.

Commercial services are also widely
Overlapping passes allows for generation of 3D imagery and
available.
topography. Excellent resource for tracking construction progress.
Interpretation of information from aerial photographs and other remote sensed data
requires experience and skill. The interpretation process combined with other
information from the published reference material often informs the interpretation of
what features may be present at the project site. Spot checking in the field is an
essential element in the interpretation of geologic features from aerial photographs.
Aerial photographs are most helpful when assessing similarities and differences
between areas. Use of these images in urbanized and develop areas is of limited
quantitative subsurface informational value. As with any aerial image, whether
photographic or remote data, vegetation and cloud cover can often obscure the
underlying topography. Recently, computer enhancements of multi-spectral imagery
have made LANDSAT data compatible with conventional aerial photography.
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2-4 GEOPHYSICAL METHODS.
2-4.1 Utilization and Applications.
With increasing regularity, geophysical investigations are being used to estimate

subsurface conditions because of improved interpretation techniques and the overall
acceptance within the professional community of the geophysical characterization
techniques. Table 2-5 provides a summary of the common geophysical testing
techniques and the objectives/characterizations that are obtained from these
techniques. Information regarding the selection of appropriate surface geophysical
testing techniques is also presented in ASTM D6429.
Geophysical methods are best suited when investigating relatively large and/or linear
sites, including dams, reservoirs, tunnels, highways, and large groups of structures.
Techniques are available for both onshore and offshore exploration. Geophysics have
been used to locate gravel deposits and sources of other construction materials,
particularly for stratified materials where properties differ substantially from adjacent
soil/rock. As shown in Table 2-5, many of the geophysical testing methods are helpful
in identifying different subsurface strata and anomalies in the subsurface.
Table 2-5 Surface Geophysical Methods and Investigation Objectives

(after NCHRP 2018, Fenning and Hasan 1995, USACE 1995a, Sirles 2006, FHWA
2006, and Anderson et al. 2008)
Seismic Electrical and Electromagnetic Potential Field

Ground-Penetrating
Electromagnetic
Refraction and
Magnetometry
Surface Wave
Self-Potential
Microgravity
Resistivity
Reflection
Radar
Information Obtained
Lithology and stratigraphy     
Bedrock topography       
Water table   
Rippability of rock 
Shear wave velocity profile 
Fault detection     
Void and cavity detection      
Subsurface fluid flow  
Ferrous anomalies   
Conductive anomalies    
Corrosion potential 
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2-4.2 Advantages and Limitations.
In contrast to borings, geophysical surveys explore large areas rapidly and

economically. Because they evaluate conditions over a large area, the results reflect
average conditions in an area rather than a specific result that one would obtain from a
series of vertically advanced borings. Geophysical testing can prove most
advantageous in geologic conditions that display a strong contrast between adjacent
strata (i.e., rock beneath soil, interface between hard and soft rock, water- or air-filled
voids in soil or rock, etc.). Geophysical testing can often detect irregularities of bedrock
surface and the interface between soil and rock strata, and may be particularly useful in
karst topography.
Geophysical surveys can often distinguish boundaries between strata, but most
methods can only indicate approximate soil properties. These “approximate” properties
should be considered the average properties within the subsurface, as delineation of
specific properties of specific strata are generally not possible.
Interpretation of geophysical testing results is often difficult and subjective to the

experience of the operator or interpreter. In many cases, there are no definite criteria
for the interpretation of geophysical testing techniques. Some techniques are highly
specialized and almost all techniques require experienced operators and interpreters for
each application. Spot checks of “interpreted” versus “actual” conditions are strongly
recommended for each site using boring methods. Previously successful techniques
and an experienced interpreter should be used.
Differences in degree of saturation, presence of mineral salts in groundwater, or

similarities of strata that effect transmission of seismic waves may lead to vague or
inaccurate conclusions. These limitations notwithstanding, geophysical testing is
anticipated to see more widespread use and acceptance in the future. Further
reference and extensive discussion are found in FHWA (2003) and NCHRP (2018).
2-5 SOIL AND ROCK EXPLORATION METHODS.
Soil borings are the most commonly used method for subsurface soil exploration in the
field. They allow a vertical profile of soil to be established at a specific location and for
the collection of samples at selected vertical intervals at specific locations. Rock drilling
and coring techniques are more specialized than those used for soils and are used less
frequently.
2-5.1 Drilling and Boring Methods.
Most geotechnical borings in soils have historically utilized either hollow-stem augers or
rotary wash techniques, where numerous variations technologies are available. Recent
advancements that are gaining popularity and acceptance include the use of direct-push
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and sonic boring techniques. Table 2-6 identifies the applicability of the several
methods for advancing soil borings. Table 2-7 provides similar information for rock.
The drilling equipment used for geotechnical investigation is selected based on a
combination of: (1) ground conditions encountered at the site (i.e., soft ground, steep
terrain, over water, etc.) and (2) the type of drilling that is selected (i.e., auger, rotary,
percussion, etc.). Table 2-8 provides a summary of various types of drilling equipment
and their application. Figure 2-1 provides a schematic of the various drilling methods.
Table 2-6 Methods of Advancing an Exploration Hole in Soil

(NCHRP 2018 and Day 1999)
Method Procedure Applications Limitations / Remarks

Auger boring Dry hole drilled with hand or power Identify geologic units and Stratification destroyed;
(ASTM auger; samples recovered from water content above water sample mixed with water
D1452) auger flights table in soil and soft rock below the water table.
Sample limited by larger
Typically used in soils that
Hole advanced by hollow-stem gravel; maintaining
Hollow-stem would require casing to
auger; soil sampled with auger in hydrostatic balance in
auger boring maintain an open hole for
place hole below water table is
sampling.
difficult.
Coarse material tends to
Light chopping and strong jetting settle to bottom of hole;
Wash-type of soil; cuttings removed by Soft to stiff cohesive materials Should not be used in
boring circulating fluid and discharged and many granular soils. boreholes above water
into settling tub table where intact
samples are desired.
Becker Hole advanced using double Typically used in soils with Skin friction of casing
hammer acting diesel hammer to drive a gravel and cobbles; casing difficult to account for;
penetration 168 mm double-walled casing into driven open-ended if sampling repeatability of test
test (BPT) the ground. of materials is desired. unclear.
Not applicable in running
Most soils above water table;
sands; used for obtaining
Rotates and advances a 600- to can penetrate harder soils than
large volumes of
Bucket auger 1200-mm diameter drilling bucket above types; can penetrate
disturbed samples; used
boring with cutting teeth; bucket retrieved soils with cobbles and
to provide access to
and emptied on the ground. boulders if equipped with a
enter a boring for
rock bucket.
observations.
Static weight and percussion used
Most cohesive and granular Recovered samples are
Direct push to advance a 90- to 115-mm
soils; near-continuous sample generally disturbed
diameter casing;
High-frequency resonant vertical
Applicable in nearly all soils
oscillations advance a 75- to 300-
and much bedrock; returns Not cost effective in very
mm diameter core barrel; recovers
continuous stratigraphy; dense and hard rock
a continuous 3.3-m long core; after
Sonic drilling applicable for conditions both where coring is desired;
sample is retrieved, overcore
above and below the water recovered samples are
barrel advanced to bottom of core
table; process does not require disturbed
barrel by similar technique and
drilling fluids
process is repeated
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Table 2-7 Rock Core Drilling Methods (NCHRP 2018 and Day 1999)
Limitations /
Method Procedure Type of sample Applications
Remarks
Outer tube with diamond Rock cylinder 22 Obtain continuous core Core loss in fractured
Rotary (or tungsten carbide) bit to 100 mm wide in sound rock (percent or variable rock;
coring of rotated to cut annular and as long as 3 of core recovered blockage prevents
rock hole in rock; core m, depending on depends on fractures, drilling in badly
(ASTM protected by stationary rock soundness; rock variability, fractured rock; dip of
D2113) inner tube; cuttings standard size is equipment, and driller bedding and joint
flushed by drill fluid 54 mm diameter. skill) evident but not strike
Same as ASTM D2113,
but core and stationary Core loss in fractured
inner tube retrieved from Better core recovery in or variable rock;
Rotary Rock cylinder 28
outer core barrel by lifting fractured rock; much blockage prevents
coring of to 85 mm wide
device or “overshot” faster cycle of core drilling in badly
rock, wire and 1.5 to 3 m
suspended on thin cable recovery and efficiency fractured rock; dip of
line long
(wire line) through large- in deep holes bedding and joint
diameter drill rods and evident but not strike
outer core barrel
Rotary Soil cylinder 28.5 Soils and soft rocks Small sample;
Similar to rotary coring of
coring of to 53.2 mm wide that swell or equipment more
rock; swelling core
swelling and 600 to 1500 disintegrate rapidly in complex than other
retained by third inner
clay, soft mm long encased air (protected by plastic soil sampling
plastic liner
rock in plastic tube tube) techniques
Not cost effective in
hard rock where
High-frequency resonant Applicable most coring is desired;
vertical oscillations Continuous core bedrock; applicable for recovered rock cores
Sonic advance a 75- to 300-mm sample when conditions both above may be disturbed in
drilling diameter core barrel; overcore barrel is and below the water fractured rock,
recovers a continuous advanced table; process does not provides good
3.3-m long core require drilling fluids recovery and
continuous
stratigraphy
To locate rock, soft
Impact drill used; cuttings
Percussive Rock dust and seams, or cavities in Drill may become
removed by compressed
Method chips sound rock; advance plugged by wet soil
air
through boulders
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Table 2-8 Soil and Rock Investigation Equipment and Their Applications
(NCHRP 2018 and Australian Drilling Industry Training Committee 2015)
Rig Type Application

Truck-mounted drill rigs Areas with easy access
All-terrain vehicles drill rigs Sites with soft ground and rugged terrain
Track-mounted drill rigs Sites with swampy and very soft ground
Skid drill rigs Sites with steep terrain
Wireline drill rigs Rock sampling
Hydraulic direct-push rigs Fast, continuous sampling, cleaner (no spoils)
Sonic rigs Continuous sampling of soil and rock
Over water drilling for shallow water depths (10 ft. [3 m] or
Barges – regular
less)
Over water drilling for areas with deep water (up to 40 ft. [12
Jack up platforms
m])
Figure 2-1 Schematic of Various Drilling Techniques for Soil and Rock
(after NCHRP 2018 and Mayne 2012)
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2-5.1.1 Boring Layout and Depth.
General guidance for preliminary and final boring layout (i.e., location and number of
borings) and the depth of the borings is presented in Table 2-9 according to the type of
structure and/or problem being investigated. Additional discussion of the spacing and
number of borings is presented in FHWA (2002). In addition to structure type, boring
layout and depth are strongly dependent on past experience in the region (or at the site)
and the site/region geology. When a project is in an unfamiliar area, at least one boring
should extend well below the zone necessary for apparent stability to verify that the
site conditions are consistent with the anticipated geology and to assure no unusual
or unanticipated condition exist at depth.
The site geology is an important factor in developing the boring layout and should
influence the arrangement of borings so that geological sections may be viewed in the
context of the final design. This requires review of geologic maps of the area and
compilation of the information in a format that allows the geology, existing topography,
current site plans, and boring locations to be presented at similar scales on the same
figure/drawing.
In cases where detailed settlement, slope stability, or seepage analyses are required,
the boring plan should include a minimum of two borings in each critical stratum to
obtain intact samples (if applicable). For some site investigation programs this may
mean that preliminary sample borings and/or cone penetration soundings may be
needed to determine the most representative location and depth for intact sample
borings.
Table 2-9 Selecting Number, Locations, and Depths of Investigation

(after NCHRP 2018, FHWA 2002, FHWA 2016, NYDOT 2013, and SCDOT 2010)
Project Minimum Number of Investigation Locations Minimum Depth of Investigationa

• One location per pier if width of foundation is • For L ≤ 2 B , use depth of 2B
less than 100 ft. • For 2 B ≤ L ≤ 5 B , use depth of 3B
Bridge - • Two locations per pier if width of foundation is • For L ≥ 5 B , use depth of 4B
shallow greater than 100 ft. • Extend below any soft compressible
foundations • Additional investigation locations should be material into competent material
included if uncertain or highly variable • Extend 10 ft. into competent rock if
subsurface conditions are encountered. encountered before the above are met.
• One location per pier if width of foundation is • In soil, extend below the anticipated tip
less than 100 ft. or base elevation the greater of 20 ft. or
• Two locations per pier if width of foundation is 2x the maximum group dimension.
Bridge - deep greater than 100 ft. • In rock, extend below anticipated tip or
foundations • Additional investigation locations should be base elevation a minimum of 10 ft. or
included if uncertain or highly variable 3x shaft diameter for isolated
subsurface conditions are encountered piles/shafts or 2x maximum group
• At each shaft location for rock sockets dimension, whichever is greater.
a
B = footing width and L = footing length
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Table 2-9 (cont.) Selecting Number, Locations, and Depths of Investigation

(after NCHRP 2018, FHWA 2002, FHWA 2016, NYDOT 2013, and SCDOT 2010)
Project Minimum Number of Investigation Locations Minimum Depth of Investigationa
• A minimum of one location for each wall. If

the wall is greater than 100 ft. long, spacing
should be 100 to 200 ft. with locations • Extend below bottom of the wall 2x the
alternating in front to behind the wall. wall height or 10 ft. into hard rock.
Retaining
• Anchored walls: Additional locations in the • Should extend below any soft
structures
anchorage zone spaced at 100 to 200 ft. compressible material into competent
• Soil nail walls: Additional locations behind the material.
wall at a distance of 1 to 1.5x the wall height;
spacing should be at intervals of 100 to 200 ft
• Along embankment centerline: spacing of 200
• Depth of 2x the embankment height
ft. in uncertain or highly variable conditions to
unless a hard stratum is encountered
400 ft. in uniform conditions
Roadway - above this depth.
• At critical locations (maximum height or
embankment • If soft strata are encountered extending
maximum depth of soft strata): a minimum of
foundations to a depth greater than 2x embankment
three locations along the transverse direction
height, extend below the soft strata into
• Bridge approach embankment: a minimum of
competent material.
one location per abutment
• Minimum depth of 15 ft. (4.5 m) below
• Along centerline of cut: spacing of 200 ft. in lowest cut elevation unless a hard
uncertain or highly variable conditions to 400 stratum is encountered before the
ft. in uniform conditions minimum depth is achieved.
Roadway • At critical locations (maximum cut depth or • If soft strata are encountered, extend
cuts maximum depth of soft strata): a minimum of investigation to a competent layer.
three locations along the transverse direction • If base of cut extends below
• For cut slopes in rock: perform geologic groundwater level, extend depth of
mapping along the length of the cut slope. investigation to determine the depth of
the underlying pervious strata.
• Spacing of 100 to 300 ft. depending on the
subsurface conditions. Closer spacing for • Minimum depth of 10 ft. from the
Pavements
uncertain or highly variable conditions and proposed top of subgrade elevation.
longer spacing for uniform conditions.
• One boring at each end of the culvert.
• Additional borings between the end of culvert • Large culverts: same criteria as for
Culverts and spaced at 100 to 300 ft. depending on the bridge foundations
pipes variability of the subsurface conditions • Small culverts: Minimum of 10 ft. below
• For culvert extensions: one boring every 50 to anticipated invert elevation
100 ft. with a minimum of one boring
• 30 ft. below the anticipated top of
Poles, masts
• One boring at each foundation location foundation in soil or 10 ft. of rock coring
and towers
whichever is shallower.
a
B = footing width and L = footing length
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2-5.1.2 Abandoning or Sealing Boreholes.
Boreholes should be backfilled. Often, backfilling with the drill cuttings is sufficient.
However, boreholes must be sealed with grout in cases where the borings are
advanced below groundwater, in all cases where artesian pressures are encountered,
and whenever environmental borings are advanced. Under these conditions, boreholes
may be left temporarily unfilled to use for water-level observations after the initial field
investigation drilling is completed. In boreholes for groundwater observations, the
casings should be placed in tight contact with walls of boreholes or the annular space
between the standpipe and borehole should be backfilled using the appropriately
graded sand or gravel. Many agencies, such as the USACE and state DOTs have
specific guidelines for sealing boreholes, and these are part of the project specifications.
Additional discussion of details regarding groundwater investigation is presented in
Section 2-8.
2-5.2 Test Pits and Test Trenches.
Test pits are commonly used to examine and sample soils in situ at relatively shallow
depths. Test pits can be used to determine the depth to shallow groundwater, thickness
of topsoil or surficial deposits, and/or to assess near surface conditions. Test pits are
often used to determine sources of construction materials for earthwork projects, such
as dams and embankments. Test pits range from shallow, hand-excavated pits or
(more commonly) machine-advanced excavations.
Test trenches are essentially long test pits and are particularly useful for exploration in
very heterogeneous deposits (e.g., rubble fills) where borings may be misleading,
meaningless, or not feasible. Test trenches are used commonly for detection of fault
traces in seismicity investigations and for investigating conditions near a slide plane in a
landslide investigation. Safety precautions need to be recognized when working in and
around test pits and trenches.
Table 2-10 provides guidance for the use and limitations of test pits and trenches.
Hand-cut, block samples are frequently obtained from these explorations and may be
necessary for sensitive soils, brittle and weathered rock, and soil formations exhibiting a
honeycomb structure.
2-5.3 Other Exploratory Techniques.
Once a hole is advanced in either soil or rock, downhole tools can be placed in the open
hole to make specific measurements or serve as carriers for geophysical testing
instruments. Borehole cameras are commonly used for open holes in rock to assess
stratigraphy, as well as strike and dip of the formation. Geotechnical performance
monitoring instruments (i.e., slope inclinometers, water pressure transducers, borehole
extensometers, etc.) can also be placed in the advanced borehole.
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2-6 SAMPLING.
Recovery of representative samples of the subsurface soil and rock for testing is
perhaps the most common goal of the techniques in Section 2-5. These samples are
commonly referenced as disturbed or undisturbed depending on how well the recovered
sample maintains the structure of the in situ material. Disturbance is initiated by the
process of removing the soil/rock from the confined conditions in the subsurface. Thus,
an “undisturbed” sample is actually a misnomer, as it (hopefully) represents a minimally
(or nominally) disturbed sample. The term intact sample has largely replaced
undisturbed sample in geotechnical engineering vernacular.
Table 2-10 Use and Limitations of Test Pits and Test Trenches
(after NCHRP 2018)
Exploration
General Use Capabilities Limitations
Method
Provides data in
Expensive, time-
Hand-excavated inaccessible areas,
Bulk sampling, in situ testing, visual consuming, limited to
test pits and less mechanical
inspection depths above groundwater
shafts disturbance of
level.
surrounding ground.
Fast and economical, Equipment access,
Backhoe Bulk sampling, block sampling, in situ
generally less than 15- generally limited to depths
excavated test testing, visual inspection, depth of
feet deep, can be up to above groundwater level,
pits and trenches bedrock and groundwater.
30-feet deep limited intact sampling.
Fast, more economical Equipment access, difficult
Pre-excavation for piles and shafts,
than hand excavated, to obtain intact samples,
Drilled shafts landslide investigations, drainage
min. 30-inches dia., casing obscured visual
wells.
max. 6-feet dia. inspection.
Bedrock characteristics, depth of
bedrock and groundwater level,
Relatively low cost, Exploration limited to
rippability, used in conjunction with
Dozer cuts create exposures for depth above groundwater
backhoes for deeper excavations,
geologic mapping. level.
used to level areas for other
exploration equipment.
Costly, time-consuming,
requires shoring, only
Definitive location of
Evaluation of presence and activity of useful where dateable
Trenches for fault faulting, subsurface
faulting and sometimes landslide materials are present,
investigations observation up to 30
features. depth limited to zone
feet.
above groundwater level.
Specialized application.
Disturbed samples are primarily used for index tests that are performed for
classification. A disturbed sample needs only to be representative of the soil
composition and moisture because the soil structure is disturbed. Intact samples are
obtained primarily for laboratory strength, compressibility, and permeability tests. The in
situ structure and composition significantly influence the strength, compressibility and
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permeability (i.e., engineering) properties of the soil. Most of the discussion in this
section focuses on sampling from terrestrial or shallow-water locations. Offshore
samplers are specialized and are treated separately in Section 2-6.3.
The number and type of samples depend on the stratification of the subsurface, the type
of material encountered, the quantity needed for testing, and the criticality of the
application. For most projects, both disturbed and intact samples are obtained for
testing.
2-6.1 Soil Sampling.
2-6.1.1 Disturbed Soil Samples.
In general, representative disturbed samples are obtained at vertical intervals of no less

than 5 feet and at every change in strata. Continuous samples are occasionally
required or justified. This may be the case when a relatively thin layer of critical material
is anticipated. Table 2-11 lists common types of disturbed samples and samplers.
Recommended procedures for obtaining disturbed samples are provided in ASTM
D1586. The split barrel (a.k.a., split spoon) sampler, depicted in Figure 2-2, is the most
commonly used sampler.
Table 2-11 Samplers to Collect Disturbed Soil Samples

Sampler Cause of
Typical Soils that Give Best
(Method of Low Remarks
Dimensions Results
Penetration) Recovery
2.0-inch outside All soils finer than gravel-
Split Barrel Gravel- Standard Penetration Test (SPT)
diameter (OD), size particles; gravels
(140-lb sized performed using this hammer and
1.375-inch invalidate drive data; soil
hammer particles sampler and hammer; samples are
inside diameter retainer may be used in
driven) and larger extremely disturbed
(ID) coarse-grained soils
3- to 16-inch Most soils above water Method of determining soil profile,
Continuous
diameter; table; will not penetrate Hard soils, bag samples can be obtained; log
helical-flight
penetration to hard soils or those cobbles, and sample depths must account
auger
depths containing cobbles or boulders for lag time between penetration of
(Rotation)
exceeding 50 ft. boulders bit and arrival of sample at surface
Up to 48-inch Most soils above water
diameter table; can penetrate Several bucket types available,
Bucket Soil too
common; with harder soils than above including those with ripper teeth
auger hard to
extensions, types, can penetrate and chopping tools; progress is
(Rotation) penetrate
depth over 80 ft. cobbles and boulders with slow when extensions are used
are possible a rock bucket
Large 2- to 3-inch ID, Sample is intact but very disturbed;
Penetration 2.5- to 3.5-inch Particles A resistance can be recorded
Test (LPT) OD samplers, large than during penetration, but is not
Sandy to gravelly soils
(Up to 300-lb (e.g., Converse coarse equivalent to the SPT N value and
hammer and California gravel is more variable due to no standard
driven) samplers) equipment and methods
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Figure 2-2 Cross Section of Split Barrel Sampler
2-6.1.2 Intact Soil Samples.
Intact (or “undisturbed”) samples are most commonly obtained using a thin-walled steel
tube (Shelby tube) that is pushed at a relatively rapid and constant rate following
procedures in accordance with ASTM D1587. Intact sampling and samplers should
provide samples that comply with the following criteria: (1) show no visible distortion of
strata, (2) include no visible openings or softened material, (3) exhibit a recovery ratio
(i.e., sample length divided by distance of sample push) that exceeds 95 percent, (4)
have an area ratio (i.e., area displaced by the sampler tube divided by the area of the
sample) of less than 15 percent, and (5) have a clearance ratio (i.e., the difference
between the diameter of inside of the tube and the diameter of the opening at the
bottom of the tube divided by the diameter of the opening at the bottom of the tube) as
small as possible but less than 3 percent. A schematic and photograph of a thin-walled
Shelby tube that meets these criteria is presented in Figure 2-3.
Figure 2-3 Cross Section of Shelby Tube Sampler with Ball-check Valve Head
In general, intact samples of clean sands and gravels cannot be obtained, even when
using thin-walled samplers. For this reason, in situ testing methods are commonly used
in these soils, and intact sampling focuses on silts, clays, and coarse-grained soils with
a significant amount of silty and clayey fines. Because fine-grained soils can vary from
very soft to very hard, different types of samplers have been developed to facilitate the
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recovery of intact samples. Table 2-12 summarizes common types of samplers used
for intact soil samples.
Table 2-12 Samplers Used to Collect Intact Soil Samples
Sampler Typical Dimensions Method of Penetration
3.0-inch OD and 2.87-inch inside diameter Pressing with relatively rapid, smooth stroke;
Shelby tube
(ID) most common; available from 2- to 5- can be carefully hammer driven but this will
(ASTM D1587)
inch OD; 30-inch sampler length standard induce additional disturbance
Fixed or 3-inch OD most common; available from 2-

Stationary to 5-inch OD; 30-inch sampler length Pressing with continuous, steady stroke
piston standard
Continuous samples with 2-inch ID; Pushed into the ground with steady stroke;
Foil Sampler
up to 65 ft. long Pauses occur to add segments to sampler
Hydraulic piston 3-inch OD is most common; available from
Hydraulic or compressed air pressure
(Osterberg) 2- to 4-inch OD; 36-inch length standard
3.5- to 7-inch OD, producing samples 2.4 to
Denison Rotation and hydraulic pressure
6.3 inches; 24-inch sampler length standard
4-inch OD; uses 3-inch diameter Shelby
Pitcher sampler Same as Denison
tubes; sample length 24 inches
Soils that Give Cause of Disturbance or

Sampler Remarks
Best Results Low Recovery
Erratic sampling pressure,
Shelby Cohesive fine- Simplest device for undisturbed samples;
hammering, gravel particles,
tube grained or soft soils; clean boring before sampler is lowered;
crimping of tube edge,
(ASTM gravelly / very stiff little waste area in sampler; not suitable
improper soil types, pressing
D1587) soils will crimp tube for hard, dense or gravelly soils
more than 80% of tube length
Soft to medium
Erratic pressure during Piston at end of sampler prevents entry of
Fixed or clays and fine silts;
sampling, allowing piston rod fluid and contaminating material, requires
Stationary not for hard, dense,
to move during press, heavy drill rig with hydraulic drill head;
piston sandy, or gravelly
improper soil types for sampler less disturbance than Shelby tube
soil
Soft sensitive clays, Samplers should not be used Samples surrounded by thin strips of
Foil
silts, and varved in soils containing fragments or stainless steel, stored above cutter, to
sampler
clays shells prevent contact of soil with tube
Needs only standard drill rods; requires
Hydraulic adequate hydraulic or air capacity to
Silts and clays, Inadequate clamping of drill
piston activate sampler; samples generally less
some sandy soils rods, erratic pressure
(Osterberg) disturbed compared with Shelby tube; not
suitable for hard, dense, or gravelly soil
Stiff to hard clay, silt, Inner tube face projects beyond outer
and sands with Improper operation of sampler; tube, which rotates; amount of projection
Denison
some cementation, poor drilling procedures can be adjusted; generally good samples;
soft rock not suitable for loose sands and soft clays
Differs from Denison in that inner tube
Pitcher
Same as Denison Same as Denison projection is spring controlled; often
sampler
ineffective in cohesionless soils
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For soft soils, a stationary (or fixed-piston) sampler (Figure 2-4) or hydraulic piston
sampler is commonly deployed. For very soft soils and varved clays, a foil sampler may
be deployed, although limited in use in the United States.
Figure 2-4 Cross Section of a Stationary or Fixed Piston Sampler
For stiff fine-grained soils, or for layers of soft and hard materials, special samplers
have been developed that have the ability to “core” around the recovered stiff materials
while capturing the softer materials in the same thin-walled tube. The Denison sampler
and the Pitcher sampler are two types of common samples for these subsurface
conditions.
2-6.1.3 Intact Samples from Test Pits and Test Trenches.
One of the advantages of test pits and test trenches is that hand-trimmed (i.e., block)
samples may be obtained from the bottom or the sidewalls of the test pits and test
trenches. These block samples are potentially the least disturbed of all types of
samples. Unfortunately, the test pits and trenches are only feasible to a limited depth.
To obtain a block sample, a column of soil is trimmed the same size or slightly smaller
than the container that will be used for transporting the sample. The container should
be placed over the top of the sample and should provide as small an annular space as
possible. This annular space ideally would be filled using wax. A tight fit in a stiff
container that can be sealed provides the ideal conditions for retrieving and transporting
block samples with least disturbance.
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2-6.2 Rock Sampling.
Rock is sampled with core barrels that have either tungsten carbide or diamond core
bits at the cutting face. Drill rods and core barrels come in a variety of standard sizes
(see Table 2-13), depending on the size of the recovered rock core.
For hard and massive rock, relatively undisturbed rock samples may be obtained using
just the core barrel to recover the sample. More commonly, a tube (or series of tubes)
is used to contain the rock core and the tube is isolated from the core barrel to minimize
disturbance. Inner tubes must be used for undisturbed rock sampling whenever the
rock includes discontinuities. Table 2-14 lists summarizes techniques for recovery of
relatively undisturbed samples of rock. Double tube core barrels are the most
commonly used in practice. Depending on the number of discontinuities in the rock, the
recovered sample may be considered either disturbed or undisturbed. Schematics of
single and double tube rock core samplers are provided in Figure 2-5(a) and (b),
respectively.
Table 2-13 Standard Size of Rock Casing, Drill Rods, Core Barrels, and
Coreholes (after ASTM D2113)
Approx.
Core Approx.
Casing, Casing Bit Drill Rod Diameter
Casing OD Barrel Bit Diameter
Core Drill Rod OD OD of
(in.) OD of Core
Barrel (in.) (in.) Corehole
(in.) (in.)
(in.)
EX E 1-13/16 1-7/8 1-7/16 1-5/16 1-1/2 7/8
AX A 2-1/4 2-11/32 1-27/32 1-5/8 1-7/8 1-3/16
BX B 2-7/8 2-31/32 2-5/16 1-29/32 2-3/8 1-5/8
NX N 3-1/2 3-5/8 2-15/16 2-3/8 3 2-1/8
The suitability of rock cores for structural property tests depends on the quality of
individual recovered samples. If the properties of the intact rock are desired, then
smaller diameter cores are recommended because large-diameter rock cores likely
include more discontinuities than small-diameter rock cores.
The percentage of core recovery (i.e., recovered core length divided by the recorded
core run) provides an indication of soundness and degree of weathering of rock. The
Rock Quality Designation, RQD , (i.e., total length of recovered core pieces greater than
4 inches in length divided by the recorded core run) provide a better indication of the
soundness and degree of rock weathering because it essentially disallows the
consideration of the fractured and weathered rock intervals. The RQD is also a major
factor in assessing the behavior of the in situ rock mass, as defined by the Rock Mass
Rating ( RMR ) of the rock. The engineer and geologist should carefully examine rock
core samples exhibiting low recovery and/or low RQD to assess the reasons for low
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recovery and the interpreted poor rock quality. Details regarding on rock classification
and rock properties are presented in Chapter 1 and in NCHRP (2018).
Figure 2-5 Rock Core Samplers (after NCHRP 2018)

Table 2-14 Common Samplers for Rock Cores (after NCHRP 2018)
Diamond Core Barrels

Best Results in Soil or Rock
Dimensions Methods of Penetration
Types
Standard sizes: 1-1/2” to 3” OD, 7/8” Hard rock. All barrels can be
to 2-1/8” core. Barrel lengths 5 to fitted with insert bits for coring Rotary drilling using water or slurry
10 feet for exploration. soft rock or hard soil.
Details for tube sampling
Causes of Disturbance Best Results in Soil or Rock

Type Remarks
or Low Recovery Types
Drill fluid must circulate around core –
Single Fractured rock. Rock Primarily for strong, sound
rock must not be subject to erosion.
Tube too soft. and uniform rock.
Single tube not often used for exploration.
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Diamond Core Barrels

Best Results in Soil or Rock
Dimensions Methods of Penetration
Types
Improper rotation or feed Has inner barrel or swivel which does not
Double Non-uniform, fractured, friable
rate in fractured or soft rotate with outer tube. For soft, erodible
Tube and soft rock.
rock. rock. Best with bottom discharge bit.
Differs from Double Tube by having an
Triple
Same as Double Tube Same as Double Tube. inner split tube liner. Intensely fractured
Tube
rock core best preserved in this barrel.
Sampling of highly (or partially) weathered, fractured, or disintegrated rock is extremely

difficult. These materials often occur near the interface between soil and rock and
represent the transition between these two materials, especially in the case of residual
soils. The best samples of these materials are obtained by experienced drillers using
double- or triple-core barrel samplers.
2-6.3 Offshore Sampling.
In some cases, samples of soil and rock must be obtained from the bottom of rivers,
lakes, or the ocean. For water depths less than about 60 feet, the conventional soil and
rock boring equipment can be used on small jack-up platforms, small barges, or barrel
floats. The challenge is that floating equipment requires suitable anchoring and is
limited to fairly calm water, although tidal fluctuations can be easily accommodated. For
deep water sites and/or extreme ocean settings, large dedicated drill ships, specialized
equipment, and experience are required to obtain quality intact samples. Table 2-15
identifies some of the specialized equipment used for underwater sampling.
Table 2-15 Common Underwater Samplers (after NCHRP 2018)
Size of Length of Water Depth Method of

Sampler Remarks
Sample Sample Limitations Penetration
Reliable grab sampler;
To 200 ft. and
Peterson intact samples may be
Grab ± 6-inch depth more with Clam shell jaw
Dredge obtained with jaws that
additional weight
precisely mate
No limit on depth
Open
2.5- to Core barrels but required
Barrel Spooled freely off
6-inch length from 6 weight, amount of
Gravity the winch drum
diameter to 30 ft line or size of
Corer
vessel may control
Core barrels Relatively light weight core
About Free fall from 10
Pflueger available in 12, for upper 1 to 3 ft. of bottom
1.5-inch From 25 to 200 ft to 20 ft. above
Corer 24 and 36 in. sediments; usually not
diameter bottom
length suitable for strength tests
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Free fall from

No depth limit
Standard calibrated height Capable of obtaining
Standard except that
Piston barrel is 10 ft. above bottom samples suitable for
corer available weight,
Gravity Additional 10 such that piston strength tests with
has 2.5- amount of line, or
Corer ft. sections can does not experienced crew; samples
in. barrel size of vessel may
be added penetrate may be seriously disturbed
control
sediments
Samples are disturbed
because of vibration and
Minimum depth
20 ft. standard, large area ratio; not suitable
Sample limited by draft of Pneumatic
Vibratory can be for strength testing;
is 3.5-in. support vessel; impacting
Corer lengthened to Penetration resistance can
diameter maximum depth vibratory hammer
40 ft. be measured; obtains
about 200 ft.
continuous samples in
marine soils
Numerous types of oceanographic samplers, both open-tube and piston types, are
available for use when drilling from ships. Some of these depend upon free-fall
penetration and are limited in the depth of exploration. Drilling and sampling from the
ocean floor can be accomplished using specialized equipment deployed remotely from
portable equipment that is deployed in underwater vessels or on underwater platforms
operating on the ocean floor. The quality of samples obtained by most oceanographic
samplers is not high because of their large length to diameter ratio and because air/gas
in the dissolved state in the underwater environment comes out of solution when the
sample is recovered at the ground surface. For detailed information on underwater
sampling equipment, refer to ASTM STP 501 (ASTM 1972).
2-6.4 Field Logging and Boring Logs.
While monitoring drilling and sampling activities, an engineer, geologist, or experienced

driller prepares a field boring log to document the findings and observations. This field
logging is an important part of documenting the soil and rock conditions that exist at the
project site. A typical field log includes all the relevant information for the boring that
was completed, including a unique boring identification number, date of drilling,
personnel on-site, boring advancement method (i.e., auger, rotary wash, direct push,
sonic), depths where samples were obtained, type of samples (i.e., split-barrel and
Shelby tube), hammer type, raw SPT N values, water level observations, and
preliminary estimates of stratigraphy. If available, the global positioning system (GPS)
coordinates should be included. The field log provides a unique designation of each
recovered sample, whether disturbed or intact, as well as a field visual classification of
the sample in accordance with ASTM D2488.
The field log, the recovered samples, and lab/field testing results are used to produce
the final boring log, which represents the official engineering record of the drilling and
sampling efforts. The boring log provides the permanent, technical documentation of
the materials encountered during drilling, sampling, and coring. The geotechnical
engineer or geologist uses the results from the field and their training/experience to
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group samples/records together based on color, soil type, and SPT N values and
identify layers or strata, which may be consistently found in the adjacent companion
borings from the site. An example of the engineering boring log is shown as Figure 2-6.
In the final engineering boring logs, soil types are categorized according to a user- or
agency-specified soil classification system. The most common soil classification
systems in the United States include the Unified Soil Classification System (USCS)
(ASTM D2487 or D2488), the AASHTO system, and the United States Department of
Agriculture (USDA) system. In addition to the soil classification, the description should
also include color, relative density (e.g., loose, dense, etc.) or consistency (e.g., soft,
medium, hard, etc.), and the presence of organics, shells, peat and/or manmade
materials. Identification of these additional features may impact engineering
performance and may prove beneficial in subsequent construction/excavation phases of
the project.
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Figure 2-6 Example Geotechnical Boring Log

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2-7 PENETRATION RESISTANCE TESTS.
Penetration resistance tests are the most common in situ testing techniques for
characterizing subsurface conditions. The most common field penetration test remains
the Standard Penetration Test (SPT), which measures resistance to the penetration of a
standard, thick-walled drive sampler in an open borehole using a drop hammer. A more
controlled and increasingly popular test is the Cone Penetration Test (CPT), which
utilizes a standard cone-shaped instrument that is pushed at a standard constant rate
from the ground surface. Another common test is the flat plate dilatometer (DMT). This
device utilizes a robust steel blade that is pushed into the ground at a constant rate and
then periodically stopped to allow the controlled measured inflation of a flexible steel
membrane. In many parts of the United States, particularly when stiff soils and//or
granular soils are encounters, a dynamic cone penetration (DCP) test is performed by
driving a standard sized cone into the ground using a drop hammer. This section
provides information regarding these four penetration tests. Section 2-9 will address
other common in situ testing methods.
2-7.1 Standard Penetration Test (SPT).
The SPT was originally developed in the 1900s and proceeds by driving a thick-walled,
split-barrel (a.k.a., “split spoon”) sampler into the ground using incremental blows from a
drop hammer. The sampler is driven a total of 18 inches into the ground. The number
of blows required to drive the sampler the 12-inch vertical interval between 6 and 18
inches is referred to as an N value or the blow count. The procedure is presented in
ASTM D1586, and a schematic of the SPT is presented in Figure 2-7.
The SPT provides a disturbed sample of the tested material and generates useful data
that can be used to correlate to many engineering properties. Many factors can affect
the SPT results, and there are several vastly superior in situ testing methods.
Nevertheless, the test is still almost universally referenced and often required in the
United States. One reason for this is the large amount of historical (i.e., legacy) data
available. Numerous correlations have been published (see Chapter 8) and their use
along with SPT represents the Standard of Practice in many parts of the country.
2-7.1.1 Corrections to Field Blow Counts.
As an improvement on older donut and safety hammers, most current SPT programs
use an automatic hammer that does not rely on an operator-dependent cathead and
rope to establish the drop height of the hammer. Modern automatic deliver consistent
energy to the sampler, which should be measured periodically. The field-recorded N
values may be adjusted to reflect the energy of the specific hammer. The adjustments
are intended to correct the N value to the 60 percent hammer efficiency that is
assumed for the older equipment and historic correlations. The energy corrected value
( N 60 ) can also be normalized to an equivalent value at a vertical stress of one
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atmosphere. The “overburden corrected” or “normalized” blow count is labeled ( N1 )60 or

N1,60 . Several correlations to normalized blow count are presented in Chapter 8 and in
McGregor and Duncan (1998).
2-7.1.2 Advantages and Limitations.
The biggest advantage to the SPT is its near-universal acceptance and use in the
United States. As a result, there is a large data set that can be used for correlation.
However, SPT blow counts are affected by many operational procedures, the presence
of gravel, and by cementation between the particle grains. In clays, the blow count
does not reflect the influence of fractures or slickensides. Table 2-16 presents a
summary of the many operational factors that are known to influence the N value
measured in the field.
Figure 2-7 Standard Penetration Test (after NCHRP 2018 and Mayne 2012)
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Table 2-16 Factors Affecting the Standard Penetration Test and SPT results
(after Kulhawy and Mayne 1990)
Influence on
Cause Effects
SPT N Value
Sampler driven in disturbed, artificially Increases
Sampler driven above bottom of casing
densified soil greatly
Test not performed in original in situ
soil; soil may become trapped in
Inadequate cleaning of base of borehole Increases
sampler and may be compressed as
sampler is driven, recovery reduced
Hammer energy varies (generally
Careless measure of drop Increases
variations cluster on low side)
Hammer strikes drill rod collar eccentrically Hammer energy reduced Increases
Lack of hammer free fall because of ungreased
sheaves, new stiff rope on weight, more than two turns Hammer energy reduced Increases
on cathead, incomplete release of rope each drop
Coarse gravel or cobbles in soil Sampler becomes clogged or impeded Increases
Use of bent drill rods Inhibited transfer of energy of sampler Increases
Hammer energy varies (driller supplies Increases or
Hammer weight inaccurate
weight; variations of 5 – 7% common) decreases
Increases or
Careless blow count Inaccurate results
decreases
Correlations with standard sampler Increases or
Use of non-standard sampler
invalid decreases
Failure to maintain adequate head of water in borehole Bottom of borehole may become quick Decreases
2-7.2 Cone Penetrometer Tests (CPT).
The CPT involves hydraulically pushing an instrumented steel probe at a constant rate
to obtain a continuous record of the penetration resistance of the cone tip and the
frictional resistance of the soil. The CPT does not produce a borehole, samples, or drill
cuttings. The original test involved a mechanically operated cone, referenced as a
“Dutch” cone (DPT). The original equipment has been superseded, modified, and
improved to allow electronic measurements.
Most modern instruments also include a piezometer near the tip. When equipped with
the proper sensors and instruments, the routine performance of the CPT also allows the
measurement of temperature, vertical alignment, electrical resistivity, acoustic
emissions, and shear wave velocity.
Testing is currently conducted in accordance with ASTM D5778. The test can be
conducted without the use of a pore pressure measurement and is referenced simply as
the CPT. Alternatively (and commonly) the test is performed using a device to measure
pore pressures behind the tip of the probe while pushing. This is referred to as the
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piezocone test (CPTu). Recent advances have allowed the ability to measure the
propagation of shear waves using a seismic piezocone, which is referred to as a
seismic CPT (SCPTu).
2-7.2.1 Equipment and Testing Procedure.
The cone penetration test requires continuous hydraulic advancement of the probe and
the simultaneous recording of multiple electronic instruments. Specific equipment and
procedures necessary for performing a CPT are summarized as follows:
• Cone Penetrometer: A standard cone penetrometer is a 1.4-inch (35.7-mm)

diameter cylindrical probe with a 60o apex at the tip, which results in a projected
tip area of 1.55 in2 (10 cm2) and a 23.3 in2 (150 cm2) instrumented sleeve surface
area. Other sizes (both smaller and larger) are available. The size of a cone is
typically identified by the projected tip area (i.e., 10-cm2 cone or a 15-cm2 cone).
A variety of tip load capacities (i.e., 2-ton, 15-ton, etc.) are available.
• Drill Rig/CPT Truck and Cone Rods: A hydraulic actuator is attached to a truck
or drill rig that can provide sufficient reaction mass to advance the penetrometer
at a constant rate of 2 cm/second. This reaction can be provided using a
conventional drilling rig, but dedicated CPT trucks typically weighing 20 to 25
tons have become the standard.
• Water Pressure Transducer: Valuable information can be provided by measuring
the pore water pressure behind the cone tip during penetration. For a CPTu, the
water pressures are monitored using a transducer and porous filter element.
• Geophone: For the SCPTu, a geophone is located along the drill string at a
distance of approximately 20 inches (500 mm) above the cone tip. The
geophone detects shear waves generated at the ground surface at specific
vertical intervals. During advancement of the seismic cone, a shear wave is
generated at the ground surface. An average shear wave velocity of the soils
between the ground surface and the geophone can be calculated.
An example record from a CPT sounding is shown in Figure 2-8. This shows a
schematic of the CPT probe and the near-continuous vertical profile of cone tip
resistance ( qt ), sleeve friction ( f s ), and pore pressure at the u2 position (behind the tip).
2-7.2.2 Soil Classification with CPT.
Regardless of the specific type of cone penetration test probe (i.e., CPT CPTu, SCPTu),
the testing concept has gained near universal acceptance and interest. As shown in
Figure 2-8, a near-continuous vertical profile of the stratigraphic variations is obtained.
A variety of engineering parameters can be estimated from CPT results. Many
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correlations of CPT data to strength, compressibility, modulus, hydraulic conductivity,

and other properties are available, some of which are provided in Chapter 8.
The CPT is able to estimate the soil type of the deposit penetrated. A common method
for doing this is shown in Figure 2-9, which relates soil behavior type (SBT) to specific
CPT results. This type of correlation is extremely useful for site characterization and
subsurface stratigraphy. Other soil type correlations are available.
The CPT provides numerous advantages, due to its popularity in engineering practice
and proliferation of useful correlations to other engineering parameters. The test can be
performed quickly. The speed of operation allows considerable data to be obtained in a
short period of time, resulting in a continuous record of soil conditions. It is particularly
helpful in assessing variability in subsurface conditions across a site.
The major limitation of all cone penetration tests is that discrete samples are not
recovered for physical observation and companion testing. The cone can be difficult to
advance in dense or stiff to hard soils, and if the operator is not experienced, the probe
can be damaged (or destroyed) when encountering these materials. The specialized
equipment and the reliance on electronic instrumentation usually requires the services
of a specialty vendor to perform the tests.
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Figure 2-8 CPT - Example Test Record and Equipment
Figure 2-9 Nine Zone (Normalized) Soil Behavioral Chart for CPT
(after Robertson 2009 and NCHRP 2018)
2-7.3 Flat Plate Dilatometer.
The flat plate dilatometer test (DMT) was developed in Italy and introduced to the United
States practice in the 1980s (Marchetti et al. 2006). It has been widely adopted
worldwide and the testing procedures have been standardized in ASTM D6635. The
test involves pushing a relatively long and thin flat plate into the ground, generally in 9-
to 12-inch vertical increments and then inflating a flexible steel diaphragm while making
two or three specific measurements (i.e., A, B, and C). The A reading is the pressure
required to lift off the membrane from the face of the blade. The B reading is the
pressure required to move the center of the membrane a distance of 0.04 inch (1.1 mm)
into the soil. The C reading is an optional reading that can be taken by deflating the
membrane until the center of the membrane again contacts the face of the blade. Many
practitioners perform the DMT using the same specialized equipment for performing a
CPT. In most cases, the test is run without an excavated borehole, so no samples or
drill cuttings are produced. However, in some cases, the DMT is lowered into a
sampled borehole, advanced approximately 12 inches past the base of the borehole
and then inflated as described above. A schematic representation of the test is
presented in Figure 2-10.
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Figure 2-10 Flat Plate Dilatometer Test Schematic

(after NCHRP 2018 and Mayne 2012)
2-7.3.1 Equipment, Procedure, and Results.
A photograph of the flat plate dilatometer is presented in Figure 2-11(a), and the control
unit used to perform the test (i.e., control inflation and deflation of the membrane) is
shown in Figure 2-11(b). The dilatometer blade is nominally 3.75-inches (95-mm) wide,
0.60-inches (15-mm) thick, and 7.5-inches (190-mm) tall with a 30° apex angle at the
tip. A 2.4-inch (60-mm) diameter stainless steel membrane is used. The membrane is
typically 0.008-inches (0.20-mm) thick and requires careful calibration. The control unit
uses bottled gas (nitrogen) supply. A CPT rig is often used to push the dilatometer at a
rate of about 0.4 to 1.2 inches/second (1 to 3 cm/s). The vertical thrust is typically
monitored and recorded during the test.
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Figure 2-11 Flat Plate Dilatometer and Control Unit (Marchetti et al. 2006)
The DMT method and calculations are summarized in Figure 2-10. For more details
about the test procedure, refer to NCHRP (2018). The reduced DMT test results
produces three values, p0 , p1 , and u0 , from which the following DMT index values are
directly calculated for each test depth:
• Material Index ( I D ) =
( p1 − p0 ) / ( p0 − u0 ) , which is used to identify soil type;
• Dilatometer Modulus ( ED ) = 34.7 × ( p1 − p0 ) in units of atmospheres, which is a
measure of soil stiffness; and
• Horizontal Stress Index ( K D= ) ( p1 − p0 ) / σ 'v 0 , which is used to assess stress
history.
These three indices are typically plotted with respect to test depth to develop a near-
continuous vertical profile. Similar to the techniques used for the CPT, these directly
calculated values are used to estimate important engineering parameters, including
strength, compressibility, modulus, lateral earth pressure) by semi-empirical
correlations. A summary of correlations to the DMT results is presented Chapter 8.
The DMT provides multiple advantages. The test can be performed relatively quickly
using a variety of insertion equipment. The probe itself is relatively simple to maintain
and training is not particularly onerous. It provides some information regarding
horizontal stress and stiffness, which the SPT and CPT are unable to provide.
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A major limitation of the dilatometer is that the thin blade and particularly, the
diaphragm, can be easily damaged when penetrating soil with particles the size of
coarse sand or larger. The diaphragm can be replaced, but the break-in and calibration
procedures must be performed, and that can often be difficult to do in the field. The
specialized equipment needs to be maintained and properly cleaned between tests, as
erratic electric signaling has been experienced when humid conditions exist beneath the
membrane. Caution should be used when using dilatometer correlations directly for
engineering design parameters.
2-7.4 Dynamic Cone Penetrometer.
Like the CPT, the dynamic cone penetrometer (DCP) has seen a historical evolution. In
the United States, the device was developed in the late 1950s in the southeastern
United States, primarily to confirm near-surface conditions for spread footings and as a
potential surrogate for the SPT. This original, heavier DCP correlate closely with the
SPT blow count but was not formally standardized. The United States Army Corps of
Engineers (USACE) developed a lightweight DCP that correlates to SPT N values and
California Bearing Ratio, CBR (Webster et al. 1992). This lightweight device has seen
more widespread use and is standardized in ASTM D6951.
2-7.4.1 Equipment, Procedure, and Results.
A schematic of the lightweight DCP equipment and details of the cone tip are presented
in Figure 2-12. A drop hammer (either 17.4 or 10.1 pounds) strikes the anvil to drive in
the cone tip. The upper and lower shaft guide the hammer and transmit the driving
force to the cone tip. The 60° apex angle cone is at the bottom of the lower shaft. Both
fixed and disposable cone tips are available. An extraction jack may be needed to
remove the cone and shaft.
The DCP is normally conduct by two people. After seating the cone tip about 1.0 inch,
the cone is advanced incrementally by successive drops of the hammer, while holding
the device vertical. After each hammer blow, the penetration of the cone is measured
and recorded to the nearest 0.1 inch. The test is terminated when a target depth is
achieved, when the full length of the lower shaft is embedded, or when the total
penetration is less than 0.1 inch/blow for 10 successive hammer blows. The extraction
jack is then used to retrieve the embedded shaft/cone.
From the recorded test results, the DCP Penetration Index (DPI) is calculated, and
tabulated versus depth. A plot can be developed of the incremental values of DPI
versus the cumulative penetration depth, providing an indication of relative
stiffness/strength versus depth. Correlations to DCP are found in Chapter 8. 2
2 DCP is often used to represent the DCP Index in equations. This convention is followed in Chapter 8.
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Figure 2-12 Schematic of Dynamic Cone Penetrometer (DCP) Equipment

(after Webster et al. 1992)
The DCP is a simple, low-cost, easy-to-use tool, which is ideal for quick or very low-cost
results, or when site access is limited. The results of the DCP can be used as
compaction acceptance criteria. It can easily assess stratigraphy, particularly in the
delineation of soft and hard layers. The equipment is easy to maintain. Perhaps the
biggest advantage is that local and regional correlations can be easily developed and
updated as needed.
A significant limitation of the test is the shallow depth of penetration. Verticality of the
shafts when driving is critical, and operator experience is valuable. Because a donut-
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style drop hammer is used, the operator (and helper) need to be avoid “pinch points”
between the hammer and anvil.
2-8 GROUNDWATER MEASUREMENTS.
Because of its importance in geotechnical analysis, location of the groundwater table is

a key element of a subsurface investigation. During drilling, depths are typically
recorded at which the water is first encountered in the borehole and at which the water
level stabilizes after drilling. The latter is often recorded after the borehole remains
open for approximately 24 hours. In some soils, the sidewalls will collapse unless they
are confined or supported (e.g., sands beneath the water table). In these cases,
perforated pipe may be used as a temporary casing to protect against borehole collapse
while allowing water to flow through the perforations. Knowledge of the seasonal
groundwater fluctuation is important, and long-term measurements can be made by
converting borings to piezometers, which can vary from open wells to electronic
transducers.
Knowledge of the local groundwater regime is required to correctly interpret

groundwater measurements, especially those from piezometers. Groundwater can
occur at different elevations in the subsurface. It may be perched in isolated zones, or it
may be confined between different low-permeability strata. In addition, groundwater
flow can affect the interpretation of water levels. Where gradients are low and the
groundwater table is relatively horizontal, groundwater depths can be directly inferred
from piezometer measurements. However, where large gradients are present, the
water pressure measured at a point in the ground cannot be directly used to calculate
the vertical depth of water above that point. Knowledge of the seepage conditions is
required (see Chapter 7-6) to make this determination. Finally, a distinction must be
made between steady state and transient conditions. Steady state conditions can be
effectively monitored using all types of piezometers. However, under transient
conditions (e.g., rapid drawdown in dams, consolidation or swell in fine-grained soils),
pore pressures may be changing significantly with time and require instrumentation with
a fast response time, such as diaphragm type transducers.
2-8.1 Types of Standpipe Piezometer.
Groundwater level monitoring involves direct measurement of water levels within open
well, open standpipe piezometer, or porous element piezometers. The common types
of standpipe piezometer for monitoring groundwater levels are depicted in Figure 2-13
and summarized in Table 2-17. The type of standpipe that is selected depends on
preferences, regulations (if applicable), and the type of subsurface soils in which the
groundwater level will be measured.
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The three basic components of a standpipe piezometer include: (1) the tip or well
screen; (2) the standpipe itself; and (3) the standpipe seal. These can be installed by
hand at shallow depths, but in most cases are installed using a drill rig after the
completion of a boring in the soil or rock.
Figure 2-13 Open Piezometers
2-8.1.1 Open Well Piezometer.
A common groundwater monitoring technique is to install a standpipe, or water tight

pipe, within an open boring as shown in Figure 2-13(a). The standpipe has a perforated
tip or screen that allows water to enter are usually small-diameter (e.g., less than 2
inches) PVC plastic pipe but may be larger for environmental sampling applications. In
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an open well, the annular space between the pipe and borehole wall is filled with filter
sand or gravel almost to the ground surface. At the ground surface, a seal of cement
grout, bentonite slurry, or other low permeability material is placed above the filter sand
to isolate the well from surface water flow. Open wells are often called groundwater
monitoring wells and are commonly used for environmental applications when samples
of the groundwater are required.
Table 2-17 Types of Standpipe Piezometers

Piezometer Type Advantages Disadvantages
Simple and reliable; long experience record; good Slow response time, particularly in fine
Open well for coarse-grained soils; large diameter may be grained soils; unable to monitor distinct
piezometer required/needed for environmental monitoring and stratum exhibiting different groundwater
groundwater sampling levels; freezing potential in winter
Simple and reliable; long experience record; able to
Open standpipe Slow response time in low permeability
monitor distinct stratum exhibiting different
piezometer soils
groundwater levels; good for coarse-grained soils
Porous element Rapid response time; good for soils exhibiting Humid air entering tubing may impact
piezometer medium permeability; good for applications readings; time consuming to make
(hydraulic) impacted by electrical interference measurements.
Porous element Relatively expensive; temperature and
Rapid response; high sensitivity; suitable for
piezometer barometric pressure correction may be
automatic readout
(electronic) required; zero drift errors can arise
Because an open well has a full-length screen or a full depth filter zone, it is best suited
for measuring water levels in relatively homogeneous deposits with high permeability.
When multiple strata are crossed, the groundwater level corresponds the stratum with
the highest total head. A significant advantage of an open well piezometer is that it can
be cleaned and “developed” by flushing water from the standpipe into the formation,
which is critical for open well piezometers used for environmental applications.
2-8.1.2 Open Standpipe Piezometer.
An open standpipe piezometer is similar to an open well, except that the screen extends
only across a specific stratum of interest. Seals are installed above and below this zone
to only allow water to enter from the stratum of interest. An open standpipe piezometer
is shown in Figure 2-13(b). Outside of the screened test section, select backfill
materials are used but not necessarily filter sand. A seal is typically placed around the
open standpipe at the ground surface.
Multiple open standpipe piezometers can be installed to measurement groundwater

levels in multiple strata within a single borehole through careful installation of multiple
seals. This approach is sometimes referred to as a nested piezometer. The vertical
location (i.e., depth, thickness, and elevation) of each seal must be accurately
measured and recorded on the well log.
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A major disadvantage of open well and open standpipe piezometers is the long
equalization time that may be required for water to flow from the formation and fill the
piezometer. Until the groundwater level stabilizes in this manner, the readings are
inaccurate. To reduce the equalization time, the diameter of the standpipe can be
reduced to less than 0.5 inches, which decreases the volume of water needed.
2-8.1.3 Porous Element Piezometers.
The primary disadvantage of open well and open standpipe piezometers is the
potentially long equalization time for the groundwater level to stabilize since the riser
pipe must fill with a considerable volume of water from the formation. Porous element
or hydraulic piezometers have a ceramic or porous metal tip attached to a small-
diameter riser pipe (i.e., standpipe). Modern versions use a porous element with pore
sizes of <50 microns, so that the tip can be used in direct contact with fine-grained soils.
One of the primary advantages of the porous element piezometer is the relatively short
equalization time periods in fine-grained soils exhibiting low permeability. However,
water must still flow from the formation through the porous element and into the
standpipe to obtain accurate groundwater level measurements.
2-8.2 Multiple or Nested Installations.
Several standpipe piezometers may be installed in a single boring with an impervious

seal separating the different measuring zones. These are called nested piezometers.
This concept represents a cost advantage, as it reduces the number of borings (but
increases the difficulties/challenges of installing seals and specific elevations) and the
number of “obstacles” during for the contractor during construction. However, if
measurements are needed in zones with 10 feet or less of vertical separation,
piezometers should be installed in separate borings.
2-8.3 Measurement of Groundwater Levels.
Groundwater levels/elevations can be obtained by either direct or indirect methods. The

direct method includes: (1) surveying the elevation of the top of the riser pipe and/or the
ground surface; (2) measuring or calculating the “stick-up” of the riser pipe above the
ground surface; and (3) measuring the distance depth from the top of the riser pipe to
the water surface inside the open pipe or the standpipe. The elevation of the
groundwater can then be easily calculated. There are several methods to measure the
distance from the water surface to the top of the riser pipe; including a plumb bob, cloth
or metal surveyors' tapes coated with chalk, or commercially available electrical
indicators. Using these direct measurements, the water level can be established
generally within a tolerance of about 0.5 inch.
An indirect, but more accurate, method for measuring the depth of water in an open well
uses an electrical transducer capable of measuring water pressure. The transducer is
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attached to a hoisting cable and an electrical cable. Markings on the electrical cable are
used to measure the length of the cable in the open pipe. Commercially, these portable
systems are called water level indicators. The transducer is lowered to a depth typically
near the bottom of the open hole such that the transducer is submerged. The electrical
cable is attached at the surface to a readout unit that measures the pressure due to
water column above the tip of the transducer. From the unit weight of water
(i.e., 62.4 pcf or 9.81 kN/m3), the depth of water above the tip of the transducer is
calculated, which combined with the length of cable in the open pipe can be used to
calculate the groundwater level in the pipe. This type of transducer can be connected to
a data collection unit and groundwater levels can be automatically collected over time.
This capability is often quite efficient and is, in fact, a requirement when groundwater
pumping tests are performed and the time-dependent groundwater elevation as a
function of pumping rate is necessary for subsequent calculations of in situ permeability
of a formation (see Section 2-9.2). The indirect method can also be used with pore
pressure transducers that are grouted or sealed directly in boreholes without a
standpipe. More information on the transducers used with this type of piezometer is
provided in Section 2-10.4.
While being simple in concept, the techniques for measuring groundwater levels have
some inherent limitations. First, standpipe piezometers require access to the top of the
vertical riser pipe, which usually extends above the ground surface and may be easily
damaged during construction. If the riser is extended vertically during construction
(e.g., installed in a constructed embankment), the extension activities must be carefully
coordinated with the earthwork contractor. Manual or direct measurement of water
levels is time consuming and may adversely impact construction. The largest source of
error for standpipe piezometers is the lag time required for the piezometer to respond to
changing groundwater levels because water must flow from the formation into the
piezometer. For this reason, groundwater piezometers are intended to measure
hydrostatic groundwater levels and are inappropriate for time-dependent pore
pressures. Other sources of error that impact piezometer readings include the
possibility of direct introduction of precipitation into the riser pipe due to a missing or
vandalized cap, infiltration of surface water into the borehole, and the formation of gas
bubbles within the pipe. Indirect groundwater measurements by porous element
piezometers without a standpipe or pore pressure transducers can alleviate many of
these limitations.
2-8.4 Detection of Combustible Gases
Gas bubbles in groundwater can influence the measurement of groundwater levels.

Gas can exist in subsurface soils and pose other hazards. Specifically, methane (CH4)
and other combustible gases may be present in subsurface soils and rock, particularly
in sites near municipal solid waste (MSW) landfills, or at sites near or over peat bogs,
marshes, and swamp deposits. Methane is a dominating combustible gas in the
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subsurface because it is one of the primary by-products of the anaerobic decomposition

of organic material. The other dominant gas is carbon dioxide (CO2). Commercially
available portable instruments, referenced as landfill gas analyzers, are used to detect
the presence and concentration of combustible methane gas in landfill gas wells and
monitoring probes. These instruments sample the air/gas from the confined space in
wells and borings above the water table. The instrument generally detector indicates
the concentration of gases in the collected gas sample. The critical concentration limits
for methane is between 5 and 17 percent by volume. A concentration of less than 5
percent methane is considered too “lean” to burn or ignite and is referenced as the
lower explosive limit (LEL). A concentration of more than 17 percent methane is
considered too “rich” to ignite and cause a flash fire (i.e., explosion) and is referenced
as the upper explosive limit (UEL). If methane concentrations are measured within the
5 to 17 percent range, all possibilities of spark generation (e.g., pile driving, grinding,
welding, smoking) should be eliminated and a venting system should be considered, to
provide worker protection.
2-9 MEASUREMENT OF SOIL AND ROCK PROPERTIES IN SITU.
Field sampling and laboratory testing can sometimes be complemented or replaced by

in situ testing, which refers to measurements conducted on soil and rock “in place.” As
a general rule, an in situ testing program can be performed faster and in many cases at
a lower cost than most laboratory testing programs. As a result, this alternative has
seen growing popularity since the 1980s. The SPT, CPT, DMT, and DCP are four of
the most popular in situ testing methods, which are often correlated to engineering
parameters. This section discusses other in situ tests that measure strength,
stiffness/modulus, and permeability of existing soils, as well as the as compacted
properties of earthwork. Methods for in situ testing of rock are also discussed.
Although not universally adopted, many practitioners find that the pocket penetrometer
and the field torvane provide useful correlations to the shear strength as measured in
the laboratory. These tests may be performed on the soil exposed at the bottom of a
recovered Shelby tube sample. Practitioners who use these tests often correlate the
results to results from laboratory strength tests, to increase the value of the field test.
Although these tests are often used in practice, the results are very inexact and should
not be used for design.
2-9.1 Strength and Deformation Properties of Soil.
The pressuremeter test, vane shear test, and the plate load test are the most commonly
used in situ testing methods for assessing strength and stiffness of soils. A summary
and comparison of these tests is presented in Table 2-18. Where ASTM standards are
available, they have been included in the table.
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Table 2-18 In situ Testing Methods Used in Soil for Strength and Deformation
(after FHWA 2002)
Applicable
Method and Applicable
Procedure Soil Limitations / Remarks
ASTM No. Soil Types
Properties
Borehole drilled and the
bottom is carefully
Clays, silts, and Preparation of the borehole most
Pre-bored prepared. The
peat; marginal E , G, important step to obtain good
Pressuremeter pressure required to
response in results; good test for calculation
(PMT) expand the cylindrical mv , su
some sands of lateral deformation
ASTM D4719 membrane to a certain
and gravels characteristics
volumetric or radial
strain is recorded
Cylindrical probe with a
pressuremeter attached
behind a conical tip is
hydraulically pushed Disturbance during advancement
Full through the soil and of the probe will lead to stiffer
Displacement paused at select Clays, silts, and E , G, initial modulus and mask liftoff
Pressuremeter intervals for testing. peat mv , su ( )
pressure p0 ; good test for
(PMT) The pressure required calculation of lateral deformation
to expand the characteristics
cylindrical membrane to
a certain volume or
radial strain is recorded
Disturbance may occur in soft
Four- blade vane is sensitive clays, reducing
pushed into the bottom measured shear strength; partial
of a borehole. The Clays, some drainage may occur in fissured
vane is slowly rotated silts and peats if clays and silty materials, leading
Vane Shear until the maximum undrained to errors in calculated strength;
Test (VST) torque required for conditions can su , St , σ ' p rod friction needs to be
ASTM D2573 rotation is recorded. be assumed. accounted for in calculation of
The vane is rapidly Not for use in strength; vane diameter and
rotated for 10 turns, granular soils torque wrench capacity need to
and the residual torque be properly sized for adequate
is recorded. measurements in various clay
deposits
A circular, rigid steel All soils and
plate is hydraulically rock, Limited depth of influence; short-
Plate Load Test pushed into the soil and particularly term test will not capture
(PLT) the relationship helpful in qult , k s consolidation impacts; not
ASTM D1196 between bearing stress unbounded typically used as part of
and vertical settlement base aggregate geotechnical site investigation
is recorded for pavements
Note: E = elastic modulus; G = shear modulus; mv = coefficient of volume compressibility; su = undrained shear
strength; St = sensitivity; σ ' p = preconsolidation stress, qult = ultimate bearing capacity; k s = modulus of subgrade
reaction
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2-9.1.1 Pressuremeter Test.
The pressuremeter test (PMT) was first developed in about 1955. In the PMT, a
membrane is inflated from a cylindrical probe against the sidewalls of an open borehole.
The radial expansion is measured, and this response is used to calculate specific
strength and deformation properties of the subsurface soils. PMT can be performed in
a wide range of materials, including sands, residual soil, tills, and soft rock, that are
usually difficult to sample. The “traditional” is called the “pre-bored” or “Menard”
pressuremeter. Other types include the self-boring pressuremeter (SBPMT) that
includes its own cutting shoe and does not require an existing borehole and the full-
displacement or cone pressuremeter (CPMT) that is pushed into the ground, usually
behind a piezocone. A schematic of the PMT equipment and test is shown in Figure
2-14.
Figure 2-14 Schematic of Pressuremeter Test (after NCHR 2018 and Mayne 2012)
As shown in Figure 2-14, the equipment needed to perform a PMT includes of an

expandable cylindrical probe, rubber membrane, outer slotted sleeve, pressure lines,
and control unit. The PMT procedures are defined in ASTM D4719 and can be
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consulted for further details on the test method. The inner membrane is hydraulically
expanded to obtain an expansion pressure versus volume curve. During loading,
disturbed soil in the borehole wall first compresses, followed by a pseudo-elastic
response and then plastic yielding. After plastic yielding is induced, a creep test is
performed by holding pressure constant until the lateral expansion falls below a
threshold. The PMT concludes with an unload-reload cycle to better define the elastic
properties.
Typical PMT pressure vs. volume change results are shown in Figure 2-15 along with
definitions of the characteristic pressures. For comparison, a typical SBPMT test result
is presented in Figure 2-16.
Figure 2-15 Typical Result and Characteristic Pressures from Pressuremeter

Test (after FHWA 2002)
2-9.1.1.1 Test Interpretation.
Pressuremeter tests have been used to estimate the coefficient of lateral earth pressure
at rest ( K O ), the soil stiffness; and undrained shear strength ( su ).
• Coefficient of Lateral Earth Pressure at Rest ( K O ): Upon initial inflation, the

membrane will expand to contact the borehole sidewalls. In the PMT, p0 is
related to the in situ total horizontal stress, which combined with the vertical
stress allows K O to be calculated. Due to unloading effects and disturbance
during drilling the borehole, the accuracy of this calculation is questionable. To
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accurately assess the in situ lateral stress and K O , a self-boring pressuremeter

should be considered. Point A in Figure 2-16 is a relatively accurate
representation of the total horizontal stress in the ground.
• Stiffness: The elastic stiffness of the soil can be estimated by the slope of the
unload-reload pressuremeter curve where the response is assumed to be nearly
elastic. One technique calculates the pressuremeter modulus ( E p ), while
another uses cavity expansion theory to calculate shear modulus, G (Gibson and
Anderson 1961, Windle and Wroth 1977).
• Undrained Shear Strength ( su ): Methods exist to estimate the undrained shear
strength from pressuremeter results, but the resulting values are less reliable
than those from other in situ tests, such as the cone penetration test or the vane
shear test.
Figure 2-16 Example Result from Self-boring Pressuremeter Test in Clay

(after Windle and Wroth 1977)
2-9.1.1.2 Limitations.
Pressuremeter testing is sensitive to test procedures. In very soft soils and in sands, it
may be difficult to maintain borehole stability before the probe is inserted. In these
cases, a self-boring pressuremeter may be necessary. Irregularities in the wall of the
borehole wall also affects test results, and the self-boring pressuremeter eliminates
some of this disadvantage. The SBPMT usually requires a specialist familiar with the
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test and the instrument. Pressuremeter test interpretation has a theoretical basis and,
therefore, either drained or undrained conditions need to be maintained. The
pressuremeter is relatively long (i.e., generally greater than 2 feet in length) and results
reflect an averaging of the soils over this length. For this reason, the PMT in best used
in relatively homogenous deposits, and it not expected to provide reliable parameters in
stratified soil. Pressuremeter equipment has many moving parts and requires
maintenance and careful handling.
2-9.1.2 Vane Shear Test.
The vane shear test (VST) is a popular and reliable in situ test that has been in use
since the 1940s. The VST involves the use of a simple four-sided blade (i.e., vane) that
is pushed into the ground and then rotated to evaluate the undrained shear strength and
sensitivity of soft to stiff clays and silts. The use of the VST should be limited to soils in
which slow (i.e., ~6° / min) rotation of the vane represents undrained shearing. A
schematic of the VST equipment and operation is presented in Figure 2-17. At failure,
the vane will cut a “cylinder” of soil equivalent to the outside dimensions of the vane and
the torque will reduce. The vane is often rotated 10 more revolutions, and the residual
torque is measured.
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Figure 2-17 Schematic of Vane Shear Test (after Mayne 2012)
The undrained shear strength ( su , fv ), the remolded undrained shear strength ( sur , fv ), and
the sensitivity ( St , fv ), can be obtained from the VST. During rotation, the maximum net
torque ( Tmax ) is measured and the undrained shear strength for a “standard” rectangular
vane with an H / D ratio of 2 is as follows:
6Tmax
su , fv = (2-1)
7π D 3
where:
D = diameter of the vane.
To measure the remolded undrained shear strength, the torque reading ( Tres ) is taken
during rotation of the vane following five to ten rapid turns of the vane. The remolded
strength is calculated by replacing Tmax with Tres in Equation 2-1. With knowledge of the
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peak and remolded values for undrained shear strength, the sensitivity of the soil from
vane shear tests can be calculated by:
su , fv
St , fv = (2-2)
sur , fv
The undrained shear strength from the VST overpredicts the shear strength mobilized in
failures of embankments, shallow footings, and slopes constructed on soft clay. In
order to account for this, su , fv should be adjusted by a correction factor ( µ R ), which is a
function of the PI of the soil tested. Three different vane correction methods are given
in the ASTM specification. The corrected shear strength can be calculated by:
= su , fv × µ R
su , field (2-3)
where:
µ R = vane correction factor.
The VST has proven to be a very reliable and repeatable in situ test and enjoys
widespread popularity due to cost and efficiency. It is perhaps the best device to
measure the in situ strength of soft to medium clays ( su < 2000 psf). The biggest
limitation of the VST is the types of soil where it can be used. The VST cannot
accurately assess the strength of fissured clays, clays with significant amounts of sand
or gravel, and soils with relatively thin laminations. The data for the VST is reduced
based on the assumption of undrained conditions. Therefore, soils that might allow
partial drainage during shear are problematic.
2-9.2 Hydraulic Conductivity of Soil.
Hydraulic conductivity is the most variable of all the material properties commonly
measured and used in geotechnical analysis, with the range extending to more than ten
orders of magnitude. Accurate measurement of hydraulic conductivity is very sensitive
to the type of soil, the disturbance of the soil, site stratigraphy, and the variability of the
soil deposit across the site. Laboratory testing of the hydraulic conductivity of soil, even
on samples of minimally disturbed recovered samples, may not reflect the hydraulic
conductivity of the natural deposit because the lab sample is quite small and certainly
not representative of a geologically placed and weathered material. If the soils at the
site are relatively uniform and can be sampled with minimal disturbance (i.e., uniform
clay soils), laboratory testing may be sufficient and adequate. However, for deposits of
coarse-grained materials, intact sample are nearly impossible to obtain, and in situ
hydraulic conductivity tests are commonly used. In particular, in situ tests are important
for uniform coarse-grained deposits. Correlations based on grain-size distribution are
also very common for coarse-grained deposits (see Chapter 8).
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The following five physical characteristics influence the performance and applicability of
in situ hydraulic conductivity tests: (1) water level position, (2) type of soil or rock, (3)
depth of the test zone, (4) hydraulic conductivity of the test zone, and (5) heterogeneity
and anisotropy of the test zone.
The difficulties inherent with in situ hydraulic conductivity testing require that great care
be taken to minimize sources of error and to correctly interpret and compensate for
deviations from ideal test conditions. Many of these difficulties can be overcome by
planning tests to isolate specific zones of material assumed to be uniform.
In situ hydraulic conductivity tests are most commonly used in geotechnical engineering
investigations for dams, hydraulic barriers, geo-environmental projects, and project with
a strong hydrogeologic component. For strictly geotechnical applications, steady-state
testing is much more common than transient state testing. In addition, saturated
hydraulic conductivity testing is more common than in situ testing for unsaturated
conditions. A brief summary of the four most common types of in situ hydraulic
conductivity tests is presented in Table 2-19.
A considerable amount of skill is necessary to correctly measured the hydraulic

conductivity of soil in situ. Specialty firms will likely provide more accurate and prompt
results compare with the personnel used for routine drilling and sampling. Specialty
firms are staffed by geotechnical and hydrogeologists who routinely conduct in situ
permeability tests.
Table 2-19 Summary of In situ Test Procedures for Measuring Hydraulic

Conductivity of Soil Deposits
Type of Test Description Advantages and Disadvantages
Water is added to an open-ended pipe Can be performed in saturated and partially
Constant Head (cased borehole) at a constant rate. saturated soils. Difficult to perform on soils with a
Test Water level in hole maintained at a very high or very low k. Only a small zone of soil is
constant level. tested (if unscreened).
Variable Head
An interval of a borehole is screened. A
Rising Head Construction and development of the well is more
“slug” of water removed from or added to
and Falling difficult than constant head test. Data reduction
the borehole. Elevation of the water
Head Tests can be complex.
level recorded over time.
(ASTM D4044)
Best for deep explorations. Can be conducted
Borehole section isolated by or sealed off
above or below the water table. Can be used to
Pressure Tests with “packers.” Elevated pressure can be
measure the hydraulic conductivity of fractured
applied to achieve increased flow.
rock.
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Install pumped well and observations

More expensive than other methods. Hard to
wells radially from the pumped well.
justify increased cost for common geotechnical
Record the amount of water pumped and
projects. Provides addition information regarding
Pumping Tests the elevation of water in the wells is over
aquifer transmissivity and storativity. Tests a larger
time. Use analytical or numerical
volume of soil than other methods. Long testing
analysis to determine the hydraulic
times.
parameters of the soil deposit.
2-9.3 Engineered Fill and Earthworks.
The as-compacted moisture/water content and unit weight (often referred to as density)
of compacted soils is a significant component of geotechnical practice involving
constructed engineered fill or earthworks (e.g., dams, embankments, etc.). For
purposes of this discussion, the terms moisture content and unit weight will be used.
Geotechnical engineers have long recognized that many of the desired engineering
properties of both fine- and coarse-grained compacted materials depend on the
compaction moisture content and/or unit weight. The compaction water content is much
more important for fine-grained as opposed to coarse-grained soils. This section
focuses on the use and limitations of the numerous techniques used in practice for
these important measurements.
It should be noted that compactors are current available that can measure the stiffness
of soils “on-the-fly” during compaction (McGuire et al. 2009). These special compactors
are becoming very popular in earthwork construction, and their use has supplanted
many of the older methods of compaction quality control incorporated into earthwork
specifications.
2-9.3.1 Measurement of As-Compacted Soil Unit Weight.
Four general strategies can be used to assess the as-compacted unit weight of soils: (1)
displacement methods; (2) drive-cylinder methods; and (3) nuclear gauge methods; and
(4) non-nuclear gauge methods. All of these methods use measurements from the
ground surface to assess the near-surface characteristics of the soil. This section
provides a discussion of each of these strategies and methods.
2-9.3.1.1 Displacement Methods.
Of the displacement methods, the sand displacement and water balloon displacement
techniques are the most widely used because of their simplicity, applicability to a wide
range of material types, and their historical performance and record of accuracy. In
both cases, a known weight of soil is excavated from the ground. Either sand or water
of a known unit weight is used to measure the volume of the excavated hole. The
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results are more accurate when a large volume is measured. Displacement methods
measure the total unit weight of the soil. The water content of the excavated soil must
be determined to calculate the dry unit weight.
The sand displacement method, usually referenced as the “sand cone” method, is the
most frequently used displacement test to assess in situ dry unit weight. A schematic of
the sand cone apparatus is shown in Figure 2-18. Originally standardized in 1958
(ASTM D1556), the sand cone method remains the recognized reference test for all
other methods used to assess in situ soil unit weight. Consistent results strongly
depend on operator experience and care in performing the test.
The water balloon displacement test (ASTM D2167) uses the same principle as the
sand cone. The excavated hole in the soil is lined with a balloon (i.e., watertight, thin
membrane), which is filled with pressurized water from a volume-calibrated container as
shown in Figure 2-19. The water balloon method should not be used in soils that
contain significant amounts of gravel that can potentially puncture the balloon. The
water balloon method generally provides consistent and accurate results when
performed correctly, although not as consistent as the sand cone (Berney et al. 2013).
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Figure 2-18 Schematic of Equipment and Process to Perform a Sand Cone Test
(after Dunn 2017)
2-9.3.1.2 Drive-Cylinder Method.
The drive cylinder method (ASTM D2937) is a convenient and rapid technique to obtain
the as-compacted total unit weight of soil. A slide hammer is used to drive a relatively
short thin-walled tube (e.g., shortened Shelby tube) into the ground to obtain a sample.
After the cylinder is driven, the soil-filled cylinder is carefully dug out of the ground, and
the top and bottom of the sample is trimmed flush with the ends of the tube. The inside
volume of the tube is the volume of the sample. A conventional drive cylinder is shown
in Figure 2-20. This method can be used as long as the soil will remain in the cylinder,
most notably fine-grained soils containing little or no gravel and moist, fine sands that
exhibit apparent cohesion. The method cannot be used in soils that contain gravel, as
the cylinder can be easily damaged. The drive cylinder measures the total unit weight,
and the water content must be determined by drying the soil sample.
Figure 2-19 Schematic of Equipment to Perform Water Balloon Test

(after Dunn 2017)
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Figure 2-20 Schematic of Drive Cylinder (after ASTM D2937)

2-9.3.1.3 Nuclear Gauge Method.
The use of nuclear technology to assess soil unit weight and moisture content
commenced in the early 1950s (Burgers and Yoder 1962). As shown in Figure 2-21,
the nuclear source (usually Cesium-231) is housed near the tip of a rod that is inserted
in a prepared hole at the test location. A low-power nuclear source is used to emit
gamma rays through the soil to detectors in the gauge, and the detection rate is
correlated to total unit weight of the soil. When not in use for testing, the nuclear source
is protected inside of the shielded gauge. Procedures for using a nuclear gauge to
measure unit weight and water content can be found in ASTM D6938. Proper, regular
calibration of nuclear gauges is essential to obtain consistent and accurate results.
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Figure 2-21 Schematic of Nuclear Gauge in Direct Transmission Mode

(after NRC 1996)
Guidelines for the safe operation and licensing are provided by the Nuclear Regulatory
Commission (NRC) in NUREG/BR-0133, Working Safely with Nuclear Gauges (NRC
1996). Technicians working with nuclear gauges should be trained in the safe operation
and handling of the gauge. They also need to wear a dosimeter that is periodically
tested to confirm no unexpected exposure, although the risks are relatively minor due to
the low energy emitted and the safety precautions built into the gauge.
2-9.3.1.4 Non-nuclear Gauge Methods.
Because of the NRC requirements regarding licensing and transportation of nuclear

gauges, several agencies (including the U.S. military) have been researching reliable
alternatives to the nuclear gauge. Interestingly, these groups are also assessing direct
measurement of the in situ strength and/or stiffness to replace the traditional density-
moisture content specification for compacted soil. Table 2-20 provides a summary of
various techniques and their performance compared to the “traditional” nuclear gauge
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based on field assessments by Berney et al. (2013, 2016). These devices are
commercially available and operate using different principles.
Table 2-20 Comparison of Non-nuclear Technologies for Assessing Soil Density

(Berney et al. 2013, 2016)
Non-nuclear Device Measuring Technology Field Performance
Electromagnetic pulses and time domain Requires third-party software,
Moisture-Density Indicator
reflectometry used to determine total unit difficult installation without causing
(M-DI)
weight and moisture content disturbance
High radio frequency waves measure the Highest precision, but average
Electrical Density Gauge soil’s dielectric constant, capacitance, accuracy, requires extensive
(EDG) impedance, total unit weight, and moisture calibration to site-specific soil to
content are calculated establish accuracy
Accurate but imprecise
Electrical impedance spectroscopy (EIS)
Soil Density Gauge measurement of density, requires
used for non-contact measurement of total
(SDG) calibration with site specific soil,
unit weight and moisture content
(i.e., grain size and Atterberg limits)
Low level nuclear source (Cesium-127) used
License-exempt soil density High correlation to results from
to measure total unit weight, exempt from
gauge traditional nuclear gauge
NRC licensing because of source size
Lightweight Falling
Various methods used to assess soil None of the devices directly
Deflectometer (LFD);
stiffness and/or strength, results can be provide unit weight or moisture
Dynamic Cone Penetrometer
correlated to unit weight and moisture content, poor correlation to unit
(DCP); Impact Soil Tester
content weight and moisture content.
(IST), Surface Stiffness (SS)
2-9.3.2 Measurement of As-Compacted Soil Moisture Content.
The standard method to measure moisture content is the laboratory oven and is the
appropriate basis of comparison for all other methods. Unfortunately, drying in the oven
requires a 24-hour time period at a temperature of 110°C ± 5°C (ASTM D2216), which
is too slow for quality control of engineered fill. At least ten alternative methods have
been developed for in situ moisture content evaluation. These methods can be
classified either as: (1) gravimetric, in which the soil is actually heated and dried; or (2)
indirect, in which the moisture content is correlated to another parameter, as the soil is
not physically dried.
Gravimetric methods for field measurement of moisture content are compared to ASTM
D2216 in Table 2-21. In all four cases, the soil is physically dried to obtain total mass
and dry mass measurements from which moisture content is calculated.
Table 2-21 Gravimetric Testing Methods for Moisture Content

(after Berney et al. 2013)
Technique
Summary Comments
(Standard)
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Laboratory Standard by which all other methods are

Sample dried in conventional oven at
Oven compared, requires long testing time period
temperature of 110°C for 24 hours
(ASTM D2216) (about 24 hours)
Standard Results sensitive to specific microwave and
(700-Watt) Sample heated and weighed repeatedly in 1- type of soil, use of 1-minute cycles minimizes
Microwave minute intervals until dry weight is constant chance of overheating, requires electricity,
(ASTM D4643) relatively rapid testing time (≈ minutes)
Field Sample heated and weighed repeatedly in 1-
Microwave minute intervals until dry weight is constant,
Same as standard microwave
(low power) more heating cycles are required compared to
(ASTM D4643) standard microwave
Sample heated in a container exposed to direct
Direct Heating flame from a field stove, heating and cooling Heating time periods will vary depending on
(ASTM D4959) cycles are used until specimen achieves size of test specimen
constant weight
Sample dried under halogen lights or infrared
Moisture Not traditionally used in geotechnical
heating elements on a dedicated laboratory-
Analyzer applications due to small size of test
scale, internal controls periodically weigh
(N/A) specimen (i.e., <50 gm)
specimen and terminate test automatically
A number of indirect methods have been developed to assess moisture content without
physically drying the soil. As summarized in Table 2-22, these methods use a surrogate
for temperature (i.e., gas pressure, dielectric constant changes, electrical impedance,
etc.). Moisture content has a benchmark test that can and should be used - the
laboratory oven. Each method has specific advantages and disadvantages, which can
be expressed in statistical terms of bias, accuracy, and precision. Table 2-23
summarizes of comparison of the techniques (Berney et al. 2012, 2013).
Table 2-22 Indirect Testing Methods to Assess As-compacted Moisture Content

(after Berney et al. 2012)
Technique
Summary Comments
(Standard)
A neutron source is used to determine Most common method used in compaction
Nuclear Density
hydrogen ion concentration by “backscatter” quality control, results can be affected by
Gauge (NDG)
method, hydrogen is assumed to be in form of chemical composition of the soil, results
(ASTM D6938)
water in soil, measures upper 4 inches biased by the soil closest to the surface.
Electrical Very dependent on type of soil and requires
High-frequency radio waves are used to
Density Gauge calibration of the equipment to the site-
measure the dielectric constant of soil, which is
(EDG) specific soil, calculates average moisture
correlated to moisture content
(ASTM D7698) content in a relatively large block of soil
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Similar to EDG because both measure

Soil Density Non-contact test uses electrical impedance
dielectric constant, results in average
Gauge (SDG) spectroscopy (EIS) to assess dielectric constant
estimate of moisture content, affected by
(N/A) of the soil
near-surface moisture conditions.
Gas Pressure A calcium carbide reagent reacts with water to Uses a small sample size that must be
Moisture Tester produce acetylene gas within a sealed pressure carefully selected, the acetylene gas by-
(a.k.a., Speedy) vessel, gas pressure is proportional to the product must be carefully vented, reagent
(ASTM D4944) moisture content must be kept dry, rapid test results
Electromagnetic An electromagnetic probe is used to measure
May be part of a license-exempt soil density
Gauge the dielectric constant by “fringing field
gauge, similar comments to EDG
(N/A) capacitance”
Table 2-23 Bias, Accuracy, and Precision of Test Methods for the As-Compacted
Measurement of Moisture Content (after Berney et al. 2012, 2013)
Bias Accuracy Precision

Method
(Slope) ( R2 ) (Standard deviation)
Lab Oven 1.00 0.995 0.087
Standard Microwave 1.11 0.973 0.109
Field Microwave 0.924 0.976 0.145
Direct Heating 1.027 0.964 0.159
Moisture Analyzer 0.731 0.915 0.044
Nuclear Density Gauge 0.922 0.970 0.091
Electrical Density Gauge 1.01 0.866 0.253
Soil Density Gauge 0.979 0.936 0.175
Gas Pressure Tester 1.405 0.867 0.056
Electromagnetic Probe 1.096 0.857 ~0.10 (similar to NDG)
Explanation Slope of trend: slope > 1 Scatter about the average

Measure of scatter about
(for comparison of field and indicates over-prediction, 2 value, standard deviation
trend, R = 1 for results
oven measurements of slope < 1 indicates under- approaches 0 for more
with no scatter
moisture content) prediction precise results
2-9.4 Rock Properties.
Strength and stiffness tests on rock core tend to reflect the characteristics of the “intact”
rock, while the engineering performance of rock in the field is governed by the rock
discontinuities (Deere and Deere 1988; Bieniawski 1989). Therefore, in situ tests that
include rock discontinuities and assess their impact are useful. The following tests are
discussed in this section: plate load, flat jack, rock dilatometer, rock borehole shear,
field direct shear, and rock joint hydraulic conductivity. Because of the specialized
nature of this testing and the cost of the equipment, many of these in situ testing
methods for rock are subcontracted to a specialty contractor.
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2-9.4.1 Strength and Stiffness Tests on Rock Masses.
Plate load testing (ASTM D4394, D4395) on rock is an in situ test method for that
evaluates the rock mass stiffness (Goodman 1989, George et al. 1999). The test can
also assess the strength of medium to low strength rock. The results from rock plate
load tests are presented in terms of stress vs. displacement (Figure 2-22). Tests often
include a series of loading and unloading cycles to help isolate the influence of fractures
and discontinuities. Generally, the results are analyzed using solutions based on elastic
theory to calculate an equivalent modulus, E ' (Hoek 2007, Goodman 1989).
Figure 2-22 Example Plate Load Test Result on Intact Limestone

(after NCHRP 2017)
The plate bearing test requires a large reaction system to apply the required force. As
an alternative, the flat jack concept uses a relatively thin (i.e., ~0.25 inches thick), flat,
hydraulic jack that is inserted into a slot in the rock, thus allowing the rock to provide its
own reaction. The flat jack test (ASTM D4729) is performed to assess the in situ state
of stress in the rock and the rock mass stiffness.
For a geotechnical engineer who is familiar with in situ soil testing, a rock dilatometer is
a misnomer, because it is actually a high-capacity pressuremeter (see Section 2-9.1.1).
Operating procedures for a rock dilatometer are generally similar to those identified in
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ASTM D4719. The rock dilatometer test involves placing a long, cylindrical probe into a
rock corehole and inflating a membrane on the probe. The membrane is expanded
laterally while measuring the radial deformation. The results are used to evaluate rock
mass stiffness.
The rock dilatometer may not be able to sufficiently stress the rock without rupturing the
membrane. The borehole jack, commonly called a Goodman jack, overcomes this
problem using small internal hydraulic jacks to induce lateral force across opposing
curved steel platens that each stress a 90°sector of the borehole wall over a length of 8
inches. Equipment description and operating procedures are presented in ASTM
D4971. The borehole jack can be used in boreholes core with NX-size coring
equipment. The hydraulic system used for the borehole jack can generate up to 10,000
psi, so it can be used on virtually any rock. The borehole jack is a common in situ rock
test that does not require extensive experience to perform and obtain reliable results.
The interpretation of the results presents some unique challenges, but also provides
some insight regarding the in situ response of rock.
The rock borehole shear test is an alternative in situ borehole method for relatively weak
rock or rock that is easily disturbed upon drilling and coring (e.g., weathered rock,
fractured rock, shale, etc.). This test is a modification of the Iowa borehole shear test
originally developed for soil (Yang et al. 2006). The rock borehole shear test measures
the shear strength of rock. While there is no recognized testing standard, guidelines for
the rock borehole shear test are presented in Lutenegger and Hallberg (1981). The
device is lowered down the hole to the desired test elevation. Hydraulic pressure is
applied to shear plates to obtain the desired normal stress, and a tether is pulled to
create a shear force on the plates. The normal stress and shear stress are recorded.
The calculated shear stress values for each of three or four normal stresses provide the
data points to construct a strength envelope.
2-9.4.2 Direct Shear Tests on Rock Discontinuities.
Rock mass behavior governs the engineering performance, as dictated by

discontinuities, which can vary from clean fractures with a certain surface roughness
caused by asperities to weathered rock joints to clay-filled fractures. Many field
applications load those discontinuities in shear (e.g., rock slopes, tunnel side walls and
crown). Undisturbed, representative samples of discontinuities can be difficult to obtain.
Direct shear testing of rock discontinuities has been developed similar to direct shear
testing for soils. The test equipment is highly specialized and quite large. Large-scale
in situ direct shear tests on rock are performed on exposed from natural rock outcrops,
in tunnels, or in excavations. In almost all cases, the test is performed to evaluate the
strength of the discontinuity.
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A description and discussion of the use of large-scale direct shear testing of rock by
USACE is presented in Zeigler (1972). Standardized testing procedures are presented
in ASTM D4554. A confining ring is place around the in situ rock specimen such that
the discontinuity is parallel to the ring. The specimen is encased in the ring using
plaster of Paris or hydrostone. Normal and shear loads are applied perpendicular and
parallel to the discontinuity, and displacements are measured.
While the performance of large-scale in situ direct shear tests on rock discontinuities
can be daunting and difficult, the interpretation is similar to the conventional direct shear
test performed on soil specimens. Pairs of normal and shear stress are plotted to
define the failure envelope and shear strength parameters. In rock and along rock
discontinuities, it is possible to measure the peak and residual strength of the
discontinuities (Goodman 1970).
2-9.4.3 Hydraulic Conductivity of Rock Discontinuities.
Water flows through rocks occurs mostly in open voids, fractures, joints, and other
discontinuities, which contribute to the “primary” porosity of the rock. Water flow in this
regime may be turbulent instead of laminar and may not be governed by Darcy’s law,
making quantification of water flow in rock discontinuities challenging. In the 1930s, the
lugeon was introduced to quantify water flow in jointed rock (Houlsby 1976). A lugeon is
defined as the flow of one liter of water per meter per minute under a pressure of 10
bars (145 psi) in a constant head double packer test as shown in Figure 2-23. For this
situation, a lugeon is approximately 1×10-5 cm/s for laminar flow conditions.
To assess the flow regime for water in rock joints, the five-step test method summarized
in Figure 2-23 was developed. Based on the five-step test results, the rock is
characterized as being in one of five groups (Houlsby 1976).
• Group A – Laminar Flow: Lugeon values relatively constant through all five steps
• Group B – Turbulent: Lowest lugeon occurs at highest pressure
• Group C – Dilation: Highest lugeon occurs at highest pressure
• Group D – Wash-out: Lugeon increases as test progresses
• Group E – Void Filing: Lugeon decreases as test progresses
These characterizations are commonly used to select the appropriate grouting strategy
for hydraulic barriers in jointed rock. The interpretation of the lugeon test and physical
characterization of the jointed rock is shown in Table 2-24.
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Figure 2-23 Double Packer Set-up to Conduct Five-step Lugeon test

(after Clayton et al. 1995)
Table 2-24 Interpretation of Lugeon Test Results (after Tunbridge 2017)

Lugeon Approx. Hydraulic Condition of Rock Mass Report Precision
Classification
Range Conductivity (cm/s) Discontinuities (lugeons)
<1 Very Low < 1 x 10-5 Very Tight <1
1-5 Low 1 x 10-5 - 6 x 10-5 Tight ±0
5-15 Moderate 6 x 10-5 - 2 x 10-4 Few partly open ±1
15-50 Medium 2 x 10-4 - 6 x 10-4 Some open ±5
50-100 High 6 x 10-4 - 1 x 10-3 Many open ±10
>100 Very High > 1 x 10-3 Open closely spaced or voids >100
2-10 FIELD INSTRUMENTATION AND MONITORING.
Monitoring the performance of geotechnical structures is a vital consideration for

individual projects and has helped guide the evolution of the practice (Peck 1969).
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Monitoring field performance starts with an assessment of what “performance” is

anticipated. Once this is established, instruments may need to be installed before,
during, and/or after construction or after a failure as part of the forensic investigation to
understand the failure mechanisms. Finally, certain instruments may be integral
components of early warning systems for sensitive structures. This section summarizes
the types of geotechnical instrumentation and their operations to help guide the
geotechnical engineer in selecting the most appropriate instrument for a given project.
In-depth details regarding geotechnical instrumentation can be found in USACE (1987,
1995b), Bartholomew et al. (1987, 1987a), FHWA (1988) and most notably Dunnicliff
(1993). The rapid evolution of the various measurement technologies and the
recognition of the importance of performance monitoring will undoubtedly result in
further expansion and utility of geotechnical instrumentation over the coming years.
2-10.1 Operating Concepts for Geotechnical Monitoring Instruments.
Making a measurement of engineering performance involves using some type of

instrument or transducer for obtaining the measurement. The major types of
instruments are summarized in Table 2-25. The transducers introduced in this table are
incorporated into instruments that are used to make specific measurements. These
measurements may include: (1) deformations (e.g., horizontal movement of a landslide,
vertical settlement, tilt of retaining wall), (2) pore pressures in soil (e.g., excess pore
pressure due to consolidation or static water levels in wells), (3) earth pressures (e.g.,
pressures acting on earth retaining structures), (4) loads (e.g., strut loads in braced
excavations, anchor loads on tiebacks, vertical loads for plate load tests), (5)
temperature (e.g., frost penetration, thermal-induced stress/deformation), and (6)
vibration (e.g., geophysical testing, blast monitoring, seismic activity).
Regardless of the type, selection of monitoring instruments must consider the

instrument range, accuracy, and precision as well as the required calibration
procedures. Geotechnical instruments have a specific measurement range, which must
encompass the values anticipated in the field application. Instruments also have a
precision, which refers to the smallest recordable unit that can be measured. For
example, a 1,000 lb. capacity load cell that records to the nearest 0.1 lb. has a precision
of 0.01% of its full range or full scale. The accuracy of an instrument refers to the ability
to obtain a correct and repeatable measurement of the desire quantity. Instruments
with a large range are not always sufficiently accurate at values near the lower end of
the range. Many instruments require calibration to convert the measured property into
the desired engineering property. While typically provided by the manufacturer, the
calibration should be confirmed and repeated on a regular basis.
Table 2-25 Types of Geotechnical Monitoring Instruments
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Instrument Type Examples Advantages Disadvantages

Low cost, simple and direct
Plumb bob, tape
measurement, readings in Continuous readings are not
Mechanical measure,
engineering units, external power possible, manual recording
micrometer
not required
Relatively rugged, and portable
Piezometers, electrical supply not required, not
Pneumatic – use Difficulty of reading systems,
earth pressure impacted by electrical signals,
air pressure time-consuming manual effort
cells relatively good for long-term
measurements
Hydraulic – use Piezometers,
Similar to pneumatic but with
water or hydraulic earth pressure Similar to pneumatic
higher pressures available
fluid cells
Bonded strain
Electrical – sensing Very stable and reliable, can be Higher cost, require a controlled
gauge,
element bonded to automated and remotely accessed, power supply, signal processing
piezoelectrics,
surface expected to low voltage and portable power is required to obtain data in
vibrating wire
strain supplies are available engineering units
devices
Require a controlled power
Micro-electro- Tiltmeter, Combine microscopic mechanical
supply, signal processing is
mechanical piezometer, parts with electric signals, can be
required to obtain data in
(MEMS) load cell automated
engineering units
Fiber optics –
Strain gauge Can measure strain along the Requires external power,
measurements of
(distributed or entire length of a structure, sensor relatively high cost for readout
strain along an
discrete), cost is low, potential for automation and signal processing, data
embedded or
temperature and dynamic analysis interpretation is required
bonded fiber
2-10.2 Linear Deformation Measurements.
There are many applications where deformation monitoring is either imperative or, at
least beneficial (see Section 2-10.9 below). The deformation could involve vertical
movement from consolidation adjacent to a deep excavation, horizontal and rotational
movements from a landslide, or outward tilt of a retaining wall. Methods for determining
linear deformation are summarized in Table 2-26 and can range from simple to
relatively complex.
Simple methods for deformation monitoring should not be overlooked. Peck (1972)
notes that the human eye is too often overlooked and “can detect most of what we need
to know about subsurface construction.” Measurements by crack pins and tape
measure can be used to monitor observed cracks in soils adjacent to slopes or cracks in
rock. If conditions warrant, distances can be continuously monitored using a
displacement transducer as shown in Figure 2-24. When considering instruments for
monitoring observed cracks in the walls of buildings, a mechanical grid crack gauge can
be monitored to show magnitude and direction of movement over time. This gauge is
simply attached across the crack using epoxy or pins though the mounting holes.
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Table 2-26 Methods of Determining Linear Deformation

Instrument Type Operation Comments
Observe conditions within the subsurface Readily available at no cost, helps develop
Human eyes
and effects on structures observational method, often overlooked
Measure cracking by distance between
Crack pins and tape Simple solution, can be automated using
pins on either side of a crack in soil or
measure electrical displacement sensors
rock
Used to monitor cracks in structures,
Grid crack gauges two-piece plastic gauge attached to Simple solution, requires manual readings
structure on opposite sides of crack
Displacement gauge –
Use to measure displacement over Very accurate, electrical instruments can
e.g., dial gauge or
relatively short time or in a protected be automated, sensitive to environmental
LVDT
environment disturbance, more expensive
(Figure 2-24)
Requires a common benchmark, survey
must be “closed,” precision should be
Use conventional surveying to locate
Field survey established by completion of two surveys,
position and elevation of points
including equipment tear-down, on the
same day
Excellent for construction-induced
Use an AMTS set up in a secure location movements, more repeatable than
Automated total station
to take measurements of targets at conventional survey, near real-time
(AMTS)
selected time interval readings, can be included in online
monitoring sites
Good long-term measurement technique,
Plate with riser pipe is installed on or
Surface settlement accurate and relatively inexpensive, riser
within the ground and surveyed over time
plates or platforms pipes must be protected during
to track settlement, riser extensions can
(Figure 2-25) construction, benchmark must be outside
be used for deep fills
of area impacted by construction
Similar to surface settlement plate with a Higher initial cost than surface settlement
Liquid level settlement
pressure transducer and without risers, plates, construction is much easier without
gauge
changes in pressure from an external risers, potential for leakage, reservoir and
(Figure 2-25)
reservoir are converted to settlement tubing must be protected from freezing
Install a flexible pipe below the
Can use a water filled pipe and a
Liquid level settlement embankment, pull transducer and water
standalone pressure transducer, provides
profiler line through the pipe, measure pressure
distributed measurement of settlement,
(Figure 2-26) at intervals, calculate settlement from
time intensive manual measurements
pressure
Measures relative position of two or more
Borehole extensometer points along the axis of a borehole, Requires a borehole, can monitor
(Figure 2-27) anchor the rod at the base or point of movements at multiple points
measurement
Where needed, more precise deformation measurements can be made using

transducers. When monitoring deformations over relatively short time periods or when
the instrument is protected, a simple dial gauge may be used. For automated readings,
an electrical transducer (i.e., linear variable displacement transformer (LVDT), direct
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current LVDT, or linear potentiometer) provides a precise and potentially highly accurate
alternative.
Conventional field survey equipment can be used to monitor deformations of several

points over large distances and often over long time periods, based on a common
benchmark. Such surveys must be “closed” by shooting the benchmark before and
after the survey. Automated total stations (AMTS) can provide accurate near real-time
monitoring for multiple points and are especially useful for monitoring construction-
induced movements. The accuracy of the AMTS is inversely proportional to the
distance from instrument to prism (or object), but can be used at distances greater than
1,000 feet. As with any surveying option, it is always desired to include a benchmark
point during each reading cycle. The AMTS uses a laser for finding and monitoring the
target so it can be used day or night.
Figure 2-24 Electrical Crack Gauge and Reference Pins (after Dunnicliff 1993)
The magnitude and time-rate of consolidation settlements during construction are often
used to direct certain construction activities (i.e., fill placement rate). Monitoring long-
term settlement is often desired. Surface settlement plates (often referred to as
platforms) are monitored using conventional surveying instruments and provide an
accurate and relatively inexpensive technique. The settlement plate is placed on the
original ground surface before fill placement commences. A figure showing a typical
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set-up is provided in Figure 2-25a. It is necessary to protect the pipe from being
damaged or tilted during construction by either vehicle impact or differential fill
placement around the riser pipe. As with all survey measurements, a non-moving
benchmark is to be shot during each cycle of readings. More advanced systems use
liquid pressure to measure change in height below a fixed reservoir (Figure 2-25b) or
probes to measure the profile of settlement in a buried tube (Figure 2-26).
Figure 2-25 Surface Settlement (a) Plate or (b) Platform (after Dunnicliff 1993)
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Figure 2-26 Liquid Level System to Continuously Profile Settlements

(after Dunnicliff 1993)
Figure 2-27 Borehole Extensometer (after Dunnicliff 1993)
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2-10.3 Angular Displacement Measurements.
Angular displacement measurements are used to measure relative displacements of

slopes or tilting of structures, such as retaining walls. These instruments can be used to
determine the location of planes of sliding and to give warnings when structures
become out-of-vertical. Types of angular displacement instruments are summarized in
Table 2-27.
2-10.4 Pore Pressure and Water Pressure Measurements.
Transducers that are capable of measuring transient water pressures are commonly
relied upon for pore pressure measurements. These transducers are often generically
referenced as piezometers. This is in contrast to the use of the term standpipe
piezometer in Section 2-8 to describe a specific type of well. As described above, the
same basic technology can be used for measuring both static and transient water
pressures, and the transducer types include pneumatic, hydraulic, electrical, and MEMS
devices. Operation and application of these piezometers are summarized in Table
2-28. All of these methods measure positive water pressures in saturated soil.
Table 2-27 Angular Displacement Instruments

Instrument
Operation Comments
Type
Uses accelerometer to determine
inclination with respect to vertical,
MEMS tiltmeters may have range up to 40 arc minutes
Tiltmeter typical operating range of a few
and precision of 1 arc second
degrees, affixed to a structure or used
as integral part of inclinometers
Frequently used to assess movement for landslides, can
Special grooving casing grouted into a
be used for vertical structures and deep excavations,
borehole, inclinometer probe with
Slope initial baseline reading obtained after grout is set, casing
biaxial accelerometer is pulled through
inclinometer should extend into a stable stratum, two passes through
the casing, measures inclination of
(Figure the casing are required for quality data, time-consuming
casing at regular intervals, integration
2-28) manual process, bias correction must be completed in
of tilt provides deformed shape of
data processing, excessive deformation prevents
casing
ongoing use of casing
Same principle as conventional
In-place Can be automated, less time-consuming, much higher
inclinometer except that the instrument
inclinometer equipment cost compared to conventional inclinometer,
has multiple segments with multiple
(Figure MEMS accelerometers can be used that are cheaper
accelerometers, the instrument
2-28) and do not require special casing
remains at same location within casing
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Figure 2-28 Slope Inclinometers – (a) Manual System, (b) Measurement Principle,
and (c) In-Place Inclinometer System (after Dunnicliff 1993)
Hydrodynamic lag time or simply lag time refers to the time required for an instrument to
respond to a change in pressure. Figure 2-31 plots the estimated lag time (in terms of
time for 90% response) for various types of wells and piezometers compared to the
hydraulic conductivity of the soil. Note the short lag time for diaphragm transducers and
the extended time lag for open piezometer wells. As indicated by the figure, open wells
cannot effectively measure hydrodynamic water pressures for transient flow situations.
Table 2-28 Piezometer Types

Piezometer Type Operation Comments
Good choice for long-term monitoring, can be
Device pushed or grouted in-place at
flushed and cycled to increase confidence,
location of interest, air pressure is used
Pneumatic power source not required, insensitive to stray
to measure water pressure on opposite
electrical signals, time-consuming manual
side of a diaphragm
readings
Similar operation to pneumatic except
Hydraulic, a.k.a., water pressure is used instead of air,
Period flushing is required for long-term
twin-tube pressure lines are filled with deaired
monitoring
(Figure 2-29) fluid, pressure is measured with
mechanical or electrical gauges
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Piezometer Type Operation Comments

Very common, can be automated and
Strain gauge, piezoelectric, or vibrating
remotely monitored, very little water flow
wire transducer attached to a diaphragm
required to move the diaphragm resulting in
Electrical and to measure pressure, transducer output
rapid response to transient conditions (i.e.,
MEMS is proportional to pressure, use high-air-
short lag time), requires power source (can be
(Figure 2-30) entry saturated filters to accurately
low voltage), MEMS sensors experience
measure changes in pore pressure
electrical drift in the signal, can recalibrate
during transient conditions
MEMS if sensor is accessible
Figure 2-29 Dual-tube Hydraulic Piezometer in Embankment Dam

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Figure 2-30 Example of Electrical Diaphragm Piezometer Transducer

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Figure 2-31 Estimated Hydrodynamic Lag Time for Various Piezometers and
Wells (after Dunnicliff 1993)
2-10.5 Earth Pressure Measurements.
Although uncommon, measurement of the actual vertical and/or horizontal stress in the
ground is sometimes desired. The concept for making these measurements is easy,
while the interpretation of the results can be challenging. An earth pressure cell (a.k.a.,
total stress cell) measures stress by deformations of a thin diaphragm or transfer of
pressure to a hydraulic cell. Deformations of the diaphragm-type cell can be measured
by a strain gauge, vibrating wire transducer, or MEMS gauge. Earth pressure cells
make an internal measurement compared with most of other transducers that monitor
soil response externally. The presence of the cell alters the stresses in the soil due to
differences between the stiffness of the cell and the soil. Good summaries of the
influence of installation and type of earth pressure cells can be found in Filz and
Brandon (1994) and Sehn (1990). As an alternative to earth pressure cells, tactile
pressure sensors are thin, flexible polymer sheets with many embedded strain sensors
that allow stress to be measured (Paikowsky et al. 2006).
From experience, earth pressure cells should be relatively large (i.e., 9 to 12 inches) in
diameter and thin, resulting in a ratio of thickness to diameter of less than 1:10. The cell
must be in intimate contact with the soil and is generally surrounded (i.e., bedded) using
fine sand to minimize potential for stress concentrations from large particles in the soil.
2-10.6 Load Measurements.
Many applications in geotechnical projects require the measurement of load. A

summary of the common load measuring instruments is provided in Table 2-29.
Table 2-29 Instruments for Measuring Load

Passive system that responds to load by internal
Load is applied to a sealed
pressure increase, relatively robust, capable of
Hydraulic Load hydraulic chamber, increase in
measuring loads >500 tons with about 0.1% accuracy,
Cell pressure is measured and
no external power required for cell with mechanical
converted to load
pressure transducer
An external pressure transducer
Same advantages and capacities as hydraulic load
Calibrated to measures pressure induced by
cells, Osterberg Cell (O Cell) is an example used for
Hydraulic Jack a hydraulic jack, pressure over a
load testing of drilled shafts
known area is converted to load
Load applied to metal gauge, Commonly used in laboratory and field applications,
bonded strain gauges or vibrating wide range of capacities and physical sizes, can be
Electric Load Cell wire transducers measure strain, custom made, robust and reliable, require external
which is converted to load via power supply, signal conditioning and data processing
calibration required to determine load
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Can be bonded strain gauges attached to surface of
Strain gauge attached to surface
structure or pre-attached (sister bars) gauges welded to
Embedded and of metal structural members,
the structural member or reinforcing cage, gauges must
surface-mounted stresses in structure are
be correctly located to measure the desired strain and
strain gauges determined from strain and can be
typically are installed in sets, must be protected during
converted to load via section size
construction and operation
2-10.7 Temperature Measurements.
As described in Dunnicliff (1993), temperature measurements in geotechnical

engineering are typically obtained for one of three specific reasons. The appropriate
instrument for the specific application should be selected and used accordingly.
• Direct Measurement: Temperature measurements may be required for projects,

such as depth of frost penetration, soil temperature beneath industrial furnaces,
temperatures of thermal piles. Thermocouples and resistance temperature
devices (RTDs) are the most common devices for these applications.
• Measurement of Temperatures that Induce Loads: Temperature changes cause
thermal expansion and contraction of materials. The loads in structural members
(e.g., struts for excavation support, tunnel support systems) will be impacted
when subjected to temperature fluctuations. Again, thermocouple or RTDs are
commonly used for these applications.
• Measurement of Temperatures that Influence Transducer Performance:
Transducer temperatures can also have significant impacts on the response of
geotechnical instruments themselves. For example, in closed hydraulic systems,
temperature can change the viscosity of the fluid and can cause expansion
and/or contraction of the fluid and the hoses. These changes will influence the
interpreted response unless appropriate corrections are made. Instrument
manufacturers will report the necessary corrections. Monitoring temperature
changes for transducers requires accurate measurement of small temperature
changes. The thermistors used for this purpose are usually built directly into the
transducer by the manufacturer.
2-10.8 Vibration Measurements.
In some specialty cases, the measurement of vibrations may be a component of a

performance monitoring program. Both steady state vibrations and transient vibrations
may be of interest.
• Steady State Vibrations: Oscillating machinery that can induce vibratory loads to
the foundation and foundation soils. These vibrations may induce cracks
concrete floors, walls, masonry, or finishes. Vibration transducers use
accelerometers (often MEMS based) to assess the magnitude and frequency of
the vibration. These transducers may be included in portable handheld meters
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that are effective at capturing the induced vibrations from the machinery. Modern
smartphones use MEMS accelerometers and can be used to assess vibrations.
• Transient Vibrations: Transient vibrations may be induced by construction
activity, such as equipment traffic, pile driving, dynamic compaction, or blasting;
machinery; or seismic activity. Transient vibration monitoring is an integral part
of many of the geophysical tests discussed in Section 2-4. Transient vibrations
are captured by accelerometers and require high sampling rates in order to
capture vibrations that may occur over a few seconds. Of particular interest for
the transient vibrations are the magnitude, duration, and frequency content.
Capturing and storing vibration data for detailed analysis usually requires dedicated
hardware, specifically portable (or permanent) seismographs. This equipment may
include geophones to measure ground velocity or accelerometers to measure ground
acceleration. The equipment usually has a user-defined threshold limit for the device to
trigger its recording and the ability to capture data a few seconds before the triggering
event. A brief discussion of the equipment to measure and capture ground vibrations is
presents in NCHRP (2018). There are engineering firms that specialize in measuring
transient vibrations.
2-10.9 Field Applications for Instrumentation.
Every instrument used to monitor geotechnical projects should be selected to answer a

particular performance question. This approach opposes the tendency to adopt a
philosophy of “if you can monitor it, you should monitor it,” which is not recommended.
Table 2-30 provides example performance questions that might be appropriate for
various project types. Significant forethought should be given before an instrumentation
plan is developed and the program should be organized around addressing specific
questions. Additional discussion on these topics is provided in Dunnicliff (1993) and
NCHRP (2018).
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Table 2-30 Example Questions for Instrumentation Decisions

Project
Example Project Types Potential Questions
Category
• Deep excavations in urban areas and/or • What is the lateral extent and magnitude of
Braced or
near historical structures ground surface deformations?
Anchored
• Excavation support system that uses • Would an instrumented test anchor load test
Excavations
high strength anchors benefit the final design?
• Strength gain and staged construction • What rate of strength gain is required to not
Embankments
are part of the design impact the construction schedule?
on Soft
• Significant settlement or long-term • What is the confidence of predicted long-
Ground
movements are anticipated term settlement?
• Excavation for rehabilitation is • During rehabilitation, how much movement
Embankment anticipated near the downstream toe is anticipated at specific locations?
Dams • Cutoff structure is part of the • What measurements can be made to
rehabilitation increase confidence of cutoff performance?
• Existing slopes that have to be • What is the anticipated movement or impact
Excavated
steepened of slope steepening?
and Natural
• Seepage appears to be impacting slope • What is the confidence that the design
Slopes
stability accurately accounts for groundwater levels?
• Tunnel will be advanced below • What is the confidence in the ability to
Underground groundwater control water?
Construction • Fractured rock conditions are anticipated • What is the confidence of the variation in
• Seams of weak materials are anticipated rock structure along alignment?
• What is the confidence of pile driving
• Deep foundation system new to the area
acceptance criteria?
Driven Piles is being considered
• What is level of vibration for structures in
• Driven piles in urban environments
close proximity?
• Larger loads than had previously been • What is the confidence in lateral load
used in area are included in the design capacity and location of bending moments?
Drilled Shafts
• Drilled shaft will be located below • What is the impact of excavation below the
groundwater groundwater table?
• Earth retaining structures not previously • What is the estimated settlement of the wall?
Earth
been used in area • What is the estimated tilt of the wall?
Retaining
• Wall height exceeds heights previously • What is the confidence in anchors loads and
Structures
constructed in the region long-term creep?
• Dewatering in urban areas • Will the drawdown be uniform?

Dewatering • Dewatering is considered on the critical • What is the confidence in pumping rate and
path of construction drawdown?
• Uncertainty regarding grout take • What is the confidence in predicted grout

• Grouting in fractured rock and karst take?
Grouting
• Grouting is initiated to minimize potential • Will post-construction verification provide
for settlement of critical structures useful verification data?
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Project
Example Project Types Potential Questions
Category
• Techniques have previously been untried • What is the confidence in long-term
Ground in region performance?
Improvement • Ground improvement is implemented to • What is the confidence in ability to predict
strengthen or stiffen the existing soil settlement and strength gain?
• Litigation is anticipated • How closely do original design assumptions
Liability • Client may be implicated based on match as-constructed conditions?
Control conditions encountered in the field during • Will additional monitoring benefit the client?
a forensic investigation What type?

Topic Reference
NCHRP. (2018). Manual on Subsurface Investigations. National Cooperative
Highway Research Program. Publication No. CRP Project 21-20.
Subsurface Exploration
Transportation Research Board, National Academies of Science Engineering,
and Medicine, Washington, DC.
Cone Penetration Testing (CPT) Design Manual for State Geotechnical
Cone Penetration Test Engineers. Report No. 2018-32, Minnesota Department of Transportation, St.
Paul, MN, 2018.
Geotechnical Site Characterization. Geotechnical Engineering Circular No. 5.
In situ Measurements Publication No. NHI-16-072., Federal Highway Administration, U.S.
Department of Transportation, Washington, D.C., 2016.
Ground Water Manual, Water Resources Technical Publication, United States
Groundwater Measurements
Department of the Interior, Bureau of Reclamation, Washington, D.C., 1995.
Geotechnical Instrumentation for Monitoring Field Performance. by J.
Field Instrumentation
Dunnicliff, John Wiley & Sons, New York, NY, 1993.
2-12 NOTATION.
Symbol Description
D Diameter of the vane - VST
ED Dilatometer modulus - DMT
Ep Pressuremeter modulus - PMT
E' Equivalent modulus
fs Sleeve friction - CPT
G Shear modulus
ID Material index
KD Horizontal stress index - DMT
KO Coefficient of lateral earth pressure at rest
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Symbol Description
ks Modulus of subgrade reaction
qt Normalized tip resistance - CPT
qu Tip resistance - CPT
RQD Rock Quality Designation
St , fv Sensitivity in undrained shear strength from vane shear test
su Undrained shear strength
su , fv Undrained shear strength from vane shear test
sur , fv Remolded undrained shear strength from vane shear test
Tmax Maximum net torque for vane shear test
Tres Residual torque for vane shear test
V0 Initial calculated volume within the uninflated membrane in PMT
µR Vane correction factor
Inflection point assumed to delineate the change from pseudo elastic to plastic response and the
pf
point where creep may be expected in PMT
Pressure at which recompression of disturbed soil in the side of the borehole is complete and
p0
expansion into undisturbed soil starts in PMT, also referred to as liftoff pressure
Yield point during the reloading portion of a PMT unload–reload cycle where recompression
pr
ends and the soil reinitiates plastic shearing
pt Limit pressure in PMT where the curve becomes asymptotic on a pressure versus volume curve
pu Minimum pressure during unloading during the PMT unload–reload cycle
σ h0 Total horizontal stress
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LABORATORY TESTING
3-1 INTRODUCTION.
3-1.1 Scope.
This chapter discusses the common laboratory tests that are used in geotechnical
engineering practice. The chapter has been written assuming that the reader will not
personally be conducting the tests, but will be engaging a commercial laboratory to do
the tests. The discussion considers the types of test that can be conducted for different
engineering parameters and important factors influencing the values obtained.
3-1.2 Evolution of Laboratory Test Procedures.
Geotechnical laboratory testing began in the early part of the last century. The test
apparatuses and procedures were developed by a variety of organizations. Certain
index tests used in geotechnical engineering were originally used in soil science and
agronomy. Many of the compression and strength tests were initially developed by
universities in the United States and Europe. An important early study published in
1946 was “The Use of the Triaxial Test in Engineering Practice” which was the
summary report for a 10-year study sponsored by the Corps of Engineers. In this study,
Professors Arthur Casagrande of Harvard, Don Taylor of MIT, and P. Rutledge of
Northwestern, developed the major categories for triaxial testing of soils which are still
used today.
During the next 40 years, testing procedures and specifications were developed by
organizations involved in constructing dams and highways. The U.S. Bureau of
Reclamation first published the Earth Manual in 1951, and newer editions were released
in subsequent years. The U.S. Army Corps of Engineers developed EM 1110-2-1906
Laboratory Soils Testing, which provided procedures for conducting and presenting the
results of a variety of geotechnical tests. AASHTO has published over thirty editions of
Standard Specifications for Transportation Materials and Methods of Sampling and
Testing, which contains many geotechnical tests. In addition, some state departments
of transportation have developed their own test specifications. The testing procedures
for all of these organizations have coalesced to procedures standardized by ASTM
International.
ASTM was initially named the American Society of Testing Materials in 1902, and the
first specifications focused on tests related to the railway industry. The organization has
expanded to manage the specifications for testing a variety of engineering materials and
consumer products. ASTM has been very active in developing standards for soil and
rock testing, and these standards have been widely adopted in U.S. and international
engineering practice. Committee D18 oversees the standards for soil and rock testing,
and dozens of subcommittees covering a wide array of special areas in laboratory
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testing. The name of the organization was changed to ASTM International to reflect that
the test standards are used internationally as opposed just in the United States. ASTM
International provides standards for hundreds of soil and rock tests and these standards
have been adopted by most of the organizations that once produced their own
standards.
ASTM standards for soil and rock testing usually begin with a “C” or “D” followed by a
three or four digit number. In geotechnical engineering practice, engineers often refer to
specific tests by the ASTM number as opposed to the test name itself. As an example,
engineers will often say “ASTM D698” as opposed to “standard Proctor compaction
test.” The letter and test number are also followed by a dash and a two digit number
reflecting the year that the standard was adopted or last reviewed. For example, ASTM
D698-12 indicates that the standard was last approved in 2012. In this manual, the date
portion of the ASTM standard is omitted since these will change during the useful life of
the manual.
ASTM standards are not static. Each standard is reviewed every five years. During
review, the standard may be reapproved, modified, or withdrawn. Individuals who are
actively engaged in soil and rock testing need to ensure that tests are conducted to the
most recent approved version of the standard.
3-1.3 Laboratory Certification.
The specifications and guidelines for laboratory tests are often quite complex. It takes a
considerable investment of time and money for a laboratory to competently conduct
many geotechnical tests and obtain reliable and repeatable results. It is often difficult
for an engineer to know the competency of a laboratory to conduct high quality tests
without conducting an assessment of the laboratory’s past performance. Two
organizations conduct routine assessments of a laboratory’s ability to conduct
standardized tests to a minimum level of competency.
The Materials Testing Center (MTC) of the U.S. Army Corps of Engineers inspects
laboratories and validates their ability to conduct tests that follow ASTM standards.
They see if the laboratory has a quality manual, certified technicians, functional
equipment, and calibration procedures. Their inspection is done for specific tests, and
they maintain an online register of validated labs and the tests that they are able to
perform.
AASHTO also has a laboratory accreditation program. They perform on-site

assessments of a laboratory’s prowess in conducting tests to AASHTO and ASTM
standards. The laboratory must demonstrate the specific tests on the apparatuses
where they will be performed. They also review the laboratory’s quality management
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program, technician training, and calibration procedures. AASHTO maintains a register

of accredited labs online.
When planning to engage a private testing firm to conduct tests for a project, it is
prudent to examine the validation or accreditation of the proposed laboratory for the
specific tests that will be conducted. Even though a laboratory has been validated, it is
still necessary to carefully review the test results to ensure that the tests were
conducted properly and that the test results are reasonable.
3-2 LABORATORY TESTS ON SOILS.
There are hundreds of ASTM standardized tests used in geotechnical engineering. A

small subset of those, perhaps 40 tests, are routinely used in geotechnical practice. It
the following discussion, the tests will be categorized as: (1) index tests, (2) strength
tests, (3) compression tests, (4) dynamic tests, and (5) permeability tests.
3-2.1 Sample Selection.
Soil samples normally can be categorized as disturbed or “undisturbed.” As explained

in Chapter 2, disturbed samples can be obtained by drive samplers, cuttings generated
by an auger, materials excavated in test pits, etc. “Undisturbed” or intact samples are
those obtained from thin-walled samplers or excavated block samples.
Remolded samples are a form of disturbed sample, and this term is normally reserved
for fine-grained soils. Clays can be remolded by mixing them with a spatula or other
stirring device at a high water content. A remolded sample has lost the structure of the
parent material. A reconstituted sample is also a form of disturbed sample, and that
term is normally reserved for coarse-grained soils. Compacted samples, which can be
formed from either fine-trained or coarse-grained soils, can also be considered
disturbed samples.
The amount of material needed to conduct a given test can vary greatly. Over 100
pounds of soil may be required for compaction tests (ASTM D698 or ASTM D1557), and
only a few ounces of soil may be needed for ring shear tests (ASTM D6467). The
individual ASTM test procedures often state the amount of material needed for a test
series. Table 3-1 provides a summary of the amount of material needed for common
soil tests.
As a general rule, fine-grained soils should not be allowed to dry out prior to testing
unless the test procedures specifically require drying the sample. In particular, oven
drying of a fine-grained soil can cause irreversible changes in mechanical properties.
Correct methods of sample storage for soils is provided in ASTM D3213 (Standard
Practices for Handling, Storing, and Preparing Soft Intact Marine Soil) and ASTM D4220
(Standard Practices for Preserving and Transporting Soil Samples).
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Tests conducted to determine important strength or compressibility parameters for in

situ soil conditions require high-quality intact or undisturbed samples. It has become
common practice to X-ray soil samples while still in the tube (prior to extrusion) to select
the best portions for test assignments. ASTM D4452 (Standard Practice for X-Ray
Radiography of Soil Samples) provides guidance for X-raying soil samples.
Table 3-1 Amount of Soil Needed for Common ASTM Tests

ASTM # Description Sample SizeA Comments
If the sample has material larger
Laboratory Compaction Characteristics 23 kg (Methods A than the No. 4 sieve (Methods A
D698 of Soil Using Standard Effort (12,400 ft- and B) and B) or greater than ¾” (Method
lbf/ft3 (600 kN-m/m3)) 45 kg (Method C) C), much more material may be
needed.
The test can be conducted with less
Specific Gravity of Soil Solids by Water
D854 100 g dry material if the small (250 ml)
Pycnometer
pycnometer is used.
This value is for a sample where
Determining the Amount of Material 100% passes the No. 4 sieve.
D1140 Finer than 75 μm (No. 200) Sieve in 200 g Considerably more material is
Soils by Washing needed for accurate results of
coarser soils.
Laboratory Compaction Characteristics 16 kg dry (Methods Moist field samples of 23 kg for
D1557 of Soil Using Modified Effort (56,000 ft- A and B), 29 kg dry Method A and B, 45 kg for Method
lbf/ft3 (2,700 kN-m/m3)) (Method C) C.
Unconfined Compressive Strength of Around 150 g for test specimen and
D2166 250 - 300 g
Cohesive Soil 50-70 g for trimmings
For maximum particle size:
75 mm: 5 kg required
Depends on particle
Laboratory Determination of Water 37.5 mm: 1 kg
size:
D2216 (Moisture) Content of Soil and Rock by 19 mm: 250 g
20 g – 5 kg
Mass 9.5 mm: 50 g
(Method A)
4.75 mm: 20 g
2 mm: 20 g
One-Dimensional Consolidation Assuming 2.5-in. diameter sample
D2435 Properties of Soils Using Incremental 300 g with height 1 in. Also including
Loading weight of trimmings.
Classification of Soils for Engineering
Exact mass not
D2487 Purposes (Unified Soil Classification
provided in ASTM
System)
For maximum particle size:
Depends on particle
Description and Identification of Soils 38.1 mm: 8 kg
D2488 size:
(Visual-Manual Procedures) 19 mm: 1 kg
110 g – 60 kg
9.5 mm: 220 g
4.75 mm: 110 g
150 g -170 g for specimen loaded
into cell for test and 50-70 g for
trimmings.
for Unconsolidated Undrained Triaxial Minimum diameter is 1.3 in. and
D2850 250 - 300 g
Compression Test on Cohesive Soils
H / D ratio is 2 to 2.5.
Considering a sample with 1.4-in.
diameter and 3-in. height.
Direct Shear Test of Soils Under Minimum specimen diameter for
D3080 250 – 300 g
Consolidated Drained Conditions circular specimens, or width for
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square specimens, is 2 in. and

minimum thickness 0.5 in.
Considering 2.5-in. diameter and
1-in. height.
A Moist sample size unless indicated otherwise
Table 3-1 (cont.) Amount of Soil Needed for Common ASTM Tests
Circular specimen with D = 54 mm
200 g each (2 in.) t D = 0.2 to 0.75.
Splitting Tensile Strength of Intact Rock (10 specimens Considering D = 54 mm, thickness
D3967
Core Specimens required of this = 27 mm, Gs = 2.7, approximate
mass)
weight for each sample is 167 g.
10 specimens required.
Cylindrical specimens with a
minimum diameter of 36 mm [1.4
Determination of the Modulus and
in.]. The height-to-diameter ratio
D3999 Damping Properties of Soils Using the 200 g
shall be between 2 and 2.5.
Cyclic Triaxial Apparatus
Considering D = 1.4 in. and H = 3
in.
Modulus and Damping of Soils by D = 7.1 cm, L = 14.2 cm. Average
D4015 650 - 700 g
Fixed-Base Resonant Column Devices mass required is 609 g.
One- Dimensional Consolidation Around 150 g for specimen loaded
D4186 Properties of Saturated Cohesive Soils 300 g into cell for test and 50-70 g for
Using Controlled- Strain Loading trimmings
Mass of specimen depends on
Maximum particle size:
Maximum Index Density and Unit
D4253 11 - 34 kg 3 in: 34 kg required
Weight of Soils Using a Vibratory Table
1.5 in: 34 kg
0.75 in. or less: 11 kg
Mass of specimen depends on
Maximum particle size:
Minimum Index Density and Unit Weight
D4254 of Soils and Calculation of Relative 11 - 34 kg
38.1 mm: 34 kg
Density
19 mm: 11 kg
9.5 mm or less: 11 kg
Liquid Limit, Plastic Limit, and Plasticity Sample should be passing No 40
D4318 150 - 200 g
Index of Soils sieve.
The size and type of sample needed
Representative
Classification of Peat Samples by is dependent on the tests to be
D4427 samples of the peat
Laboratory Testing performed and the coarseness and
should be used.
moisture content of the peat.
One-Dimensional Swell or Collapse of Considering 2.5-in. diameter and 1-
D4546 250 - 300 g
Soils in. height
Percentage retained not more than
10 % of sieve:
Determination of Water Content of Soil No 10: 100 – 200 g required
D4643 100 - 1000 g
and Rock by Microwave Oven Heating No 4: 300 – 500 g required
¾ in: 500 -1000 g
Rock/gravel samples: 500 g
For vane of diameter 0.5 in. and
Laboratory Miniature Vane Shear Test height 1 in.,
D4648 200 - 250 g
for Saturated Fine- Grained Clayey Soil Minimum Sample required has
diameter 2 in. and height 2 in.
Considering 1.4-in. diameter
Consolidated Undrained Triaxial
D4767 200 – 250 g samples with height around 3 in.
Compression Test for Cohesive Soils
Including weight of trimmings.
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Sample is first air dried or oven

D4829 Expansion Index of Soils 1 kg (dry)
dried.
Considering 2-in. diameter and 2-in.
Measurement of Hydraulic Conductivity
height.
D5084 of Saturated Porous Materials Using a 200 g
Minimum height and diameter are 1
Flexible Wall Permeameter
in.
Table 3-1 (cont.) Amount of Soil Needed for Common ASTM Tests
Cylindrical specimens with min
Load Controlled Cyclic Triaxial Strength Diameter of 2 in. and ratio
D5311 750 - 800 g
of Soil
H / D = 2-2.5.
The height of each specimen shall
be greater than the thickness of the
Performing Laboratory Direct Shear shear (test) zone and sufficient to
Exact mass not
D5607 Strength Tests of Rock Specimens embed the specimen in the holding
provided in ASTM
Under Constant Normal Force rings. Specimens may have any
shape such that the cross-sectional
areas can be determined.
Determination of the Point Load Preferred diameter/width of 50 mm
D5731 Strength Index of Rock and Application 700 - 800 g each (2 in). Average Length = 4 to 5 in.
to Rock Strength Classifications 10 samples required
Soil passing through No. 40 sieve.
Torsional Ring Shear Test to Determine
Including soil for water content.
D6467 Drained Residual Shear Strength of 100 g
Around 40-45 g needed for test
Cohesive Soils
specimen.
Consolidated Undrained Direct Simple Considering 30 mm height and 78
D6528 400 – 450 g
Shear Testing of Fine Grain Soils cm2 cross sectional area.
For maximum particle size (99%
passing):
0.425 mm: 50 g required
2 mm: 50 g
50 g – 70 kg 4.75 mm: 75 g
Particle-Size Distribution (Gradation) of
D6913 (depends on 9.5 mm: 165 g
Soils Using Sieve Analysis
particle size) 19 mm: 1.3 kg
25.4 mm: 3 kg
38.1 mm: 10 kg
50.8 mm: 25 kg
76.2 mm: 70 kg
Consolidated Drained Triaxial Cylindrical sample, including
D7181 250 – 300 g
Compression Test for Soils trimmings
Laboratory Determination of Density Considering 2.8-in. specimens on all
D7263 500 - 600 g
(Unit Weight) of Soil Specimens sides.
Particle- Size Distribution (Gradation) of
Passing No. 10 sieve and retained
D7928 Fine-Grained Soils Using the 50 g (dry)
on No. 200 sieve
Sedimentation (Hydrometer) Analysis
3-2.2 Index Property Tests.
Index properties are used to classify soils and to group soils in major strata. In general,
soils with similar index properties behave similarly, so index properties are often used in
empirical correlations. Index property tests are normally much more inexpensive than
other types of tests, so they are conducted in greater numbers than the more complex
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tests. Index property tests can be conducted on both disturbed and intact samples.
However, it is prudent to save intact samples for tests that specifically require them, and
to use disturbed samples for index property tests. One of the simplest and least
expensive index property tests is ASTM D2216 (Standard Test Methods for Laboratory
Determination of Water (Moisture) Content of Soil and Rock by Mass). This test
requires a relatively small sample, and it is good practice to conduct this test on every
disturbed sample that is taken, particularly of fine-grained soils. Important engineering
information can be gleaned from plots of water content versus depth.
Other index tests are essential to obtain the parameters required for soil classification.
As discussed in Chapter 1, the most common classification scheme used in U.S.
geotechnical engineering practice is ASTM D2487 (Standard Practice for Classification
of Soils for Engineering Purposes (Unified Soil Classification System)). In order to
correctly classify fine-grained and coarse-grained soils, ASTM D4318 (Standard Test
Methods for Liquid Limit, Plastic Limit, and Plasticity Index of Soils) and ASTM D6913
(Standard Test Methods for Particle-Size Distribution (Gradation) of Soils Using Sieve
Analysis) must be conducted. Determination of the Atterberg limits for fine-grained soils
(ASTM D4318) is especially important because they are used in many of the
correlations developed for strength and compressibility properties. The parameters
from the gradation curve (ASTM D6913) are frequently used in correlations with fluid
flow properties of granular soils.
Index tests provide the information necessary to calculate the phase relationship
parameters used to characterize various aspects of densities and saturation conditions
of soils. The phase relationship parameters are import parts of almost all soil tests. A
summary of the phase relationship calculations for soils and rock is given in Table 3-2.
Some index tests are required for specific purposes. As an example, the specific
gravity of soils (ASTM D854 - Standard Test Methods for Specific Gravity of Soil Solids
by Water Pycnometer) is necessary to calculate the void ratio and porosity of a soil or
rock. It is also needed as part of a hydrometer test (ASTM D7928 Standard Test
Method for Particle-Size Distribution (Gradation) of Fine-Grained Soils Using the
Sedimentation (Hydrometer) Analysis) in the data reduction procedure. Some
commercial laboratories may include the cost of conducting certain index tests within
the cost of the more elaborate strength and compressibility tests. For example, a
laboratory may include a specific gravity test (ASTM D854) automatically as part of a
consolidation test (ASTM D2435 Standard Test Methods for One-Dimensional
Consolidation Properties of Soil Using Incremental Loading). The water content test
(ASTM D2216) is normally automatically performed in many other tests. There is a
separate test to measure the unit weight of soils (ASTM D7263 – Standard Test
Methods for Laboratory Determination of Density (Unit Weight) of Soil Specimens), but
determination of the unit weight is a by-product of most of the strength and
compressibility tests on intact soil specimens.
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Table 3-3 includes a list of the common ASTM index tests that are available and the
associated parameters that are determined.
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Table 3-2 Summary of Phase Relationship Calculations
Saturated: Unsaturated:
Supplementary Formulae Relating Measured and
Property Ws , Ww , Gs Ws , Ww , Gs , V Computed Parameters
are known are known
ws V Vv
Vs = Volume of solids Gs γ w
(
V − Va +Vw ) V ⋅ (1− n ) 1+ e e
Vw = Volume of water
Ww
γw
Vv − Va S ⋅ Vv S Va
1+ e
S ⋅ Vs ⋅ e
Volume Components
(1− S ) Va
Va = Volume of air zero (
V − Vs +Vw ) Vv − Vw (1− S ) ⋅ Vv 1+ e
Ww Ws
Vv = Volume of voids
γw
V −
Gs γ w
V − Vs
Vs n
1− n
V e
1+ e
Vs ⋅e
Vv (1+ e )
V = Total volume Vs + Vw Measured Vs + Vw + Va
Vs
1− n
Vs ⋅ (1+ e) e
Vv Vs Ws e
n = porosity 1− 1−
V V Gs V γ w 1− e
Vv V Gs V γ w w Gs n
e = Void ratio −1 −1
Vs Vs Ws S 1− n
Ws = Weight of solids Measured

WT
Gs ⋅ V ⋅ γ w ⋅ (1+ n ) Ww Gs
Weights for
1+ w e S
Specific
Sample
e Ws S
Ww = Weight of water Measured w Ws⋅ S ⋅ γ w ⋅ Vv Gs
WT = Total weight Ws + Ww Ws ⋅ (1+ w)

Ws WT Gs γ w
Ws Gs γ w
γ d = Dry unit weight w Gs
V (1+ w ) 1+
Weights for Sample of Unit
Vs +Vw V 1+ e S
(1+ w ) γ w
Ws +Ww Ws +Ww WT ( Gs + S e ) γ w
γ T = Total (wet) unit weight w 1
+
Vs +Vw V V 1+ e S Gs
Volume
Ws +Ww Ww +Vv γ w Ws 
e 
 ( Gs + e ) γ w (1+ w ) γ w
γ sat = Sat. unit weight +  γ w+ 1
Vs +Vw V V 1+ e  w 1+ e Gs
 
 1− 1 
γ b = Buoyant unit weight or Ws 
1 
  
 Gs + e −1 γ  Gs 
γ sat − γ w − γ   γ w
1+ e  w  1+ e 
w
submerged unit weight V  w+ 1 
 Gs 
Ww WT S e  
w = water content γ 1 
S  w − 
Combined
−1
Relations
Ws Ws Gs
 γ d Gs 
w
Vw Ww w Gs
S = Degree of saturation 1.00 γ w 1 
− 
Vv Vv γ w e 
 γ d Gs 
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Ws S e
Gs = Specific gravity Vs γ w w
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Table 3-3 Index Property Tests and Engineering Parameters Obtained

ASTM # Description Parameters Comments
Laboratory Compaction γ d − max Sometimes it is useful to conduct compaction
D698 Characteristics of Soil Using Standard wopt tests at an effort less than D698. This can be
Effort (12,400 ft-lbf/ft3 (600 kN-m/m3)) done by using 15 to 20 tamps per lift.
Required when void ratios are to be
Specific Gravity of Soil Solids by
D854 Gs calculated or when hydrometer tests are
Water Pycnometer
performed.
Determining the Amount of Material
Useful when classifying a soil in lieu of
D1140 Finer than 75 μm (No. 200) Sieve in % Fines
conducting a complete gradation.
Soils by Washing
Laboratory Compaction The normal maximum mold diameter is 6
D1557
Characteristics of Soil Using Modified γ d − max inches. If larger molds are to be used,
Effort (56,000 ft-lbf/ft3 (2,700 kN- wopt calculations are necessary to guarantee the
m/m3)) correct compactive effort.
Note that lower oven temperatures are used
for organics. Oven drying can cause
Laboratory Determination of Water
irreversible changes in the mechanical
D2216 (Moisture) Content of Soil and Rock
w% behavior of clay soils. Specimens which have
by Mass
been oven dried should not be used for other
tests.
Classification of Soils for Engineering
Classification
D2487 Purposes (Unified Soil Classification Classification of peat (Pt) is in D4427
Symbol
System)
Classification
Description and Identification of Soils Visual classification is used on boring logs
D2488 Symbol
(Visual-Manual Procedures) and most laboratory test reports.
Description
Important when examining erosion potential
Dispersive Characteristics of Clay Soil
D4221 % Dispersion of fine-grained soils for use in levees and
by Double Hydrometer
dams.
Maximum Index Density and Unit γ d − min A vibratory compactor standard for maximum
D4253 Weight of Soils Using a Vibratory emin density may be available. Concerns exist on
Table proper calibration of the vibratory table.
Minimum Index Density and Unit γ d − min
D4254 Weight of Soils and Calculation of
Relative Density emax
PL
Liquid Limit, Plastic Limit, and In international practice, the fall-cone device
D4318 LL
Plasticity Index of Soils is often used for LL and sometimes PL.
PI
Ash content
Classification of Peat Samples by
D4427 Fiber content
Laboratory Testing
Acidity
Determination of Water Content of This normally is used for field compaction
D4643 Soil and Rock by Microwave Oven control tests. D2216 is normally used in
w%
Heating standard laboratory practice.
Identification and Classification of Important when examining erosion potential
Dispersive
D4647 Dispersive Clay Soils by the Pinhole of fine-grained soils for use in levees and
classification
Test dams.
Important when examining the compatibility of
D4972 pH of Soils pH steel and other engineering materials in
contact with soil.
Determining Dispersive Important when examining erosion potential
Dispersive
D6572 Characteristics of Clayey Soils by the of fine-grained soils for use in levees and
classification
Crumb Test dams.
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Table 3-3 (cont.) Index Property Tests and Engineering Parameters Obtained
Cc
Particle-Size Distribution (Gradation)
D6913 Cu
of Soils Using Sieve Analysis
Dxx
γd
Laboratory Determination of Density
D7263 γt
(Unit Weight) of Soil Specimens
w%
Particle- Size Distribution (Gradation)
D7928 of Fine-Grained Soils Using the % > 2µ m
Sedimentation (Hydrometer) Analysis
3-2.3 Compaction Tests.
Two types of compaction tests are available for determining the compaction
characteristics of soils. For soils having greater than 15% fines, impact compaction
tests are the most appropriate. Impact compaction tests provide a compaction curve for
the soil, and the maximum dry density ( γ d − max ) and optimum water content ( wopt ) can be
determined from the curve. The as-compacted density of the soil can be characterized
by relative compaction (RC). Relative compaction is defined as:
γd (3-1)
RC
= ⋅100%
γ d − max
where:
γ d = dry density of the soil to be characterized, and
γ d − max = maximum dry density from the compaction curve for a particular effort.
For soils having less than 15% fines, the density of a soil can be characterized by
relative density ( Dr ). Relative density is defined as:
emax − e γ  γ d − γ d − max 
D
=r ⋅ 100%
= d − max   ⋅ 100% (3-2)
emax − emin γd  γ d − max − γ d − min 
where:
e = void ratio of soil to be characterized,
emax = maximum index void ratio,
emin = minimum index void ratio,
γ d = dry density of soil to be characterized (corresponding to e ),
γ d − max = maximum index dry density (corresponding to emin ), and
γ d − min = minimum index dry density (corresponding to emax ).
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Soils in their loosest states have a relative density equal to 0% and soils in their densest
state have a relative density equal to 100%. For these soils, tests must be conducted to
determine the maximum and minimum index void ratios and index densities. These
tests are not “compaction tests” in the strictest sense of the term, but are index values
corresponding to the loosest and densest states of the soil. The maximum dry density
used in relative density calculations is not necessarily the same as that obtained in
ASTM D1557 that is described below.
3-2.3.1 Impact Compaction Tests.
ASTM D698 (Standard Test Methods for Laboratory Compaction Characteristics of Soil
Using Standard Effort (12,400 ft-lbf/ft3)) and ASTM D1557 (Standard Test Methods for
Laboratory Compaction Characteristics of Soil Using Modified Effort (56,000 ft-lbf/ft3)).
These tests differ by the amount of effort that is applied to compact the soil. The ASTM
standards provide test procedures for compaction molds having diameters of 4 inches
and 6 inches. Following the specifications, both of these impact compaction tests
should be limited to soils having no more than 30% retained on the 3/4-inch sieve,
although corrections exist for materials having as much as 70% retained on the 3/4-inch
sieve. However, if impact compaction test results are required for larger grain-size
materials, the mold diameters and hammer weights should be scaled up to keep the
compactive effort the same.
3-2.3.2 Index Density Determination.
For soils having less than 15% fines, there are specific tests that can be conducted to
determine the maximum and minimum index densities so that the soil can be
characterized in terms of relative density. ASTM D4254 (Standard Test Methods for
Minimum Index Density and Unit Weight of Soils and Calculation of Relative Density)
provides three methods to determine emax and γ d − min depending on the grain size of the
soil tested. Methods are available for soils have a maximum grain size of 3 inches or
less. ASTM D4253 (Test Methods for Maximum Index Density and Unit Weight of Soil
Using a Vibratory Table) is used to determine emin and γ d − max . 3 The apparatus used to
perform this test is expensive and many commercial laboratories are not able to perform
this test.
3-2.4 Strength Tests.
There are a variety of strength tests that have ASTM specifications for both drained
(effective stress) strength parameters and undrained (total stress) strength parameters.
The most common drained shear strength parameters are the effective stress cohesion
( c ' ) and the effective stress friction angle ( φ ' ). Total stress shear strength parameters
3Some engineers use the dry density obtained from ASTM D4253 for calculating relative compaction for
a soil deposit in the same manner as done using the maximum dry density from impact compaction tests.
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are su (undrained shear strength for a φ = 0 envelope) for saturated soils, and total
stress cohesion ( c ) and total stress friction angle ( φ ) for partially saturated soils.
The shear strength parameters listed above are for linear failure envelopes. It also is
possible to determine parameters for non-linear failure envelopes for many of the
strength tests.
Shear strength parameters are often needed for undisturbed or intact specimens,
compacted specimens, remolded specimens, and reconstituted specimens. The
different strength tests will be assessed based on their ability to accommodate these
different types of test specimens. Table 3-4 lists the ASTM strength tests and the
parameters obtained.
Table 3-4 Laboratory Strength Tests with ASTM Standards

Unconfined Compressive Best used as an index test as opposed to obtaining
D2166 qu
Strength of Cohesive Soil shear strength for design.
su
Field Vane Shear Test in St Perhaps the best all-around tests for undrained
D2573
Saturated Fine-Grained Soils strengths of soft, saturated clays.
Strength
anisotropy
Unconsolidated Undrained c −φ Best results are obtained when all samples come
D2850 Triaxial Compression Test on from the same depth, which can be obtained using
su
Cohesive Soils 5-in. diameter sampling tubes.
Direct Shear Test of Soils
Good test for measuring fully softened shear
D3080 Under Consolidated Drained c '−φ '
strength.
Conditions
Splitting Tensile Strength of
D3967
Intact Rock Core Specimens
σt Indirect measurement of tensile strength.
su
Laboratory Miniature Vane
St Very good test for soft clay samples that are not
D4648 Shear Test for Saturated Fine-
trimmable for UU triaxial tests.
Grained Clayey Soil Strength
anisotropy
Consolidated Undrained c '−φ ' Good for effective stress strength parameters of
D4767 Triaxial Compression Test for sands and clays. Use caution when using total
Cohesive Soils su for σ '3con stress strength parameters for stability analyses.
Laboratory Direct Shear c '−φ '
Strength Tests of Rock joint
D5607
Specimens Under Constant roughness
Normal Force coefficient
Determination of the Point
Load Strength Index of Rock Is
D5731 Used for rock classification and other applications.
and Application to Rock I s (50)
Strength Classifications
Torsional Ring Shear Test to
Best test for residual shear strength for clays.
Determine Drained Residual
D6467 φ 'r Staged tests can save a lot of time. Very small test
Shear Strength of Cohesive
specimen.
Soils
Consolidated Undrained Direct The undrained strength measured is a conservative
D6528 Simple Shear Testing of Fine
su approximation of the maximum shear stress at
Grain Soils G failure.
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Consolidated Drained Triaxial Tests on CL and CH clays may take a very long time
D7181 c '−φ '
Compression Test for Soils to complete. D3080 may be a better choice.
3-2.4.1 Drained or Effective Stress Strength Tests.
There are four recommended tests listed in the ASTM specifications for measuring
drained or effective stress strength parameters: (1) Direct Shear Test, (2) Consolidated
Drained Triaxial Test, (3) Consolidated Undrained Triaxial Test, and (4) Ring Shear
Test.
3-2.4.1.1 Direct Shear Test (ASTM D3080).
The direct shear test (ASTM D3080 – Standard Test Method for Direct Shear Test of
Soils Under Consolidated Drained Conditions) is one of the oldest and most common
strength tests. Figure 3-1 shows the basic elements of a direct shear tests and the data
collected. In U.S. engineering practice, the most common specimen cross-sections are
2 inch × 2 inch and 4 inch × 4 inch square specimens and 2.5-inch diameter circular
specimens. A few commercial laboratories have direct shear apparatuses that can
accommodate 12 inch × 12 inch specimens. The 2.5-inch diameter cylindrical shear
box is very popular since intact specimens can be easily trimmed from the common 3-
inch diameter Shelby tubes. It is also easier to compact test specimens in a cylindrical
specimen container than a square specimen container. The direct shear apparatus can
use intact, compacted, or remolded test specimens.
The direct shear test is most often used for clay soils and for some sandy soils. ASTM
D3080 requires that the maximum grain size of the test specimen be no greater than
1/10 of the shear box width and no greater than 1/6 of the specimen height. Based on
the common sizes of the shear box, it is not appropriate to test materials larger than
medium sands.
Direct shear tests are relatively easy to conduct. A consolidation stress is first applied
to the test specimen. After all excess pore water pressures are dissipated, the sample
is sheared at a constant displacement rate very slowly to ensure drained conditions are
maintained. Direct shear tests do not produce stress-strain results, but rather stress-
displacement results. Moduli cannot be determined from direct shear tests.
Since the failure plane in a direct shear test is constrained to the horizontal plane, then
the shear strength parameters measured might be lower than those measured in other
tests that have an inclined shear plane for soils that exhibit horizontal layering (Duncan
et al. 2014). Soils that are deposited in water; such as lacustrine, alluvial, and marine
soils; may exhibit this inherent anisotropy whereby the shear strength is a function of
the orientation of the failure plane. This issue of the failure plane orientation is
insignificant for remolded and compacted soils. Direct shear tests also may suffer from
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progressive failure, whereby the shear strength is not fully mobilized on the entire failure
plane at the same instant. This can also result in lower peak shear strength parameters
for materials that exhibit brittle stress-displacement curves, such as heavily
overconsolidated clay.
Figure 3-1 Basic Elements for a Consolidated Drained Direct Shear Test Along
with Example Data Collected for One Test Specimen
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Direct shear tests are especially well-suited for testing remolded clays. It is easy to
form test specimens in the shear box and the device allows the application of small
consolidation stresses. It is the best test available for measuring the fully softened
shear strength, which is the peak shear strength of remolded, normally consolidated
clays.
3-2.4.1.2 Consolidated Drained (CD) Triaxial Test (ASTM D7181).
The consolidated drained (CD) triaxial test (ASTM D7181 – Standard Test Method for
Consolidated Drained Triaxial Testing of Soil) is one of the oldest types of triaxial test.
The procedure was defined in the Army Corps of Engineers 1930s study. This type of
triaxial tests was also called an S triaxial, with S denoting slow, since this test has the
slowest shear phase of the major categories of triaxial tests.
Common specimen sizes in U.S. practice are 1.4-inch diameter for fine-grained soils
and 2.8-inch diameter for sandy soils or for test specimens directly extruded from
Shelby tubes without trimming 4. Many laboratories can test 2-inch diameter and 4-inch
diameter specimens as well. The larger 6-inch diameter and 12-inch diameter
apparatuses are rarer in commercial laboratories. ASTM D7181 requires that the
maximum grain size of the soil tested should be 1/6 of the test specimen diameter, so
the diameter of the test specimen should be selected based on the grain-size
distribution of the soil.
Figure 3-2 shows the basic elements of a manual CD triaxial test. The test specimen is
consolidated by applying a cell pressure. Most tests specimens are consolidated to
isotropic stress conditions, but anisotropic consolidation is also possible. The sample is
back-pressure saturated by using an elevated pore pressure (back pressure), and the
final consolidation stress is the cell pressure minus the back pressure. The sample is
sheared very slowly, so that excess pore pressures are not developed, by increasing
the vertical stress at a constant displacement rate. The volume change of the test
specimen is measured by recording the level of the burette as the load is applied.
These tests require considerable skill to perform correctly.
CD triaxial tests are more applicable to sandy soils because a high permeability allows
the test specimen to be sheared in a reasonable time. If clayey soils are tested, the
sample must be sheared very slowly, and the test may take weeks to months to
complete. Some laboratories will not agree to conduct these tests on clay soils.
Triaxial tests can be used to test intact, compacted, and remolded test specimens.
There is some difficulty in testing very soft remolded test specimens since the soil
4Although ASTM D7181 allows directly extruded samples to be tested, it is better practice to trim
samples to a smaller diameter to reduce the disturbance caused by sampling.
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needs to have sufficient strength to allow trimming and mounting in the triaxial cell.
Remolded soils often need to be consolidated outside of the triaxial cell to allow a
strength gain prior to trimming. This procedure can limit the consolidation stress that
can be applied to the test specimen if normally consolidated conditions are desired.
Figure 3-2 Basic Elements of a Consolidated Drained Triaxial Test Along with
Example Data Collected for One Test Specimen
3-2.4.1.3 Consolidated Undrained (CU) Triaxial Test (ASTM D4767).
The consolidated undrained (CU) triaxial test (ASTM D4767 – Standard Test Method for
Consolidated Undrained Triaxial Compression Test for Cohesive Soil) is another of the
three major types of triaxial tests defined by the U.S. Army Corps of Engineers’ 1946
study. This test was also called an R test if pore pressures are not measured and an
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R if pore pressures are measured. 5 The R test has become obsolete and it is difficult
to justify a case where this test should be conducted. Figure 3-3 shows the basic
elements of a CU triaxial test and example data.
Figure 3-3 Basic Elements of a Consolidated Undrained Triaxial Test Along with
Example Data Collected for One Test Specimen
The CU triaxial test is very similar to the CD triaxial test, and most of the information
provided for the CD test is also true for the CU test. The test procedures are identical
until it is time to shear the test specimen. In the CD test, the drainage valve is open
during shear, and in the CU test, the drainage valve is closed during shear. Volume
change is measured during a CD test, and pore pressure is measured during a CU test.
In a CD test, the strain rate ( ε ) is very slow because the pore pressured generated
during shear must dissipate throughout the test specimen. In the CU tests, the strain
rate is considerably greater since the goal is to allow equalization of pore pressures as
5 It is not clear why the letter R is used for this test. It may be that the other tests are referred to Q and S
tests, and R is the letter in the alphabet that falls between these. In the past, engineers would refer to the
QRS triaxial test types.
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opposed to dissipation. In other words, CU test specimens can be sheared much faster
that CD tests test specimens, and this is a big advantage in clay soils.
Pore pressures are measured during shear during CU triaxial tests, and this allows the
effective stresses in the sample to be determined. Since effective stress values are
available, drained or effective stress strength parameters ( c ' and φ ' ) can be
determined. However, there is an important difference in the volume change of the test
specimen in CU and CD tests and this can influence the shear strength parameters
determined. In the CD test, volume change is allowed, and the void ratio of the test
specimen can change during shear. In the CU test, there is not volume change during
shear, so the test specimen has the same void ratio at failure as it did prior to shear
(after consolidation). The void ratios of CU and CD test specimens can be different at
failure, even though they may have started at the same consolidation pressure and void
ratio, and this may cause differences in the effective stress shear strength parameters.
In engineering practice, this difference is usually neglected, and results from CU and
CD tests are often used interchangeably.
3-2.4.1.4 Ring Shear Test (ASTM D6467).
The ring shear test (ASTM D6467 – Standard Test Method for Torsional Ring Shear
Test to Determine Drained Residual Shear Strength of Soil) is similar to a direct shear
test in that it is a consolidated drained test with shearing occurring on a horizontal
plane. A ring shear test should be only used to determine the residual shear strength,
which is the shear strength of a soil at very high strains or displacements. Residual
shear strength is used in geotechnical designs for situations where a failure has already
occurred and considerable movement has taken place.
The ring shear apparatus uses an annular test specimen. The specimen size used the
most in the U.S. has an inside diameter of 2.8 inches, and outside diameter of 4 inches,
and a thickness of 0.2 inches. The basic elements of a ring shear tests are shown in
Figure 3-4. The specimen tested is markedly smaller than the other strength tests
discussed. The specimen volume is equivalent to about 4 teaspoons. Only remolded
clay samples are tested in common ring shear apparatuses, and these are placed in the
apparatus in the form of a paste.
The ring shear apparatus allows staged tests to be conducted 6. A single test specimen
can be reconsolidated and sheared multiple times. The normal test procedure is to
place a specimen in the apparatus and consolidate it to the lowest vertical stress. The
specimen is then sheared until the residual strength is achieved. Next, a higher
6 Although some older testing manuals provide instructions for conducting staged CU and CD triaxial
tests, the results of these tests are very unreliable. These types of tests should not be specified and the
data from these tests should not be used.
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consolidation pressure is applied, and the sample is sheared again. Normally, three
different normal stresses can be used for one test specimens.
It is possible to measure residual shear strength properties using a direct shear

apparatus if the direction of shear is reversed so that enough shear displacement can
be accumulated to obtain residual conditions. Although this test is reported in
geotechnical literature, there is not an ASTM procedure for conducting repeated direct
shear tests. The ring shear tests is a better test for measuring the residual friction angle
since the direction of shear does not need to be reversed during the test, and the area
of the shear surface remains constant.
Figure 3-4 Basic Elements of a Ring Shear Test Along with Sample Data
3-2.4.2 Undrained or Total Stress Strength Tests.
There are five laboratory tests with ASTM standards for measuring the undrained shear
strength of soils. Since undrained strengths are normally used for fine-grained soils,
owing to their low permeability, these tests mainly address clayey and silty soils. The
value of undrained shear strength can vary considerably from test to test. The two
fundamental undrained strength envelopes that are used for fine-grained soils are
shown in Figure 3-5.
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For saturated soils undergoing undrained loading, the shear strength is not a function of
the normal stress on the failure plane, thus φ = 0 . For partially saturated soils
undergoing undrained loading, the shear strength is a function of the normal stress on
the failure plane, thus φ > 0 . The envelope for the partially saturated soil usually
deviates from a straight line, but a linear interpretation over the appropriate range of
normal stress if often used in engineering analysis. 7
The undrained strength of a soil can depend on many different factors. Primary factors
include the major effective consolidation stress prior to undrained loading and the
overconsolidation ratio ( OCR ). There are many secondary factors, including the
orientation of the failure plane, the rate of loading, the amount of sample disturbance,
the system of stresses imposed by the field loading condition or the laboratory test, etc.
For the envelopes shown in the figure, only one of the ASTM tests (D2850) provides
multiple points to define the envelopes. The other ASTM tests only provide one point to
define the shear strength envelope.
Figure 3-5 Undrained Shear Strength Envelopes for Saturated and Partially
Saturated Soils
For layers of saturated fine-grained soils, the goal of an undrained strength testing
program is often to determine the variation of undrained strength with depth, such as
shown in Figure 3-6. For these cases, enough in situ or laboratory tests should be
conducted so that the undrained shear strength values, shown as the squares, are
determined in order that the variation with strength with depth can be estimated, shown
as the dashed line.
7A saturated soil is often called a “ φ = 0 soil” and a partially saturated soil is called a “ c − φ soil” by
geotechnical engineers.
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3-2.4.2.1 Unconfined Compression Test (ASTM D2116).
Unconfined compression tests, UCTs, (ASTM D2166) are one of the oldest and
simplest strength tests. The quality and care in conducting the test can vary greatly
from specimens being tested using portable load frames in the field to testing the
specimen in a membrane in a temperature-controlled laboratory. The basic premise of
the test is that if the soil is saturated, the undrained shear strength ( su ) should be the
same if the soil is tested with zero confining pressure ( σ 3 ) as it would be with an
elevated confining pressure. This assumption might be valid for high quality test
specimens carefully tested, but actual test results can deviate from this assumption
considerably. In general, UCTs provide lower strengths than would be determined from
higher quality tests.
Figure 3-6 Example Distribution of Undrained Strength Versus Depth

Relationship for a Hypothetical Saturated Clay
In practice, UCTs are also conducted on partially saturated soils. These tests are
difficult to correctly interpret since the results would represent one shear strength value
on a c − φ envelope, and one point cannot define the envelope. Shortcomings of the
UCT have been recognized for over 70 years, and this test should be considered as an
index test as opposed to a viable method to measure reliable shear strengths.
3-2.4.2.2 Unconsolidated Undrained (UU) Triaxial Test (ASTM D2850).
The unconsolidated undrained triaxial test (ASTM 2850 – Standard Test Method for
Unconsolidated-Undrained Triaxial Compression Tests on Cohesive Soil) has been the
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most popular test for measuring undrained shear strength in U.S. geotechnical
engineering practice. The basic procedure for this test was outlined in the Corps of
Engineers Triaxial Test report in 1946. This test is often called a Q triaxial, with Q
standing for “quick” since this is the fastest triaxial test. The test specimen is sheared at
a strain rate of 1% axial strain per minute, so the shearing phase of the test only lasts
about 20 minutes. 8
UU tests on saturated fine-grained soils are normally conducted using three test
specimens to define an envelope. All three test specimens should come from the same
depth, with all three having the same in situ consolidation pressure and OCR . All three
specimens should have the same shear strength, which would verify the φ = 0 failure
envelope. UU tests on saturated soil should always be interpreted with a φ = 0 failure
envelope, regardless of any slope implied by the tests.
A schematic of the basic test apparatus is shown in Figure 3-7. The sample is sealed in
thin rubber membranes, but there are not any porous stones or drainage lines. Special
triaxial cells are commercially available just for UU tests. UU tests are most often
conducted on 1.4-inch and 2.0-inch diameter trimmed specimens or 2.8-inch directly
extruded specimens.
One deficiency of the UU test is that it is challenging to test very soft soils. For soils that
have an undrained shear strength less than about 250 psf, it can be very difficult to trim
a test specimen, to mount the specimen in a triaxial cell, and to place a membrane over
the specimen. It is very easy to disturb the specimen, and that tends to lower the shear
strength even more. For very soft materials, it is best to use the laboratory miniature
vane shear test (discussed below) or a fall cone test.
8ASTM D2850 suggests using a strain rate of 0.3% for brittle soils, but most laboratories use 1% per
minute for all soils.
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Figure 3-7 Basic elements of a UU Test Apparatus with Sample Data for a Single
Test
UU tests can also be conducted on partially saturated soils, and have often been used
to determine the shear strength parameters for compacted clays. UU tests are the only
viable test to determine the values of c and φ for partially saturated soils for use in end-
of-construction analyses. Special compaction equipment is available to form triaxial test
specimens of compacted soils. 9
There are critics of the UU test, and many of the criticisms are valid (Ladd 1991).
However, strengths resulting from UU tests have been validated by back analysis of
failed slopes and found to be representative, and it remains a very popular test in
engineering practice.
3-2.4.2.3 Consolidated Undrained (CU) Triaxial Test (ASTM D4767).
Since the CU triaxial test is sheared undrained, it is possible to obtain undrained or total
stress strength parameters. However, this test has been misused for undrained
strength determination in the past. It is not possible to determine viable values of c and
9There are test apparatuses available for conducting triaxial tests on partially saturated soils, but few
commercial labs are equipped to perform these tests.
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φ from this test, yet these values are commonly reported by commercial laboratories.
The correct way to use the CU test for undrained analyses is to associate the value of
su (half the deviator stress at failure) with the isotropic consolidation stress. The CU
triaxial test provides undrained shear strength values that are normally too high to be
used in most analyses. Details of the use of the CU test for undrained strength
determination is presented by Duncan and Wong (1983).
3-2.4.2.4 Consolidated Undrained Direct Simple Shear Test (ASTM D6528).
The direct simple shear (DSS) test (ASTM D6528 – Standard Test Method for
Consolidated Undrained Direct Simple Shear Testing of Fine Grain Soils) was
developed in the 1950s. Commercially-available apparatuses test a cylindrical
specimen that is nominally 2.5 inches in diameter and up to 1 inch tall. A schematic of
the basic elements of the DSS test is shown in Figure 3-8. The specimen is confined
with a wire-reinforced membrane or a set of thin, stacked rings, often coated with
Teflon, located outside of an unreinforced latex membrane. The intent of the confining
rings is to prevent lateral strain from occurring when the vertical consolidation stress is
applied, and to allow the test specimen to deform in the manner of pure shear. After the
test specimen is consolidated, it is sheared undrained (technically at constant volume)
by translation of the top platen relative to the bottom platen. The undrained shear
strength ( su ) is assumed to be the maximum value of shear stress applied to the
horizontal plane of the test specimen.
The DSS test was not a common test in geotechnical engineering for fifty years after its
development. Few commercial laboratories were able to conduct the test. There are
valid criticisms of the DSS test (Saada and Townsend 1981). However, the popularity
of this test has increased in recent years, and many more laboratories are able to run
the test. The DSS apparatus provides an undrained shear strength that is comparable
to that obtained with the field vane shear test (for the same vertical consolidation stress
and OCR ) and is appropriate for many engineering design cases. This test can only be
conducted on saturated soils, and the ASTM standard only addresses fine-grained soils.
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Figure 3-8 Basic Elements of the Direct Simple Shear Test (ASTM D6528)
3-2.4.2.5 Laboratory Miniature Vane Shear Apparatus (ASTM D4648).
The laboratory miniature vane shear apparatus (ASTM D4648 – Standard Test Methods
for Laboratory Miniature Vane Shear Test for Saturated Fine-Grained Clayey Soil) is a
scaled-down version of the field vane shear test. The vane sizes range from 0.5 inch ×
0.5 inch to 1.0 inch × 1.0 inch. Vanes having a smaller diameter than the length can
also be obtained to aid in determining anisotropic strengths. A photograph of a vane
shear apparatus is shown in Figure 3-9. The vane is inserted into an intact or remolded
test specimen and rotated at a constant rate while the torque is measured. The vane
can be rotated by a hand crank or an electric drive unit. The torque can be measured
by calibrated springs or by electronic load cells. Most legacy data for the miniature
vane apparatus has been collected using the calibrated springs, and this is the
preferred method.
Lab vane shear tests are not very common in geotechnical engineering practice, but this
type of test can be especially useful in very soft clays. If a clay sample is too soft to be
trimmed for a UU triaxial test, then a laboratory miniature vane test is a viable
alternative.
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Figure 3-9 Laboratory Miniature Vane Shear Apparatus
3-2.4.2.6 Other Strength Tests.
There are other tests available to measure the laboratory undrained shear strength of
soils that do not have specific ASTM standards available, or they are variations of the
conventional ASTM tests.
There are two variations of the CU triaxial test that have limited use in geotechnical
practice. The conventional CU triaxial test is normally conducted on test specimens that
have been isotropically consolidated (vertical stress = horizontal stress during
consolidation). These are often referred to as ICU triaxial tests. It is possible to
anisotropically consolidate test specimens where the vertical stress is different than
(usually greater) the horizontal stress. These are called ACU triaxial tests. The ASTM
specifications address the basic components of ACU tests. ACU tests produce
essentially the same effective stress strength parameters as ICU tests. ACU tests
normally provide different undrained shear strengths than ICU tests, so the main
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usefulness of the test would be for projects where special undrained strengths are
required.
A separate type of ACU test can be conducted where the ratio of the effective stresses
during consolidation (minor effective stress/major effective stress) is equal to the at-rest
earth pressure coefficient ( K 0 ). These tests are sometimes called CK0U triaxial
compression tests. For these tests, the exact stresses are not specified, but determined
to be the stresses necessary for no lateral strain to occur during consolidation. It is
difficult to conduct these tests on manual triaxial apparatuses, but they can be easily
conducted on fully-automatic apparatuses. These tests also provide essentially the
same effective stress shear strength parameters as conventional ICU triaxial tests.
Their main utility would be when undrained strengths are needed for special projects.
Another special type of triaxial test that is occasionally used in engineering practice is
the stress path triaxial test. While most triaxial tests involve loading a test specimen
axially while the cell pressure is constant, stress path tests vary both the vertical stress
and the horizontal stress simultaneously to follow prescribed loading paths. The loading
path is often selected to match field loading conditions. In some cases, the intent is to
measure the strains obtained in the test specimen after the loading path has been
applied. In other cases, the intent is to measure the strength of the test specimen for a
specified system of stress changes. Fully-automated triaxial test apparatus are
normally required to conduct these types of tests.
An alternative to the miniature laboratory vane shear test (ASTM D4648) for measuring
the shear strength of very soft clay is the fall cone test. Although there currently is not
an ASTM standard for this test, it is very popular in Europe, and there are standards in
Norway, Germany, the U.K., and other countries. A photograph for the Norwegian
apparatus is shown in Figure 3-10. This test involves dropping a weight cone onto the
surface on an intact or remolded soil specimen and measuring the penetration. Cones
are available with different weights for different penetration depths for soils having
various consistencies. This test is also used to determine the liquid limit of soils in
international geotechnical engineering practice.
3-2.5 Dynamic Tests.
Geotechnical earthquake engineering is a specialty area within geotechnical

engineering. The methods of analysis used to predict the performance of structures
during earthquakes can often be quite complex, and the tests to measure soil properties
for use in these analyses can likewise be complex. Only a few geotechnical
laboratories have the equipment to conduct these tests. Table 3-5 lists the common
dynamic tests for soils.
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Figure 3-10 Fall Cone Apparatus
Table 3-5 Dynamic Tests for Soils

E
Determination of the Modulus and D Can be used for secant modulus and damping
D3999 Damping Properties of Soils Using coefficients.
the Cyclic Triaxial Apparatus
ε DA Tests can be stress or strain controlled.
ε SA
Modulus and Damping of Soils by D

Some apparatuses can allow anisotropic
D4015 Fixed-Base Resonant Column G consolidation and large torsional strains.
Devices γ
Multiple tests should be conducted to
Load Controlled Cyclic Triaxial multiple
D5311 determine the number of cycles to failure for
Strength of Soil plots
different cyclic stress ratios.
Consolidated Undrained Cyclic Multiple tests should be conducted to

G
Direct Simple Shear Test under determine the number of cycles to failure for
D8296 γ DA different cyclic stress ratios.
Constant Volume with Load Control
γ SA Tests can be stress or displacement controlled.
or Displacement Control
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3-2.5.1 Cyclic Triaxial Test (ASTM D5311).
The cyclic triaxial test (ASTM D5311 – Standard Test Method for Load Controlled Cyclic
Strength of Soil) is the most common of the dynamic tests, and it is one of the oldest,
first being run since the 1960s or before. The cyclic triaxial test is a consolidated
undrained (CU) test, and the initial portion of the test is the same as the static CU test
(ASTM D4767). The back-pressure saturation and consolidation phases are essentially
the same. The difference is in the manner of loading. Cyclic triaxial tests are normally
loaded with a sinusoidal loading function, with the maximum stress difference specified
by a cyclic stress ratio ( CSR ). The CSR is one-half of the applied deviator stress
divided by the isotropic consolidation stress. An example of the loading function is
shown in Figure 3-11. The loading frequency is ideally 1 Hz, but slower frequencies are
often used owing to the difficulty of many commercial apparatuses in maintaining a
constant 1 Hz throughout the test. The ASTM standard allows frequencies as slow as
0.1 Hz. Tests are normally conducted at three or four different values of CSR to
determine the number of cycles until failure. A plot of the applied CSR versus the
number of cycles until failure is used to define the cyclic strength of the soil.
Failure in a cyclic triaxial test can be defined in several different ways. The
conventional definition of cyclic tests conducted on clean sands was when a 100% pore
pressure ratio was achieved (pore pressure = confining stress). Soils with a significant
amount of fines may not achieve failure by this definition, so it has become common to
define failure based on axial strain. Strain values of ±2.5%, ±5% and ±10% axial strain
have been used to define failure.
3-2.5.1.1 Cyclic Direct Simple Shear Test.
The cyclic direct simple shear test (CYCDSS) is popular in geotechnical earthquake
engineering, and an ASTM standard is available (D8296). The test specimen can be
loaded by either a cyclic stress or a prescribed displacement. These tests are normally
performed using constant volume conditions where the height of the test specimen is
not allowed to change during the test. A sinusoidal loading function is applied to the
test specimen as a horizontal force to the top or bottom platen for a load-controlled test.
A loading frequency of 1 Hz is desired, but some tests are conducted much slower
frequencies, particularly for fine-grained soils.
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Figure 3-11 Loading Function and Stresses Applied for a Cycle of Loading in a
Cyclic Triaxial Test for a Cyclic Stress Ratio of 0.2
CYCDSS tests are also specified in terms of cyclic stress ratio. The cyclic stress ratio is
defined as:
τ cyc
CSR = (3-3)
σ 'v
where:
τ cyc = applied peak cyclic shear stress, and
σ 'v = vertical effective consolidation stress.
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Failure is defined as a pore pressure ratio of 100% or a limiting cyclic shear strain. As
with the cyclic triaxial test, the cyclic strength is normally represented as a plot of the
applied cyclic stress ratio versus the logarithm of the number of cycles until failure.
3-2.5.1.2 Resonant Column Test (ASTM D4015).
The resonant column test (ASTM D4015 – Standard Test Methods for Modulus and
Damping of Soils by Fixed-Base Resonant Column Devices) is a dynamic test that
provides values of shear modulus for low shear strain amplitudes. The test is
conducted in a modified triaxial cell and can be conducted on intact and remolded test
specimens. A cylindrical test specimen is loaded by applying a cyclic torque to the top
of the specimen while the resulting angular displacement is measured. The cyclic load
normally follows a sinusoidal function, and the frequency of the load is varied until the
resonant frequency of the test specimen is determined. This is a complex test and the
results are used in specialized earthquake engineering analyses.
3-2.5.1.3 Cyclic Triaxial Test for Modulus and Damping (ASTM D3999).
A different version of the cyclic triaxial test can be used for determining the secant
Young’s Modulus and Damping Coefficients (ASTM D3999 - Standard Test Methods for
the Determination of the Modulus and Damping Properties of Soils Using the Cyclic
Triaxial Apparatus). This test can be conducted on intact or reconstituted saturated and
partially saturated test specimens. Both fine-grained and coarse-grained soils can be
tested. The main purpose of this test is to determine the dynamic properties for strains
ranging from about 0.01% to 0.5%. The test specimen may be back-pressure
saturated, or it may be tested in a partially saturated condition.
3-2.6 Compressibility Tests.
There are four ASTM tests used to measure the volume change of soils, which are
summarized in Table 3-6. Two of these are categorized as consolidation tests, where
the volume change of the soil is determined for a change in applied stress. The basic
information obtained from a consolidation test is shown in Figure 3-12. The remaining
two tests can be categorized as response to wetting tests, where the volume change of
the soil is measured if the soil is given access to water or if the water content is
reduced.
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Table 3-6 Tests for Volume Change with ASTM Standards

One-Dimensional Consolidation cv , mv , av
Can be performed on samples that are
D2435 Properties of Soils Using Cc , C r , σ ' p
initially partially saturated.
Incremental Loading Cα
One- Dimensional Consolidation Provides a good compression curve for
cv , mv , av
Properties of Saturated Cohesive determination of σ ' p
D4186 Cc , C r , σ ' p
Soils Using Controlled- Strain Fast, compared to D2435.
Loading Cα Test specimens must be saturated.
One-Dimensional Swell or εs Used to determine the “response to
D4546
Collapse of Soils εc wetting” of compacted or intact soils.
D4829 Expansion Index of Soils EI Often cited in building codes.
3-2.6.1 Incremental Loading Consolidation Test (ASTM D2435).
The incremental loading or incremental stress consolidation test (ASTM D2435 –

Standard Test Methods for One-Dimensional Consolidation Properties of Soil Using
Incremental Loading) is over 80 years old and is a very common test in geotechnical
engineering practice. Figure 3-13 shows the basic elements of the fixed-ring and
floating-ring consolidometers. The test is normally conducted on saturated fine-grained
test specimens, but it may also be conducted on partially-saturated soils, compacted
soils, and remolded soils. The most common test specimen size in U.S. geotechnical
engineering practice is a cylindrical specimen with a 2.5-inch diameter and 1-inch
height, although other size apparatuses are commercially available. The test specimen
is contained in a rigid ring that prevents lateral expansion of the soil during loading, thus
all displacement is vertical. This type of test is also called a one-dimensional
compression test. A porous stone is usually located above and below the test specimen
to allow for double drainage. The fixed-ring consolidometer is the most common in
geotechnical engineering practice, but the floating-ring consolidometer can be used if
high friction between the ring and soil is anticipated (e.g. sandy clay).
As loads are applied to the test specimen, the displacement of the top platen is
measured over time. Each load is normally applied for a specific time period (i.e. 24
hours) or until the end of primary consolidation (EOP) is achieved. An example time-
deformation curve or time curve for one load increment is shown in Figure 3-12. The
time curve is important in that the value of the coefficient of consolidation ( cv ) is
determined from this curve. Details on calculating the value of cv from the time curve
can be found in Chapter 5 or ASTM D2435.
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Figure 3-12 Basic Information Obtained from a Consolidation Test
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Figure 3-13 Fixed-Ring and Floating-Ring Consolidometers
The load applied to the test specimen is usually doubled for each load increment. The
Load Increment Ratio (LIR) is used to quantify the change in load to the test specimen,
and is defined below:
∆σ
LIR = (3-4)
σ0
where:
∆σ = change in applied stress, and
σ 0 = initial total stress.
An LIR = 1 corresponds to doubling the load on the test specimen. For unloading, an
LIR = -0.75 is often used, which means the immediately previous load is skipped. For
reloading, an LIR = 4 is used, which follows the same stresses as the unloading cycle
until the past load has been reached.
Each load applied to the test specimen; for unloading, rebound, and reloading cycles;
provides a data point for the compression curve. An example compression curve is
shown in Figure 3-12. The conventional method used to plot the compression curve is
using void ratio (y-axis) and the logarithm of the effective vertical stress (x-axis). 10 The
compression curve is used to determine the preconsolidation pressure or maximum
past pressure ( σ ' p or Pp ) and the compression index ( Cc ) and the recompression index
( Cr ). The compression curve can also be plotted using axial strain instead of void ratio.
For plots using strain, the compression parameters are Cε c and Cε r (instead of Cc and
10 The compression curve is often called the “e-log p” curve in older geotechnical publications.
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Cr ). The strain or void ratio used in plotting the compression curve can either be the
value at the end of the time increment or the value at the end of primary consolidation.
It is important that the method used is indicated for the plot.
If a 24-hour load cycle is used for the incremental stress consolidation test, the test can
take two to three weeks to complete, depending on the value of the maximum stress
and the number of unload-reload loops. Laboratories often try to decrease the amount
of time required by reloading the test specimen at the end of primary (EOP)
consolidation as opposed to constant time intervals. If the test is conducted using an
automated apparatus, the time corresponding to EOP is often determined by a
computer program. The use of computer-calculated EOP times may incur errors,
especially for low stresses where strains are small. If the test specimen is reloaded too
quickly at early stages of the test, the remaining data may not be useable and the
quality of the test may be compromised.
3-2.6.2 Constant Rate of Strain Consolidation Test (ASTM D4186).
The constant rate of strain (CRS) consolidation test (ASTM D4186 – Standard Test
Method for One-Dimensional Consolidation Properties of Saturated Cohesive Soils
Using Controlled-Strain Loading) can be used as an alternative to the incremental
stress consolidation test. The test specimens for the CRS test are the same size as for
the incremental stress tests. One of the main advantages of the CRS test is that the
compression curve can be obtained in much less time than the incremental stress test.
A schematic of the basic test elements and the data acquired is given in Figure 3-14.
The CRS consolidation test can be conducted on intact soil specimens, and also on
compacted and remolded soil specimens. However, it is necessary that the test
specimen be saturated prior to compression. As shown in Figure 3-14, the
consolidometer for the CRS test is very similar to a triaxial cell. The test specimen is
back-pressure saturated in the same manner as a triaxial specimen. The cell pressure
applied in the cell serves as back pressure in the process. The test specimen is drained
only at the top, and during the loading process, the excess pore water pressures are
measured at the bottom. Based on an assumed parabolic distribution of pore water
pressure in the test specimen, the average effective stress can be calculated.
The test specimen is loaded at a constant strain rate ( ε ). The strain rate is selected
such that the excess pore water pressure measured at the base of the test specimen
does not exceed 15% of the applied stress. The rate for unloading the test specimen is
much slower than loading, and unload-reload loops may slow down the test
considerably.
One advantage of the CRS consolidation test is that the compression curve is defined
by many more data points than the incremental stress consolidation test. There are so
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many data points taken that the results are usually portrayed as a curve instead of
discrete data points. This allows an increased resolution of the compression curve in
the vicinity of the preconsolidation pressure, and allows a more accurate determination
of its value. The CRS consolidation test also allows the determination of the coefficient
of consolidation ( cv ) over the entire load range as long as the excess pore pressures
are in an acceptable range. This method of determining cv removes some of the
subjectivity involved in determining cv using time curves with the incremental stress
consolidation test.
Figure 3-14 Basic Elements of a Constant Rate of Strain Consolidation Test
3-2.6.3 Swell and Collapse Test (ASTM D4546).
When a partially saturated natural soil or a compacted soil is given access to water, the
soil can swell at low stresses or collapse at high stresses. ASTM D4546 (Standard Test
Methods for One-Dimensional Swell or Collapse of Soils) allows the volume change of
the test specimen to be measured as a function of the applied stress. If the soil has no
confining stress applied, then free swell may occur. Often, free swell is measured under
a nominal stress of 20 psf. If the pressure is varied to prevent any swell or volume
change from occurring, this pressure is called the swell pressure for the soil. ASTM
D4546 allows the amount of free swell, swell pressure, and volume change for other
stresses to be determined.
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The test is very similar to an incremental stress consolidation test, and the same
consolidometer is used. Loads can be applied to the test specimen, and after the test
specimen has achieved equilibrium under the load, the specimen is inundated. The
swell or collapse volume change is measured after the test specimen achieves a new
equilibrium. Shown in Figure 3-15 is an example curve that gives the volume change of
the soil as a function of the stress at inundation.
Figure 3-15 Volume Change of Soil as a Function of Stress at Inundation
3-2.6.4 Expansion Index Test (ASTM D4829).
The expansion index test (ASTM D4829 – Standard Test Method for Expansion Index of
Soils) determines the swell potential of a soil, but it is less rigorous than D4546. It does
not provide engineering design values to calculate volume change, but provides a
simple index to assess the swelling potential of soil. This test is conducted on
compacted soils. The main use of the expansion index test is to assess if a compacted
fill might pose problems if structures are constructed on top of it. The compaction mold
is approximately half the height of the 4-inch diameter (1/30 ft3) compaction mold used
for ASTM D698 and the same compaction hammer as used for D698 is used. The goal
is to compact the test specimen using a D698 effort at a degree of saturation of 50%. A
vertical stress of 1 psi is applied, and the specimen is allowed to equilibrate for 10
minutes. The specimen is then inundated and allowed to swell for 24 hours or until the
swell has essentially ceased. The Expansion Index ( EI ) is equal to the percent swell
multiplied by 10. The expansion potential is assessed using the criteria shown in Table
3-7.
Table 3-7 Potential Expansion for EI Values

Expansion Index, EI Potential Expansion
0 – 20 Very Low
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21 – 50 Low
51 – 90 Medium
91 – 130 High
>130 Very High
3-2.7 Hydraulic Conductivity (Permeability) Tests.
Test to measure the permeability or hydraulic conductivity of soils have been used since
the 1930s. In general, permeability tests have been more widely used for fine-grained
soils as opposed to coarse-grained soils. There are many correlations available for the
permeability of coarse-grained soils, but few reliable correlations exist for fine-grained
soils. Currently, there are two ASTM standardized tests for hydraulic conductivity.
However, hydraulic conductivity can be calculated from incremental stress and constant
rate of strain consolidation tests.
3-2.7.1 Compaction Mold Test (ASTM D5856).
The compaction mold test (ASTM D5856 – Standard Test Method for Measurement of
Hydraulic Conductivity of Porous Material using a Rigid-Wall Compaction-Mold
Permeameter) is intended to be used on compacted soils having a hydraulic
conductivity less than 10-3 cm/sec.
This test has a major deficiency in that the test specimen cannot be saturated, therefore
the measured hydraulic conductivity may be too low. There also is a problem that
leakage can occur between the compacted test specimen and the wall of the
compaction mold. This test is best suited as a quality control test for compacted clay
liners for landfills and reservoirs.
3-2.7.2 Flexible Wall Test (ASTM D5084).
The flexible wall test (ASTM D5084 – Standard Test Methods for Measurement of
Hydraulic Conductivity of Saturated Porous Materials Using a Flexible Wall
Permeameter) is the most common permeability test. The apparatus used is also called
a triaxial permeameter because it is essentially a triaxial cell without the loading rod.
The sample is enclosed in a flexible membrane, therefore the problem of side-wall
leakage experienced in D5856 is not a problem. The membrane conforms to
irregularities in the sides of the test specimen. ASTM D5084 also allows for test
specimens to be back-pressure saturated in the same manner as CU and CD triaxial
test specimens. The flexible wall permeability test also allows control of the effective
consolidation stress. This is not possible with the compaction mold test.
This method is recommended only for soils having a hydraulic conductivity less than
10-4 cm/sec, so it is best suited to fine-grained soils or soils with a significant percentage
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of fines. Intact, remolded, and compacted test specimens can be used for this test.
The test takes considerable skill to properly conduct, and there are six different methods
to promote flow though the test specimen. The test measures vertical hydraulic
conductivity, but intact samples can be trimmed at different orientations to measure
anisotropy of hydraulic conductivity.
3-2.7.3 Hydraulic Conductivity from Consolidation Tests.
One of the purposes of conducting a consolidation test is to determine the coefficient of

consolidation ( cv ). For incremental stress consolidation tests, cv is calculated from the
time curves using one of several different methods. For the CRS consolidation test, cv
can be calculated at every point where the excess pore water pressure, average stress,
and strain are known.
The hydraulic conductivity of a soil is related to the value of cv by the equation below:
k = cv ⋅ mv ⋅ γ w (3-5)
where:
k = hydraulic conductivity or permeability,
cv = coefficient of consolidation,
mv = coefficient of volumetric compressibility, and
γ w = unit weight of water.
The coefficient of volumetric compressibility can be determined by plotting the strain (y-
axis) versus the arithmetic effective stress (x-axis) and determining the slope of the plot
corresponding to the stress where cv is calculated. Consolidation tests are rarely
conducted just to determine the value of permeability, but if these data are available for
a project where additional values of permeability are useful, little effort is required to
calculate the permeability.
It is also possible to determine the permeability from the consolidation phase of CU and
CD triaxial tests. This normally requires that the test specimen only be drained at the
ends (no filter paper drainage strips) and it may be necessary to consolidate the
specimen in stages leading up to the final consolidation stress.
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3-3 LABORATORY TESTS ON ROCK.
ASTM also addresses laboratory tests on rock specimens, but there are much fewer
tests for rocks than for soils. Most of the common tests on rock focus on the strength in
compression and tension. Table 3-8 lists the common rock tests that have ASTM
standards available.
Table 3-8 Laboratory Rock Strength Tests with ASTM Standards

Splitting Tensile Strength of Intact Rock
D3967 σt
Core Specimens
Laboratory Direct Shear Strength Tests c −φ Mainly interpreted as total stress
D5607 of Rock Specimens Under Constant
c '− φ ' strength parameters.
Normal Force
Determination of the Point Load Strength Is Strength index often corrected for
D5731 Index of Rock and Application to Rock
I s (50) specimen diameter of 50 mm.
Strength Classifications
Compressive Strength and Elastic Moduli
of Intact Rock Core Specimens under
D7012 σu
Varying States of Stress and
Temperatures
3-3.1 Unconfined Compression Test (ASTM D7012).
ASTM D7012 (Standard Test Methods for Compressive Strength and Elastic Moduli of
Intact Rock Core Specimens under Varying States of Stress and Temperatures)
encompasses more than just unconfined compression tests. There are four different
methods of testing outlined in the standard, and two address the unconfined
compressive strength. The other two address triaxial compression of rock, which is
used less frequently than for soil.
The basic form of the unconfined compression test (Method C) does not measure axial
strain during loading, and provides only the unconfined compressive strength. The
resulting strength can be used for design or as an index property for the rock. The test
is normally conducted on rock cores that are 1.85 inches in diameter, and they are
trimmed to be at least 3.7 inches in height. This specimen diameter corresponds to an
NQ core barrel, but larger rock cores can be used as well. The specimen can be loaded
at a constant rate of load or a constant rate of strain which are chosen to cause failure
in 2 minutes to 15 minutes. Unconfined compression tests of rock are normally
conducted on many test specimens since there can be wide variations in the
compressive strength due to the effects of planes of weakness (joints, fractures, and
faults) and other inhomogeneities is rock. The main result of the unconfined
compression test in the uniaxial compressive strength ( σ u ).
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3-3.2 Split Cylinder Test (ASTM D3967).
The split cylinder test (ASTM D3967 – Standard Test Method for Splitting Tensile
Strength of Intact Rock Core Specimens) is used as an alternative to the direct tensile
test. The test specimen is loaded diametrically via hardened steel end platens. The
test specimen thickness can range between 0.2 to 0.75 times the specimen diameter.
The test specimen is loaded at a rate sufficient to obtain failure in 1 to 10 minutes. The
main result of this test is the splitting tensile strength ( σ t ). Although the tensile strength
resulting from the split cylinder test should be essentially equal to that measured from
direct tension tests, it is customary to preface the former with “splitting.” Owing to the
variability in the test results, split cylinder tests are often run on numerous test
specimens.
3-3.3 Rock Direct Shear Test (ASTM D5607).
Rock direct shear tests (ASTM D5607 - Standard Test Method for Performing
Laboratory Direct Shear Strength Tests of Rock Specimens Under Constant Normal
Force) can be conducted on intact rock specimens, as well as on joints and
discontinuities. Unlike the direct shear test conducted on soils, the rock direct shear
test is normally considered to be an undrained test. The basic elements of this test are
the same as for the soil direct shear tests. Rock direct shear tests are often conducted
for a range of normal stresses to determine the strength envelope for the material. One
major difference is that the test specimen is often encapsulated in a super strength
gypsum cement to fix its position in the shear box. An example of the test fixture is
Figure 3-16 Specimen Container for Rock Direct Shear Test (after ASTM D5607)
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Figure 3-17 Rock Direct Shear Apparatus for High Normal and Shear Loads
Rock direct shear tests are more complicated than the soil counterpart because of the
great variety of loads and displacements the apparatus is required to measure. For
large normal stresses, it can take 50,000 lbs. to fail an intact rock direct shear test
specimen, and failure may occur at very small (<0.01 inch) displacements. For rock
joints at low normal stresses, the failure load might be less than 100 lbs., and the
displacement at failure may be greater than 0.1 inch. Shown in Figure 3-17 is a rock
direct shear apparatus for high loads and normal stresses.
3-3.4 Point Load Test (ASTM D5731).
The point load test (ASTM D5731 – Standard Test Methods for Determination of the
Point Load Strength Index of Rock and Application of Rock Strength Classifications)
provides an index value of the rock strength. This test can be performed on rock cores
or irregular pieces of rock having diameters in the range of 1 inch to 3 inches. A
photograph of the point load apparatus is shown in Figure 3-18. The diameter of the
test specimen is considered to be the thickness of the test specimen from loading platen
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contact points. The load platens are truncated cones, with the point being a
semicircular arc of 0.2-inch radius.
Figure 3-18 Point Load Apparatus for Rock Index Testing
The point load test provides an uncorrected point load strength index ( I s ). This value
can be corrected to reflect differences in test specimen sizes, and is often normalized to
an equivalent core diameter of 2 inches (50 mm). This corrected value is called the size
corrected point load strength index ( I s (50) ). The results of the point load test are used
for rock classification and can be correlated to the uniaxial compressive strength, but
the values are not considered to have sufficient reliability for design.
3-4 OTHER SOIL AND ROCK TESTS.
This section of the manual has addressed the tests that are the most common in
geotechnical engineering practice in the U.S., but there are hundreds of other ASTM
standardized tests. There are groups of tests that address partially saturated soils, soil-
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cement mixtures, peats and organic soils, geosynthetics, and many other materials
used in engineering projects.

Topic Reference
Head, K. H. 2008. Manual of Soil Laboratory Testing, Vol. 1., 3rd Ed., Whittles, 416 pp.
Head, K. H. and R. J. Epps 2011. Manual of Soil Laboratory Testing Vol. II, Permeability
General Laboratory
Shear Strength, and Compressibility Tests, 3rd Ed., Whittles, 512 pp.
Testing
Head, K. H. and R. J. Epps 2014, Manual of Soil Laboratory Testing Vol. III, Effective
Stress Tests, 3rd Ed., Whittles, 448 pp.
Lambe, T. W. and R. V. Whitman 1969. Soil Mechanics, John Wiley and Sons, Inc. 553
pp.
Laboratory Shear Testing of Soils, ASTM STP 361, American Society for Testing and
Materials, 505 pp., 1964.
Laboratory Shear Strength of Soils, ASTM STP 740, R. N. Young and F. C. Townsend,
Eds., 717 pp, 1981.
Shear Strength
Research Conference on Shear Testing of Cohesive Soils, ASCE, University of
Colorado, Boulder, CO, 1164 pp, 1960.
Saada, A. S. and Townsend, F. C. 1981. "State of the Art: Laboratory Strength Testing
of Soils," Laboratory Shear Strength of Soils, ASTM STP 740, R. N. Yong and F. C.
Townsend, Eds., ASTM, pp. 7-77.
Advanced Triaxial Testing of Soil and Rock, ASTM SPT 977, Robert T. Donague,
Ronald C. Chaney, and Marshal L. Silver, Eds., American Society for Testing and
Materials, Philadelphia, 896 pp, 1988.
Triaxial Testing
Bishop, A. W. and D. J. Henkel (1957), Measurement of Soil Properties in the Triaxial
Test, Edward Arnold, Ltd., London, 190 pp.
Lade, P. 2016. Triaxial Testing of Soils, John Wiley & Sons, Ltd., 500 pp.
3-6 NOTATION.
Symbol Description
Total stress cohesion

c
Cc Compression index
Cr Recompression index
cv Coefficient of consolidation
Effective stress cohesion

c'
D Diameter
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Symbol Description
Dr Relative density
e Void ratio
emax Maximum index void ratio
emin Minimum index void ratio
Gs Specific gravity
H Height
Is Uncorrected point load strength index
I s (50) Size corrected point load strength index
k Hydraulic conductivity or permeability
K0 At-rest earth pressure coefficient
L Length
mv Coefficient of volumetric compressibility
n Porosity
Pp Maximum past pressure
RC Relative compaction
S Degree of saturation
St Sensitivity
su Undrained shear strength for a φ = 0 envelope for saturated soils
t Thickness
V Total volume
Va Volume of air
Vs Volume of solids
Vv Volume of voids
Vw Volume of water
Water content
w
wopt Optimum water content
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Symbol Description
Ws Weight of solids
WT Total weight of sample
Ww Weight of water
γb Buoyant unit weight or submerged unit weight
γd Dry unit weight

Maximum dry density from the compaction curve for a particular effort or maximum index dry
γ d − max density (corresponding to emin )
γ d − min Minimum index dry density (corresponding to emax )
γT Total or wet unit weight
γ sat Saturated unit weight
γw Unit weight of water
Change in applied stress

∆σ
ε Strain rate
σ0 Initial total stress
σ3 Zero confining pressure
σ 'p Preconsolidation pressure
σ 'v Vertical effective consolidation stress
τ cyc Applied peak cyclic shear stress
Total stress friction angle

φ
Effective stress friction angle
φ'
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DISTRIBUTION OF STRESSES
4-1 INTRODUCTION.
4-1.1 Scope.
This chapter describes the analysis of stress conditions within the ground, including
stress at a point, changes in stress caused by the application of soil and structural
loads, and empirical methods for estimating loads on buried pipes, conduits, shafts, and
tunnels. The calculation of stresses and changes in stress using numerical methods,
such as the finite element method, is also discussed.
4-1.2 State of Stress.
The state of stress within the ground can be analyzed assuming that either elastic or
plastic conditions prevail. Elastic solutions are most appropriate for cases in which the
shear stresses throughout the soil mass are significantly below the shear strength of the
soil and shear failure is not likely. If the shear stress in the soil is less than about one-
third of the ultimate shear strength, the stresses within the soil mass will be roughly
equal to values calculated from elastic theory (Davis and Selvadurai 1996). The stress
conditions calculated using the most of the methods in this chapter assume that elastic
conditions prevail.
Plastic solutions assume full mobilization of the soil’s shear strength within a soil mass
or along a specified failure surface. Plastic equilibrium is used for problems, such as
slope stability (Chapter 7), foundation bearing capacity, and lateral earth pressures,
where shear strength may be fully mobilized.
4-2 STRESS CONDITIONS AT A POINT.
4-2.1 Stress Conditions in Soil.
Soil consists of a compilation of discrete particles, water, and air in varying proportions.
Similarly, rock may contain a combination of the mineral components and any void
space that may be filled with water or air. These discrete systems are idealized in
stress analysis by assuming the soil acts as continuous solid mass without holes or
gaps. In this continuum manner, stress is simply conceived as force per unit area and
the contact forces at the soil particle level are not considered.
Stress in soil is the result of forces from the self-weight of the overlying and surrounding
soil plus any external loading, such as structures or ponded water. For a given plane, it
can be particularly useful to consider stress in terms of its normal and shear
components. The normal stress can be defined as the sum of the forces acting
perpendicular to a plane divided by the area of that plane. Similarly shear stress in a
particular direction is ratio of the force acting tangent to a plane divided by its area.
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The total normal stress on any plane within the ground is based on the sum of all forces
acting on the plane in question. The total stress may be divided into two parts: the
effective normal stress and the pore pressure.
4-2.1.1 Total Vertical Stress.
The total vertical stress (or overburden stress) is the normal stress acting on a
horizontal plane at some depth within the soil. The total vertical stress, σ v , at a
particular depth is calculated by multiplying the thickness ( zi ) of all overlying materials
by the total unit weight ( γ t ,i ) of each material:
n
σ v = ∑ ziγ t ,i (4-1)
i =1
It is imperative to include the weight of water resting on the ground surface (i.e., ponded
water) in calculations of total vertical stress. Ponded water can be considered by
adding a layer to the total stress calculations with thickness equal to the water depth
and unit weight equal to the unit weight of water.
The calculation of total vertical stress is illustrated in Figure 4-1.
4-2.1.2 Pore Water Pressure.
The energy present in ponded water or groundwater is often expressed in terms of the
total hydraulic head, which has pressure, elevation, and velocity components. The
velocity component is typically ignored in most geotechnical applications. The pore
water pressure ( u ) can be found from the pressure head ( hp ) as
u = hpγ w (4-2)
where:
γ w = the unit weight of water (62.4 pcf or 9.81 kN/m3).
When water is static (not flowing), the total head is constant throughout the system, and
elevation head converts directly to pressure head. This is referred to as a hydrostatic
condition, and the pressure head is simply equal to the distance below the groundwater
table or phreatic surface.
Flowing water loses energy as it flows through the soil and the total head decreases in
the direction of flow. For flowing water conditions, a flow net or some other type of
seepage analysis must be performed. Pore water pressure at any point can be
determined by first calculating the total head and the elevation head at any point in the
ground. The pressure head for flowing water is found by subtracting the elevation head
from the total head. Water pressures act in all directions equally because the water
does not sustain shear stress. For this reason, orientation of the pore water pressure is
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not important. In some older publications, pore water pressure is also called neutral
stress.
Figure 4-1 Calculation of Vertical Stresses for Hydrostatic Conditions
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4-2.1.3 Effective Vertical Stress.
The effective vertical stress ( σ 'v ) is found by subtracting the pore water pressure at any
point from the total vertical stress at the same point
σ=
'v σ v − u (4-3)
4-2.1.4 Horizontal Stress.
Horizontal stress in a soil mass is influenced by the effective vertical stress, the geologic
stress history, and lateral confinement conditions. Horizontal stress cannot be
calculated directly from the soil profile and is typically calculated as a proportion of the
effective vertical stress:
σ 'h= K ⋅ σ 'v (4-4)
The lateral earth pressure coefficient ( K ) depends on stress history and lateral
confinement conditions. Common types, applications, and sources of lateral earth
pressure coefficients are summarized in Table 4-1.
The total horizontal stress can be found by adding the pore water pressure at any point
onto the effective horizontal stress
σ=
h σ 'h + u (4-5)
Table 4-1 Lateral Earth Pressure Coefficients
Lateral earth pressure

Example applications Method to obtain
coefficient
Level, natural ground Estimate based on φ ' and OCR

At-rest, K 0
Unyielding retaining wall Measure with field tests
Near crest of slopes

Active, K A
Behind yielding retaining walls Calculate with analytical methods
(see DM 7.2)
Near toe of slopes Estimate based on experience
Passive, K P
In front of retaining wall toe
4-2.1.5 Applied Loads.
Many civil engineering applications must consider the effects of external (non-soil) loads
applied at the surface or at some depth within the soil mass. The influence of existing
loads must be included in total stress calculations. New loads cause changes in total
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stress within the soil. These load changes will cause changes in the pore water
pressures as discussed in more detail in Chapter 5. The duration of the changes in
pore water pressure will depend on the permeability and compressibility of the soil.
Analytical methods for calculating changes in stress are provided in Section 4-3,
including methods for point loads, line loads, and uniformly loaded areas. In Section 4-
6, numerical methods for calculating changes in stress are summarized.
Changes in total stress caused by applied loads should be within the soil mass to at
least the critical depth. The critical depth is the depth over which soil compression
caused by the changes in stress contributes to significant surface settlement. The
critical depth in fine-grained soils corresponds to the depth at which the change in
stress is less than 10% of the existing vertical effective stress. In coarse-grained soils,
the critical depth occurs when the change in stress is less than 20% of the existing
vertical effective stress.
Interactions between the applied load and the soil foundation must be considered,
especially for changes in stress very close to the load. The flexibility of the structure
that applies a distributed load to the soil affects the distribution of the change in stress.
A completely flexible load, such as a soil fill, will apply a uniform stress to the soil
because the load can deform in proportion to the soil. The elastic solutions presented in
Section 4-3 assume that the load is completely flexible.
A foundation that is completely rigid with respect to the soil must undergo a uniform
deformation. When a rigid foundation deforms uniformly into an elastic solid (i.e.,
undrained conditions for clay), the load must shift to the edges of the foundation,
resulting in a pressure distribution that increases toward the edge (see Figure 4-2). In
contrast, a rigid foundation on sand will cause yielding near the edges, resulting a
pressure distribution that decreases toward the edge.
Figure 4-2 Variation in Contact Pressure – a) Rigid Foundation and

b) Completely Flexible Foundation
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4-2.2 Mohr Circle of Stress.
The normal and shear stress on a plane at any point within the ground depends on the
orientation of that plane with respect to the orientation of the stress system. At any
point, the will be three mutually perpendicular planes that have no shear stress, which
are referred to as the principal planes. The normal stresses that act on these planes
are defined as the principal stresses. The major principal stress ( σ 1 ) has the largest
magnitude. The minor principal stress ( σ 3 ) has the smallest magnitude. The
intermediate principal stress ( σ 2 ) falls between σ 1 and σ 3 . For two-dimensional
problems, σ 2 is either assumed equal to σ 3 or ignored. For level ground conditions, the
principal stresses are often assumed to be aligned with the horizontal and vertical
directions with the horizontal normal stress being equal in all directions. The sign
convention used herein assigns positive values to compressive stress, shear stress that
causes counterclockwise rotation, and counterclockwise angles.
A Mohr circle of stress can easily be plotted from σ 1 and σ 3 , or the normal and shear
stresses on any two perpendicular planes. More information on the use of the Mohr
circle can be found in Parry (2004). From the Mohr circle for a point, the normal and
shear stress conditions on any plane can be determined by rotating an angle 2α about
the center of the circle, where α is the angle between the major principal plane and the
plane of interest. Figure 4-3 illustrates the Mohr circle and mathematical relationships
between common stresses.
4-3 ELASTIC SOLUTIONS FOR STRESSES DUE TO APPLIED LOADS.
4-3.1 Use and Applicability.
The elastic solutions presented in this section are useful for simple analyses of changes
in stress, especially consolidation settlement. These methods are also useful for
understanding the principles of stress distribution and for checking more complicated
numerical analyses.
4-3.2 Semi-Infinite Elastic Conditions.
4-3.2.1 Assumptions.
The Boussinesq and related solutions assume the soil is a homogeneous, elastic
material in which continuity is maintained and static equilibrium is satisfied. The applied
load is completely flexible and is applied at the surface of the material. For
embankment loading, the load is strictly vertical and no shear stress is applied to the
foundation by the embankment. As discussed in Section 4-2.1.5, the stress distribution
below a rigid foundation is not uniform and may not conform to the assumptions of the
Boussinesq solutions.
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Figure 4-3 Mohr Circle Relationships
Loads with a length to width ratio ( L B ) of at least five are commonly assumed to result
in plane strain conditions, at least near the middle of the length. Under plane strain,
deformation only occurs perpendicular to the long axis of the load and the changes in
stress do not depend on the elastic properties of the material.
The Boussinesq solutions are not typically applicable to the calculation of shear stress
for conditions where shear stress is becoming critical. In this case, the soil is
approaching a state of plastic equilibrium and the assumption of elasticity no longer
applies. In such cases, stability analysis methods, such as those in Chapter 7 and in
DM 7.2, should be used.
4-3.2.2 Stress Distribution Formulas.
Formulas for homogeneous, semi-infinite, isotropic foundations are summarized in

Table 4-2. These formulas can be used for hand calculations for simple computations
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and to check the results of more complex numerical analyses. Such formulas can also
easily be programmed into a spreadsheet solution. Additional formulas for other
geometric and loading conditions are summarized in Poulos and Davis (1974).
Horizontal and shear stresses caused by applied loads can also be determined from
elastic solutions. In many cases, these calculations require a value of Poisson’s ratio
(ν ) for the soil. Many of the common figures assume ν = 0.5 making it important to
verify the value of ν that was used. One application of horizontal stress calculations is
for the loading of unyielding walls as discussed in DM 7.2. For conditions where elastic
solutions are suitable, the calculation of shear stress is typically not required.
Table 4-2 Equations for the Calculation of Change in Vertical Stress

Below Various Loading Conditions
Loading Condition Stress Diagram Equation
3
3Qz
σz = 5
2π R
Point Load
R= x2 + y 2 + z 2
2 P z3
Uniform Line Load of σz =
π R4
Infinite Length
R
= z 2 + x2
q
Uniform Strip Load σz= [
0 α + sin α cos α (α + 2γ )
]
π
(Figure 4-4)
q0  −1 yx xyz  1 1 
σz
=  tan +  2 + 2 
2π  zR3 R3  R1 R2  

Uniformly Loaded
Rectangular Area
(Figure 4-5)
2 2 1/2 2 2 1/2
(y + z )
R1 = (x + z )
R2 =
2 2 2 1/2
R3 = ( x + y + z )
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Table 4 2 (cont.) Equations for the Calculation of Change in Vertical Stress

Below Various Loading Conditions
Loading Condition Stress Diagram Equation
Uniformly Loaded Circular  

 1 
Area σ z q0 1−
= 
3
(Figure 4-6)

(
 1+( r z )

2
) 2 


q0  xα a +b − x 
Triangular Load =σz + β
π  a b 
q0
Slope Load σz
= [ xβ + z ]
πa
q0
Terrace Load σz
= [ aβ + xα ]
πa
q0  xz 
Semi-infinite Uniform Load σz
= β+
π  R 2 
4-3.2.3 Chart Solutions for Vertical Stress beneath Regular Loads.
Figure 4-4 to Figure 4-7 provide chart solutions for regularly shaped loads and
Boussinesq theory. These charts are all based on the assumption of ν = 0.5 , where
required, to determine the change in vertical stress. Example calculations are provided
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in Figure 4-9. Additional guidance on the calculation of changes in stress under elastic
conditions can be found in Poulos and Davis (1974).
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Figure 4-4 Vertical Stress Contours from Strip and Square Loaded Areas –
Boussinesq
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Figure 4-5 Influence Factors for a Rectangular Loaded Area – Boussinesq
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Figure 4-6 Influence Factors for a Circular Loaded Area – Boussinesq

(after Ahlvin and Ulery 1962, Poulos and Davis 1974)
In the past, changes in stress caused by irregularly shaped loaded areas were
calculated using chart solutions such as those proposed by Newmark (1942) and
Jimenez Salas (1948). However, for complex loading conditions, numerical analysis
has become the most common means to evaluate changes in stress.
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Figure 4-7 Influence Factors for Embankment Loading – Boussinesq

(after Poulos and Davis 1974)
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4-3.2.4 Superposition.
The assumption of linear elasticity inherent in the Boussinesq solutions allows for
superposition of stresses that result from applied loads. This means that the change in
stress from one load can be added or subtracted from those caused by other loads,
provided the same point is being considered in all cases. This principle is especially
useful for determining the change in stress below and outside of loaded areas as
illustrated in Figure 4-8.
Figure 4-8 Use of Superposition to Determine Change in Vertical Stress
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Figure 4-9 Stress Distribution Examples
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4-3.2.5 St. Venant’s Principle.
St. Venant’s principle is another useful concept for the calculation of change in stress
due to applied loads. According to this principle, the change in stress caused by two
statically equivalent loads becomes equal as the distance from the load becomes
sufficiently large. Practically, this means that a square or circular load can be replaced
with a point load, or a strip load can be considered a line load. It is commonly assumed
that St. Venant’s principle can be applied to the calculation of change in vertical stress,
∆σ z , for depths greater than three times the width of the applied load.
4-3.3 Layered or Anisotropic Foundations.
While the Boussinesq solutions offer a relatively simple means to calculate changes in
stress, soil is not a homogeneous, isotropic, and semi-infinite medium. For example,
different layers typically have different values of elastic modulus. Soil layers are often
more rigid horizontally than vertically. These deviations from the assumptions of
Boussinesq have led to the development of other methods for the calculation of
changes in stress, most notably the Westergaard type of analysis.
4-3.3.1 Westergaard Analysis.
Westergaard analysis assumes that the soil below the load is reinforced by closely
spaced horizontal layers that prevent horizontal displacement. This reinforcement effect
causes the changes in stress predicted by Westergaard to be less than those calculated
by the Boussinesq assumptions. The Westergaard type of analysis is most appropriate
for soil profiles that have alternating layers of stiff and soft soils, such as soft clay with
intermittent horizontal layers of sand. Figure 4-10 provides influence factors for points
below the corner of a rectangular loaded area, assuming Westergaard theory.
4-3.3.2 Layered Foundations.
Soil profiles may have layers of significant thickness and very different elastic
properties. The changes vertical stress induced in these cases differs significantly from
that predicted by Boussinesq assumptions. Analytical and chart-based solutions have
been suggested to account for such differences using rigidity factors (e.g., Mehta and
Veletsos 1959). For these conditions, numerical analysis is the preferred method to
determine changes in stress or as a means of comparison to simpler solutions.
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Figure 4-10 Influence Factors for a Rectangular Loaded Area – Westergaard
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4-4 SHALLOW PIPES AND CONDUITS.
4-4.1 General.
The stresses on shallow pipes and conduits due to applied loads is one important
application of concepts presented in this chapter. The factors influencing these
stresses include the relative rigidity of the pipe to the soil, the depth of cover, the type of
loading, the maximum width (span) of the structure, the method of construction, and the
shape of the pipe.
This section presents simple empirical procedures based on observations. More

detailed analysis can be conducted numerically or by consulting Moser (1990) or
American Lifelines Alliance (2001).
4-4.2 Vertical Loads on Rigid Pipe.
Rigid pipes are those made of precast concrete, cast-in-place concrete, or cast iron.
4-4.2.1 Dead Load.
Estimates of the load caused by vertical stress on a rigid pipe can be made using the
approach suggested by Marston and Anderson (1913) and subsequent work by
Spangler (1948). The load per length of pipe ( Wd ) in a trench can be calculated as:
Wd = Cd γ t Bd2 (4-6)
where:
Cd = a load coefficient,
γ t = the total unit weight of the soil, and
Bd = the width of the trench.
The value of Cd can be calculated as:
− K µ ' H 
Bd 
1− e 
Cd = (4-7)
2K µ '
where:
H = the depth of the trench above the top of the pipe,
K = a lateral earth pressure coefficient, and
µ ' = the coefficient of friction for the trench fill.
Recommended values of γ t , K , and µ ' for use with this equation are provided in Table
4-3 as summarized by Moser (1990).
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Table 4-3 Recommended Values for Trench Load Coefficient (after Moser 1990)
Unit weight, Lateral earth Coefficient of

Soil type γ (pcf) pressure K×µ'
friction, µ '
coefficient, K
Partially compacted moist topsoil 90 0.33 0.5 0.17
Saturated top soil 110 0.37 0.4 0.15
Partially compacted moist clay 100 0.33 0.4 0.13
Saturated clay 120 0.37 0.3 0.11
Dry sand 100 0.33 0.5 0.17
Saturated sand 120 0.33 0.5 0.17
4-4.2.2 Live Load.
The primary live loads considered in the design of buried pipes are those from vehicles,
including trucks, railroad, and airplanes. The stress at the top of the pipe is typically
calculated using Boussinesq theory multiplied by an impact factors that account for
dynamic effects. The equations presented in Table 4-2 can be used to estimate the
vertical stress transferred to the top of the pipe from surface loading. Appropriate live
load impact factors are summarized in Table 4-4.
Live load pressures (including an impact factor of 1.5) are summarized in Table 4-5.
The values in this table indicate that the changes in stress become negligible below
depths of 8 feet, 30 feet, and 24 feet for standard truck, railroad, and airport loads,
respectively.
Table 4-4 Impact Factors for Live Loading of Buried Pipe

(from American Lifelines Alliance 2001)
Depth of cover Taxiways and

Highway Railway Runway
above pipe (ft) aprons
0 to 1 1.50 1.75 1.00 1.50
1 to 2 1.35 1.50 1.00 1.35
2 to 3 1.15 1.50 1.00 1.35
Over 3 1.00 1.35 1.00 1.15
4-4.3 Vertical Loads on Flexible Pipe.
Flexible pipes include corrugated metal, plastic, and thin-wall smooth steel pipes.
These pipes deform when loaded and develop horizontal restraining pressures on the
sides that may be approximately equal to the vertical pressure if the backfill is well-
compacted. The vertical pressure on the top of the pipe depends on the surrounding
soil. In highly compressible soil, the vertical pressure may exceed the overburden
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pressure. In contrast, arching in coarse-grained soils may significantly reduce the

overburden pressure.
Table 4-5 Live Load Pressures from Various Vehicle Loading

(after American Lifelines Alliance 2001)
Depth of Live load transferred to pipe (psi)

cover
above pipe Airport
Truck Load Railway Load
(ft) (180 kip gear
(AASHTO HS-20) (Cooper E-80)
assembly)
1 12.5 Not recommended
2 5.56 26.39 13.14
3 4.17 23.61 12.28
4 2.78 18.4 11.27
5 1.74 16.67 10.09
6 1.39 15.63 8.79
7 1.22 12.15 7.85
8 0.69 11.11 6.93
10 7.64 6.09
12 5.56 4.76
14 4.17 3.06
16 3.47 2.29
18 2.78 1.91
20 NA 2.08 1.53
22 1.91 1.14
24 1.74 1.05
26 1.39
28 1.04 NA
30 0.69
4-4.3.1 Dead Load.
For very flexible pipe with outside diameter ( Bc ) the Marston load theory predicts a load
( Wc ) of:
Wc = Cd ⋅ γ t ⋅ Bc ⋅ D (4-8)
where:
Cd = the load coefficient for rigid pipe (see Equation 4-7),
γ t = the total unit weight of the trench backfill, and
D = the outer diameter of the pipe (Moser 1990).
This method assumes that the pipe stiffness and soil stiffness are equal, which may
lead to unconservative values of Wc .
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The prism load is a more conservative approach for determining the dead load on a
buried flexible pipe. The prism load ( W p ) is simply the total weight of the soil above the
pipe and is equal to:
Wp = γ t ⋅ H ⋅ D (4-9)
where:
γ t = the backfill total unit weight,
H = the depth of soil cover, and
D = the outer diameter of the pipe.
4-4.3.2 Live Load.
The Boussinesq approach described in Section 4-4.2.2 should also be used to calculate
live loads on flexible pipes.
4-4.4 Long Span Metal Culverts.
The previously discussed methods do not apply to the calculation of stresses and
design of long span metal culverts. The use of finite element analysis software
specifically formulated for culverts, such as CANDE, is likely required. For additional
guidance see Duncan (1979).
For the design of long span metal culverts, the engineer must distinguish between
shallow and deep cover conditions (Duncan 1979). Shallow cover conditions apply to
cases where the cover is less than one-fourth of the culvert span. Culverts with shallow
cover must be designed for flexural stresses caused by live loads. The factor safety is
calculated by comparing the predicted axial stresses and moment loading to the culvert
capacity.
In contrast, deep cover conditions occur when the cover is greater than one-fourth of
the span. Deep cover culverts only require design for ring compression, such that the
seams don’t collapse under the design loads. Design axial ring loads are higher than
those calculated solely based on ring compression theory. This occurs because the
culvert is much stiffer in ring compression than the surrounding soil and a negative
arching condition occurs.
4-5 DEEP UNDERGROUND OPENINGS.
4-5.1 General Factors.
Prior to excavation of a tunnel, the rock or soil is typically at a state of equilibrium under
the stresses imposed by overburden and external loads. Excavation disturbs that
equilibrium condition and requires load to transfer to surrounding rock, soil, or tunnel
support system. The soil and rock will always exhibit an immediate response to the
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changes in stress caused by the excavation. Often a time-dependent response also

occurs, especially in saturated compressible soils. A good discussion of the
development of equilibrium conditions during and after tunneling in soft ground, as well
as tunneling and support methods, is provided in FHWA (2009).
Changes in stress are typically accompanied by displacement of the rock, soil, or

support structure. Similar to retaining walls, some movement is desirable to create a
suitable balance between the load carried by the structure and the load distributed to
the soil. Due to the effects of this soil-structure interaction, the type of support system
and tunneling method used will significantly affect the deformations experienced during
and after tunneling, and therefore the loading imposed on tunnel support.
The stresses acting on an underground opening will also depend on the depth of the
opening below the ground surface and the characteristics of the surrounding soil or
rock. One common distinction between the terms deep and shallow compares the
depth of cover to the diameter of the opening. Openings for which this ratio is less than
2 should be considered shallow and arching of the soil or rock should be ignored. Deep
openings have a cover to diameter ratio greater than 3 and benefit from the effects of
arching.
For deep underground openings, deformation toward the opening allows the release of
stress and the development of arching in the surrounding soil or rock. For this reason,
the stresses are heavily dependent on the amount of deformation allowed during
construction and the degree of restraint provided by the lining.
Numerical methods, such as finite element analysis, can be used to calculate stresses
and deformations of underground openings. These methods can be quite accurate and
account for significant complexity provided the soil and rock are characterized properly
and the construction sequence is adequately modeled. The methods presented in this
section are useful for simple calculations and to check the results of more complex
numerical models.
4-5.2 Openings in Rock.
Rock can be separated into two groups for the purposes of determining stresses: (1)
sound, non-swelling rock that can sustain considerable tensile stress and (2) fractured,
blocky, seamy, squeezing, or swelling rock. For more detailed explanations of rock
properties, see Chapter 1. The behavior of these two groups is distinguished primarily
by the ability of the rock to resist tensile stress and/or significant deformation.
4-5.2.1 Sound Rock.
Elastic analysis can be used to determine stresses surrounding tunnels or openings in

intact, isotropic rock (e.g., crystalline igneous, homogeneous sandstone and limestone).
Analytical methods are summarized in rock mechanics texts such as Goodman (1989).
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For these materials, stresses in rock surrounding spheroidal cavities are lower than
those for tunnels with the same cross-section. Elastic analysis can be used to
determine the best arrangement of openings and pillars, such that supports are
provided at locations of stress concentration.
4-5.2.2 Broken and Fractured Rock.
Pressure on tunnels in chemically or mechanically altered rock must be analyzed by

approximate rules based on experience, such as those presented in Table 4-6. The
rock conditions used in Table 4-6 are compared to other common rock quality indices in
Table 4-7.
4-5.2.3 Squeezing and Swelling Rock.
Rocks in categories 7 to 9 of Table 4-6 and Table 4-7 are the result of clay deposits that
have been heavily preloaded during their geologic history. The transition from very
dense soil to soft rock is not well-defined. In some cases, very dense clays that have
not fully lithified may be included in this category.
Rock properties are closely tied to the properties of the minerals from which it is
comprised. The rocks in this category contain significant amounts of clay minerals with
properties ranging from the non-swelling kaolinite group to the highly swelling
montmorillonite group. Soil and rock that has a high fraction of clay minerals will tend to
expand, absorb water, and lose shear strength while undergoing stress relief. Thus,
rocks with significant amounts of clay minerals will tend to swell as a result of the stress
relief around an underground opening. Swelling leads to a loss of shear strength and a
tendency of the tunnel walls to squeeze into the opening.
4-5.3 Loads on Underground Openings in Rock.
4-5.3.1 Vertical Rock Load.
A common starting point for the estimation of vertical roof pressure is found in Table
4-6. It should be noted that the values presented in this table were largely based on
observations by Terzaghi (1946) for tunnel widths in the range of 16 to 32 feet (5 to 10
meters) prior to the advent of “modern” tunneling methods. Table 4-6 provides an
approximate means of calculating the vertical pressure or rock load that must be
supported by roof lining. The height of rock ( H p ) that must be supported is a function of
the tunnel width ( B ) for high quality rock and also the tunnel height ( H t ) for lower
quality rock. The total vertical pressure can be found by multiplying H p by the total unit
weight of the rock.
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Table 4-6 Approximate Overburden Rock Load Carried by Roof Support
Rock Load H p
Rock Condition Remarks
(same units as B )
Sometimes spalling or
1. Hard and intact 0
popping occurs.
2. Hard stratified or schistose 0 to 0.5 B Light pressures.
Load may change erratically
3. Massive, moderately jointed 0 to 0.25 B
from point to point.
4. Moderately blocky and seamy 0.25 B to 0.35 ( B + Ht ) No side pressure.
5. Very blocky and seamy 0.35 to 1.10 ( B + H t ) Little or no side pressure.

Considerable side pressure.
6. Completely crushed, chemically intact 1.10 ( B + Ht ) Softening effect of seepage
towards bottom of tunnel.
7. Squeezing rock, moderate depth (1.10 to 2.10) ( B + Ht ) Heavy side pressure.
8. Squeezing rock, great depth (2.10 to 4.50) ( B + Ht ) Heavy side pressure.

Up to 250 ft, not related to
9. Swelling rock Very heavy pressure.
value of ( B + Ht )
Notes:
1. After Proctor and White (1977) based on observations by Terzaghi (1946).

2. Rock loads apply to tunnels at depth greater than 1.5(B + Ht).
3. The roof of the tunnel is assumed to be located below the water table. If the tunnel is located permanently
above the water table, the values given for Conditions 4 to 6 can be reduced by fifty percent.
4. Some very dense clays which have not yet acquired properties of shale rock may behave as squeezing
or swelling rock.
5. Where sandstone or limestone contain horizontal layers of immature shale, roof pressures will
correspond to rock condition “very block and seamy.”
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Table 4-7 Approximate Relationship Between Rock Quality Indices

(after Deere et al. 1970, Barton et al. 1974, Bieniawski 1990, Hemphill 2012)
Rock Tunneling Rock Mass
Rock Condition RQD
Quality Index, Q A Rating, RMR B
1. Hard and intact 95 to 100 ≥ 200 > 80
2. Hard stratified or schistose 90 to 99 25 to 50 65 to 75
3. Massive, moderately jointed 85 to 95 10 to 20 60 to 65
4. Moderately blocky and seamy 75 to 85 2 to 6 50 to 60
5. Very blocky and seamy 30 to 75 0.4 to 1 40 to 50
6. Completely crushed but chemically intact 0 to 30 0.04 to 0.08 20 to 25
7. Squeezing rock, moderate depth NA 0.01 to 0.03 10 to 20
8. Squeezing rock, great depth NA 0.001 to 0.004 <5
9. Swelling rock NA 0.001 to 0.003 0

A After Barton et al. (1974)
B After Bieniawski (1990)
4-5.3.2 Horizontal Pressures.
Horizontal pressures acting on a tunnel can be approximated using an active wedge

type analysis, such as Coulomb, with the diagram shown at the bottom of Table 4-6.
The vertical forces included in these calculations are the weight of the active wedge and
the weight of the rock for a height ( H p ) above the wedge. Shear strength parameters
can be assumed or selected using guidance in Chapter 3. The critical inclination of the
active wedge may be determined by either the shear strength of the rock or by the rock
structure. The possibility of movement along a weak bedding plane or stratum should
be considered.
4-5.3.3 Other Methods for Tunnel Support Pressures.
Alternate empirical methods for estimating tunnel support pressures are available. The
two most common are the rock tunneling quality index, Q , presented in Barton et al.
(1974) and the rock mass rating system (e.g., Bieniawski 1976). These systems are
described in more detail in Chapter 1. The roof pressure and support requirements for
tunnels can be estimated from the value of Q and some of the joint properties. While
not solely intended for tunneling applications, the RMR can be related to stand-up time,
unsupported active span length, and roof pressure. See Bieniawski (1990) for
additional comparison of methods for estimating tunnel support requirements.
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4-5.4 Openings in Soft Ground (Soil).
4-5.4.1 Ground Behavior.
Selection of an appropriate method for tunnel construction in soft ground (i.e., soil
tunneling) depends upon the response of the soil during and after excavation. This
response is often referred to in terms of stand-up time, which is the amount of time the
soil will support itself prior to the installation of tunnel supports. The stand-up time
depends on the type of soil, the position of groundwater, and the size of the opening.
Terzaghi’s (1950) Tunnelman’s Ground Classification provides a commonly used
means of describing various types of ground response. The types of ground behavior
for tunneling are summarized in Table 4-8.
Table 4-8 Types of Ground Behavior

Type of
Applicable Soil Types Support / Ground Behavior Comments
Ground
Loess above water; hard

No initial support required.
clay; marl; lightly stressed
Firm Construction final lining before None perceptible
cemented sand and
movement occurs.
gravel
Chunks or flakes of soil fall as a

Sand with clay binder Stand-up time decreases with
Raveling result of loosening, overstress,
(slow above water table, size of opening. Raveling
(slow to or brittle fracture. Time to start
fast below); stiff fissured ground can become running
fast) of raveling may be a few minutes
clays ground if the water table rises.
(fast) or more (slow)
Stand-up time is zero. In moist

Clean, dry coarse-grained Support should be prior to
soils, suction may allow soil to
soils unable to stand at an excavation. Removal of side
Running stand briefly before running.
angle greater than angle supports results in inflow of
This is referred to as “cohesive-
of repose. material.
running.”
Silt, sand, and gravel Without support, material flows

without clayey fines below into opening from all sides like a
Flowing Material acts like a thick liquid.
water table; disturbed viscous fluid. If unchecked, may
highly sensitive clay completely fill the tunnel.
Very soft to medium stiff Stand-up time adequate.

clay at shallow depths; Soil moves into tunnel gradually Behavior results from plastic
Squeezing stiff to hard clay at great without indication of rupture or flow caused by overstress.
depth; soil with low change in water content. Rate of advance is related to
frictional strength the degree of overstress.
Soil absorbs water over time,

High OCR clays with increases in volume, and Advances into opening occur
Swelling swelling minerals and PI expands toward the tunnel. due to an increase in volume
greater than about 30. Pressure on support members allowed by stress relief.
may increase with time.
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In addition to the descriptions provided in Table 4-8, the behavior of fine-grained soils
and silty sands above the water table can be evaluated using Table 4-9. The undrained
stability factor ( N crit ) is used to assess the ground behavior as suggested by Peck
(1969). This factor is defined as:
σ 'v − σ t
N crit = (4-10)
su
where:
σ 'v = effective overburden pressure at the tunnel centerline,
σ t = interior applied pressure from compressed air or breasting, and
su = undrained shear strength.
Table 4-9 Ground Behavior for Clayey Fine-Grained Soils and Silty Sand
(after FHWA 2009)
Soil Type Stability Factor, N crit Ground Behavior
1 Stable
2 to 3 Small amount of creep

Clayey Fine-
Grained 4 to 5 Creeping, usually slow enough to permit tunneling
May experience general shear failure. Clay likely

6
to invade tail space too quickly to handle.
0.25 to 0.33 Firm

Silty Sand
above Water 0.33 to 0.5 Slow raveling
Table
0.5 to 1 Raveling
For coarse-grained soils, the ground behavior depends on the grain-size distribution,
relative density, and the amount of clayey fines (binder) as indicated in Table 4-10.
Uniform, loose materials ( Cu < 3 and N < 10 ) with round grains run more freely than
well-graded, dense materials ( Cu > 6 and N > 30 ) with angular particles. Soils with
properties between those listed in Table 4-10 will tend to exhibit intermediate ground
behavior. Very high flowrates should be expected in tunnels below the water table
through soils with relatively large particles, such as gravel and medium to coarse sand.
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Table 4-10 Ground Behavior for Coarse-Grained Soils (after FHWA 2009)
Typical Ground Behavior

Relative Density (SPT
Soil Description
blow count)
Above Water Table Below Water Table
Loose ( N < 10 ) Cohesive running Flowing

Very fine clean sand
Dense ( N > 30 ) Fast raveling Flowing
Loose ( N < 10 ) Rapid raveling Flowing

Fine sand with clay binder
Dense ( N > 30 ) Firm or slowly raveling Slowly raveling
Loose ( N < 10 ) Rapid raveling Rapid raveling or flowing

Sand or sandy gravel with
clay binder
Dense ( N > 30 ) Firm Firm or slow raveling
Sandy gravel and medium

Any Running Flowing
to coarse sand
4-5.4.2 Soft Ground Support Loads.
Support pressures on tunnels in soft ground are governed by many factors, including
the unit weight of the overlying material, the groundwater level, soil properties, the
amount deformation allowed during excavation, the interaction between soil and the
supports, the opening shape, and the length of time between excavation and lining
installation. Other factors should also be considered include the presence of other
nearby openings, superimposed loads from neighboring structures, and the possibility of
changes in groundwater conditions.
Figure 4-11 illustrates the loading mechanism surrounding soft ground tunnels. In
coarse-grained soils, arching occurs above the tunnel. Arching transfers some of the
overburden load to the surrounding ground so that only a portion of the total load above
the tunnel is applied to the tunnel. In clay soils, undrained conditions tend to control. In
this case, the undrained shear strength can be considered to provide support of a
portion of the load above the tunnel.
A simplified approach to the selection of tunnel support loads is provided in Table 4-11
(FHWA 2009). More detailed guidance for the selection of tunnel support loads is
summarized in Table 4-12. These tables are used by first determining the ground
behavior type using Table 4-8 to Table 4-10.
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Figure 4-11 Loading Mechanisms for Soft Ground Tunneling
4-5.4.3 Loss of Ground.
As underground excavation is made, the surrounding ground starts to move toward the
opening. These displacements occur as the soil around the opening expands due to
stress release in addition to soil lost to the tunnel from raveling, runs, flows, etc. The
resulting loss of ground causes settlement of the ground surface. The loss of ground
associated with stress reduction can be predicted reasonably well, but the ground loss
due to raveling, runs, flows, etc. requires a detailed knowledge of the subsurface
conditions to avoid unacceptable amounts of settlement. A summary of methods to
predict surface settlement resulting from lost ground can be found in FHWA (2009).
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Table 4-11 Simplified Tunnel Support Loads based on Ground Behavior

(FHWA 2009)
Design Load Thickness, H p

Ground Behavior
Circular TunnelA Horseshoe TunnelA
H 
H
Running ground min  min  , See Note B
B 2 B

H H

Flowing ground in air free min  min  , See Note C
2 B 4 B

Raveling ground above water H H
min  min  , See Notes B and C
table B 2 B
Raveling ground below water H 
H
min  min  , See Note C
table 2 B 4 B

Squeezing ground Depth to tunnel springline
Swelling ground Same as raveling ground

A
B is the tunnel width
B Floor is required in a horseshoe tunnel if compressed air is used, otherwise ignore
compressed air.
C Stiff floor required in horseshoe tunnel
4-5.5 Pressure on Vertical Shafts.
In contrast to the methods presented in Section 4-3, the stress calculations for vertical
shafts represent either active or passive earth pressure. These limiting earth pressure
conditions correspond to a plastic rather than elastic state of stress within the soil.
4-5.5.1 Shafts in Coarse-Grained Soil.
During excavation of a vertical cylindrical shaft in coarse-grained soil, the horizontal

pressures around the shaft approach active earth pressure values. If outward-directed
forces from a structure (e.g., buried silo) move the structure walls into the soil, the earth
pressures will approach passive conditions. Earth pressures are discussed in more
detail in DM 7.2.
Active earth pressures for cylindrical shafts have been determined using analytical, limit
equilibrium, slip line, numerical, and experimental methods. The active earth pressure
coefficient depends on the shaft dimensions and the soil strength. For shallow shafts
(i.e., depth ≤ two diameters), theoretical solutions tend to be applicable while the effects
of horizontal arching become significant at greater depths (Tobar and Meguid 2010).
Horizontal arching is taken into account in some solutions by the coefficient λ , which is
the ratio of the circumferential stress to the vertical stress. The value of λ is equal to 1
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in analytical solutions, such as Terzaghi (1943) but may be as low as K0 . Cheng et

al.’s (2007) solution is plotted in Figure 4-12 for λ = 1 and λ = K0 .
Active pressures must be modified if rigid bracing at the top of the shaft prevents
development of an active state. For restrained vertical shafts, horizontal pressures may
be as large as the at-rest pressure on a long wall with plane strain conditions.
Table 4-12 Soft Ground Tunnel Support Loads for H > 1.5( B + H t )
Type of
Tunnel Conditions Design Load, H p
Ground
Loose: 0.5 ( B + H t )
Running Above water Medium: 0.4 ( B + H t )
Dense: 0.3 ( B + H t )
Disregard air pressure; H p equal to that for running ground, above water
Running Compressed air
table with equal density

 H
Flowing Free air min 

 2( B + Ht )
Above water Multiply H p for running ground by T −t

T
Below water,
Raveling Multiply H p for running ground by T −t
free air T
Below water,  T −t  pc
Using H p for running ground: 2 ⋅  H −
compressed air  T  p γt
p Hsu
Homogenous clay H− c −
γ t γ t ( B+2 Ht )
pc Hsu
SqueezingA Soft roof, stiff sides H − −
γt γtB
pc Hsu
Stiff roof, soft sides H − −
γt γ t ( B +6 H t )
Intact clay Very small
SwellingB
Fissured clay Use H p for raveling ground with same standup time
pc = tunnel air pressure, su = undrained shear strength, γ t = total unit weight of soil
t = stand up time, T = elapsed time between excavation and completion of permanent structure
Variables: H = vertical distance between ground surface and tunnel roof,
H p = design load in terms of depth (multiply by γt to determine design pressure),
H t = height of the tunnel, and B = width of the tunnel
A After complete blowout, pc = 0
B Permanent roof support should be completed within a few days after excavation
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Figure 4-12 Radial Stress at the Sides of a Vertical Shaft in Sand

(based on Cheng et al. 2007)
4-5.5.2 Shafts in Clay.
No support is needed from the ground surface to a depth of zcrit for shafts in clay. The
critical depth ( zcrit ) is:
2su
zcrit = (4-11)
γt
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where:
su = the undrained shear strength of the soil, and
γ t = the total unit weight of the soil.
At greater depths, the ultimate horizontal pressure ( σ h ) on a shaft lining in soft clay can
be estimated as:
σ h = γ '⋅ z − su (4-12)
where:
γ ' = effective unit weight of the soil,
z = depth below the ground surface, and
su = undrained shear strength of the soil.
This pressure will likely occur after several months of unsupported excavation. The
stability factors for fine-grained soils in Table 4-9 can be used as guidelines for the
behavior of vertical shafts in clay.
In stiff, intact or fissured clays, the initial horizontal stress on vertical shaft walls will be
small. Over time the pressure may increase to a value several times larger than the
effective vertical stress (and ultimately to the swelling pressure if the shaft lining is
sufficiently rigid). Local experience is important to provide useful information for soil
pressures on vertical shafts in stiff clays.
4-6 NUMERICAL SOLUTIONS FOR STRESSES IN SOIL.
The analytical and chart-based solutions presented in this chapter are an excellent
starting point for evaluation of stresses within a soil mass. However, their applicability is
limited by the constraints and assumptions of each. These methods also struggle to
effectively model complex subsurface profiles and loading conditions.
Computer programs and numerical solutions are an important part of geotechnical

engineering practice. This section provides a brief overview of the application of
numerical methods to the evaluation of stresses.
4-6.1 Numerical Analysis Types.
Some computer programs (e.g., Settle3D, CONSOL, SETOFF) are available that
directly rely upon elastic solutions, such as those presented in this chapter, to calculate
changes in stress. These programs are specifically formulated for the solution of
consolidation settlement problems discussed in the next chapter. The benefit provided
by these programs is the automated calculation of changes in stress and the ability to
consider time-dependent changes. However, the solutions depend on the same
assumptions as the elastic solutions on which they are based.
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Many continuum-based numerical analysis techniques are available. The most

common of these in geotechnical engineering are the finite element and finite difference
methods, with the former being somewhat more popular for stress analysis problems.
These approaches divide the soil or rock into elements or grid points. The relationships
between the external and gravitational forces acting on the elements (or grid) and the
corresponding displacements are defined using constitutive (stress-strain) laws and
failure criteria.
The following sections provide a condensed overview focused on the calculation of

stresses using the finite element method. The use of finite element analysis (FEA) to
determine stresses is relatively straight-forward. Calculation of accurate deformations
requires significantly more expertise and experience. More detail on numerical methods
can be found in DM 7.3 (NAVFAC 1983). For a practical introduction to the use of FEA
in geotechnical engineering see Bradley and VandenBerge (2015). A more in-depth
perspective can be found in the books by Potts and Zdravkovic (1999, 2001).
4-6.2 Linear Elastic Stress Analysis.
The simplest FE analyses use linear elastic constitutive theory, which relates changes
in stress linearly to strains (or displacements). For a linear elastic analysis, the only
required material parameters are the elastic modulus, E , and Poisson’s ratio, ν (or any
other two elastic constants such as shear modulus or bulk modulus). For problems with
only one material type (and one value of E ), the calculated stresses will be independent
of the value of E selected. However, when more than one material type is present, the
relative values of E assigned to each material in the analysis can impact the calculated
stresses because of arching and similar phenomena.
4-6.3 Nonlinear Elastic Stress Analysis.
The stress-strain behavior of geological materials is truly linear over only a small range
of strains. A properly selected nonlinear constitutive model will provide a more accurate
prediction of behavior but will require additional input parameters and expertise.
VandenBerge et al. (2014) found that major principal effective stresses calculated for
embankments with linear elastic analysis were typically within 10% of the values
calculated using more rigorous and time-consuming nonlinear procedures.
One of the earliest and most common nonlinear constitutive theories for soil is the
hyperbolic model proposed by Duncan and Chang (1970) and described in more detail
in Duncan et al. (1980). This model can also consider stress dependent variations in
Poisson’s ratio (or bulk modulus). It can be used with either effective stress (drained) or
total stress (undrained) problems, provided the model parameters are determined using
the appropriate type of test.
The Duncan-Chang model is based on the following observations;
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• The principal stress difference ( σ 1 − σ 3 ) tends to vary in a hyperbolic manner with

strain,
• The initial modulus ( Ei ) increases with increased confining stress in a manner
that can be described by a power function,
• The ultimate value of ( σ 1 − σ 3 ) predicted by the hyperbola tends to be greater
than the value measured by the test,
• The bulk modulus ( Bt ) of soil increases with increased confining stress in a
manner that can be described by a power function.
Soil tends to respond in a stiffer, approximately linear manner with modulus ( Eur ) when
unloaded and reloaded after some amount of shearing. The required parameters are
summarized and illustrated in Table 4-13 and Figure 4-13.
Table 4-13 Summary of Model Parameters for Duncan-Chang Model

Duncan-Chang
Meaning / Use Equation
Parameter
K Controls rate of increase of Ei with σ '3 n
σ ' 
Ei = KPa  3 
n Controls nonlinearity of Ei relationship  Pa 
Reduces ( σ1 −σ 3 ) from its ultimate hyperbolic value (σ1 −σ 3 )max

Rf Rf =
to match maximum value measured by testing (σ1 −σ 3 )ult
Kb Controls rate of increase in Bt with σ '3 m
σ ' 
Bt = Kb Pa  3 
m Controls nonlinearity of Bt relationship  Pa 
n
σ ' 
K ur Controls rate of increase of Eur with σ '3 Eur = Kur Pa  3 
 Pa 
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Figure 4-13 Parameters Used in Duncan-Chang Model
4-6.4 Numerical Modeling Best Practice.
Both linear and nonlinear finite element analysis can be used to make accurate
predictions of stresses. The following guidelines have been found to yield the most
consistent and uniform stress predictions (VandenBerge et al. 2014).
• Use isoparametric elements (i.e., 8-noded rectangles or 6-node triangles).

• Use a uniform mesh with elements of approximately the same shape, especially
within any zones of interest.
• Use elements with an appropriate aspect ratio, preferably longest to shortest
dimension, less than or equal to five.
• Keep the element size as small as practical with respect to the overall
dimensions. A maximum element height of approximately 1 to 2% of the height
of the problem domain is preferred.
• Remember to include the boundary water pressures from any impounded water
present in the model.
4-6.4.1 Initial or Geostatic Stresses.
The initial stress state in a FE model is dependent on the process used to “turn on”
gravity and stress within the model. Vertical stresses are governed mostly by gravity
loading. As the mesh deforms in response to gravity, horizontal stresses will develop
and tend toward at rest ( K 0 ) conditions for a level, laterally-constrained mesh. For a
linear elastic model, the value of K 0 will be equal to ν (1 −ν ) . Calculation of initial
horizontal stresses in this manner will yield correct results, and K 0 will be less than 1.0,
which is always the case for primary loading. The calculation of initial horizontal
stresses in this way will lead to initial deformation of the model and the layer
thicknesses may no longer match the in situ conditions. These displacements are a
numerical artifact and should be zeroed or removed prior to examining the effects of
new loading conditions. Overconsolidated conditions with K 0 greater than 1.0 can be
modelled by loading and then unloading the model, following the process by which
K 0 > 1.0 conditions occur in nature.
In many cases, the details of the initial stress process are program specific. The
engineer using FEA should become familiar with the various options available so that
appropriate methods are applied.
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4-6.4.2 Staged Construction or Stress History.
Finite element analysis allows the sequence of loading and construction staging to be
modeled numerically. To a lesser extent, geologic stress history can also be modeled.
Within the model, this occurs by adding and subtracting elements (or their weight) in
steps. Staged models are especially important for nonlinear analysis because the
properties of the material change with stress level.
For example, a staged approach is required to predict stresses and displacements

around open excavations and tunnels. The initial ground conditions should be modeled
after which the effects of removing soil or rock can be considered.
For cases where only stress distributions are required, such staging is unnecessary in
linear elastic analysis. The use of a staged model is required, even for linear elasticity,
to predict correct patterns displacement.
4-6.5 Evaluation of Stress Due to Applied Loads.
Changes in stress due to applied loads can be evaluated using a staged FE model.
The first stage(s) of the model are used to create the desired initial state of stress that
best represents the in situ conditions. At this point, the new loading can be added to the
model in various forms, including distributed loads, point loads, and new soil layers.
The predicted changes in stress can then be evaluated by comparing the predicted
stress at convenient points within the model between subsequent stages. Where
possible, the changes in stress predicted by FE models should be checked with
analytical solutions such as those presented earlier in this chapter.
Engineers should be aware of the limitations of their numerical analyses. For example,
two-dimensional FEA are useful for predicting changes in stress below long
foundations, embankments, and large area fills because such problems can be
analyzed in a plane strain manner. However, a three-dimensional program would be
required to predict changes in stress below more complex conditions, such as a
rectangular foundation.
4-6.6 Evaluation of Stress within Embankments and Slopes.
The calculation of stresses is more complex for slopes and embankments compared
with relatively level ground. First, no closed-form analytical solution is available for
comparison because the soil in a slope is not laterally restrained. The lack of lateral
restraint also means that K 0 conditions will not be present, especially close to the face
of a slope.
Stresses near natural slopes are best evaluated by starting with level ground in the FE
model and progressively removing (excavating) elements to form the slope. Multiple
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stages of analysis may be required to establish the initial conditions. The removal of
elements mimics the process by which the slope was formed in nature.
Likewise, stresses in an embankment can be modeled by adding layers of elements in

stages. The fill zone can be “built” within the FE model in thin layers, mimicking the
actual construction process. Deformations caused by the initial application of gravity
loading should preferably be removed. However subsequent deformations of each
layer of fill under the weight of the overlying material are realistic. This approach allows
the true pattern of displacements within the embankment to be examined.
Topic Reference
Parry, R. H. 2004. Mohr circles, stress paths and geotechnics, CRC Press,
Stress and Mohr Circles
2004.
Poulos, H. G. and Davis, E. H. 1974. Elastic Solutions for Soil and Rock
Elastic Solutions
Mechanics, John Wiley & Sons Inc.
Moser, A. P. 1990. Buried Pipe Design, McGraw-Hill Inc., (third edition also
Stress on Pipes
available by Moser and Folkman).
FHWA. 2009. Technical Manual for Design and Construction of Road

Underground Openings Tunnels – Civil Elements, FHWA-NHI-09-010, Federal Highway
Administration.
Bradley, N. and VandenBerge, D. R. 2015. Beginner’s Guide for
Geotechnical Finite Element Analyses, CGPR Report No. 82, Center for
Geotechnical Practice and Research, Virginia Tech.
Numerical Stress Analysis
Potts, D. M. and Zdravkovic, L. 2001. Finite Element Analysis in
Geotechnical Engineering: Theory and Application, ICE Publishing.
4-8 NOTATION.
Symbol Description
B Width of a foundation, loaded area, or tunnel
Bc Diameter of a flexible pipe
Bd Width of trench in pipe loading calculations
Bt Bulk modulus of soil
Cd Load coefficient in pipe loading calculations
Cu Coefficient of uniformity (from grain-size distribution)
D Outer diameter of pipe
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Symbol Description
E Elastic modulus
EA Active earth pressure force
Ei Initial tangent modulus
Eur Unload-reload modulus for Duncan-Chang model
hp Pressure head
H Depth of soil cover or vertical distance between ground surface and tunnel roof
Hp Design load thickness in terms of depth
Ht Tunnel height
I Influence factor for change in stress calculations
Ka Active earth pressure coefficient
Kb Bulk modulus parameter for Duncan-Chang model
K0 At-rest earth pressure coefficient
KP Passive earth pressure coefficient
K ur Unload-reload modulus parameter Duncan-Chang model
N Standard Penetration Test blow count
N crit Undrained stability factor
pc Tunnel air pressure
p 'f Center of Mohr circle at failure (MIT stress path space)
Pa Atmospheric pressure
u Pore water pressure
qf Radius of Mohr circle at failure
q0 Applied pressure or load
Q Rock tunneling quality index
r Horizontal distance from centerline of a foundation
Rf Reduction factor for Duncan-Chang model
RQD Rock quality designation
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Symbol Description
t Stand up time for tunneling in raveling soils
T Elapsed time between excavation and completion of permanent structure
Wc Flexible pipe load
Wd Rigid pipe load
Wp Prism load on pipe
z Depth below an applied load
zcrit Critical depth for unsupported shafts in clay soils
Zi Soil layer thickness
α Angle between the major principal plane and the plane of interest
∆σ z Change in vertical stress
φ' Effective stress friction angle
γ' Effective unit weight
γt Total unit weight
λ Ratio of the circumferential stress to the vertical stress
µ' Coefficient of friction for trench backfill
ν Poisson’s ratio
σ1 Major principal stress
σ2 Intermediate principal stress
σ3 Minor principal stress
σh Total horizontal stress
σ 'h Effective horizontal stress
σt Interior tunnel pressure from compressed air or breasting
σv Total vertical stress
σ 'v Effective vertical stress
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ANALYSIS OF SETTLEMENT AND VOLUME EXPANSION
5-1 INTRODUCTION.
5-1.1 Scope.
This chapter explains the practical aspects of the process of volume change in soil.
Many of the cases relate to the compression of soil layers due to changed conditions,
such as placement of an engineered fill, foundation loading, or lowering of the
groundwater table. Compression of soil results is settlement, which is vertical
displacement of the ground surface or a structure supported by the ground. Both
immediate and long-term settlement will be considered, along with tolerable settlement
criteria, the rate of settlement, and methods to reduce or accelerate settlement.
Swelling soils can also change volume by expansion, which is often referred to as
heave.
Guidance on special cases, such as unsaturated soils, collapsing soils, and sanitary
landfills, is provided in DM 7.2 (NAVFAC 1982) and DM 7.3 (NAVFAC 1983). Chapter 2
provides guidance on methods for monitoring settlement.
5-1.2 Occurrence of Settlement.
Settlement is the result of three primary mechanisms: (1) immediate distortion that
occurs in response to the application of a new load, (2) consolidation, which is
compression of the soil skeleton in response to changes in effective stress, and (3)
secondary compression, which is rearrangement of the soil structure under constant
effective stress. All three mechanisms will be explained in more detail in later sections
of this chapter.
In saturated fine-grained soils, the settlement associated with the three mechanisms
can be distinguished and separated for the practical purpose of estimating settlement
magnitude. This separation is possible because these soils have low hydraulic
conductivity and relatively high compressibility, which causes consolidation to occur
over a measurable period of time. The processes and magnitudes of settlement
typically associated with fine-grained soils are summarized in Table 5-1.
Coarse-grained soils are much less compressible than fine-grained soils and have
higher hydraulic conductivity. Because of these characteristics, consolidation occurs
very quickly and can be difficult to separate from immediate deformation. Much of the
compression of coarse-grained soil is related to particle rearrangement under changed
stress. Vibrations from earthquakes, blasting, or machinery can also cause settlement
of coarse-grained soil. Submergence and soaking of coarse-grained soils, particularly
fill materials, can lead to settlement as discussed in Section 5-9.
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5-1.3 Occurrence of Heave.
Heave or swell occurs primarily due to the reduction of total vertical stress or the
reduction of matric suction, which both lower the effective stress on the soil and
therefore allow an increase in the void ratio. The reduction of total stress occurs as the
result of excavation or erosion of soil, as well as the removal of man-made loading. The
matric suction in a soil depends on the soil type and the past and present atmospheric
conditions. Especially important is the degree of saturation of the soil. The factors
affecting the degree of saturation include climate history, topography, vegetation, and
groundwater level. The amount of swelling that occurs as a result of either mechanism
depends on the size and type of minerals that comprise the soil. Clay minerals with
very small particles and high specific surface area, such as montmorillonite and
vermiculite, are most susceptible to swelling. Heave can also occur as a result of the
formation of ice lenses in frozen soil. Volumetric expansion from chemical reactions,
such as pyrite, can also cause heave.
5-1.4 Applicability.
The methods to analyze settlement that are presented in this chapter apply to
conditions where shear stresses are well below the shear strength. In addition, these
analyses of consolidation magnitude and rate as well as secondary compression
assume that the soil is saturated.
Table 5-1 Settlement Calculation Methods for Different Soil Types

(after Coduto et al. 2011, Salgado 2008)
Soil
Time Frame Process Relative Magnitude Method of Calculation
Type
Distortion Negligible to small

Semi-empirical immediate or
Short-Term
“elastic” settlement
Consolidation Small to moderate
Coarse-
Grained Secondary
Long-Term Negligible to small Semi-empirical methods
Compression
Vibration,
Other Small to moderate Specialized methods
submergence
Semi-empirical immediate or
Short-Term Distortion Negligible to small
“elastic” settlement
Fine- Primary consolidation

Consolidation Moderate to large
Grained calculations
Long-Term
Secondary Secondary consolidation
Small to large
Compression calculations
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5-2 MECHANICS OF CONSOLIDATION.
5-2.1 Consolidation Process.
Consolidation of soil is caused by changes in effective stress. For this reason, it is

necessary to consider the process by which effective stresses change within a soil
mass. Saturated soil consists of two relatively incompressible components, the mineral
particles and water. In comparison, the overall soil structure is compressible because
the particles can rearrange to encompass more or less void space. The compressible
soil structure must strain or deform in order to support a change in stress.
When a soil mass experiences an abrupt change in the applied total normal stress, the
soil must respond in one of two ways: (1) the compressible soil structure strains
instantaneously, or (2) a pressure change occurs in the incompressible water within the
soil voids. The first option is not possible because strain of the soil structure requires a
change in the void ratio of the soil. The void ratio of a saturated soil can only change if
the amount of water in the soil changes, and time is required for water to leave or enter
the soil voids. Instead, the instantaneous soil response follows the second option, and a
change in the pore water pressure occurs. This temporarily altered pressure is often
referred to as excess pore water pressure, u x . For saturated, one-dimensional
conditions, the magnitude of u x will be roughly equal to the change in total vertical
stress, ∆σ z . Because ∆σ z is balanced by u x , the instantaneous change in effective
stress is negligible, and there is no instantaneous settlement as a result of
consolidation.
The excess pore water pressure in the soil creates a hydraulic gradient between the
conditions within the soil and those at its boundaries. The gradient causes water to flow
out of or into the soil, as time progresses following the application of ∆σ z . When ∆σ z is
positive, the soil volume decreases as the water flows out and the excess pore water
pressures decrease. After a relatively long period of time, the excess pore water
pressure dissipates completely (i.e., u x = 0), all of the change in total vertical stress is
transferred to the soil structure, and the consolidation settlement is complete. The time
required to reach this state can range from seconds or minutes for thin layers of coarse-
grained soils to years for thick layers of fine-grained soil.
Estimates of the magnitude of consolidation require (1) knowledge of the initial vertical
stress state, (2) prediction of the change in total stress caused by new loading, (3) an
understanding of the stress history of the soils impacted by the changes, and (4)
knowledge of the compressibility characteristics of those soils.
Estimates of the rate of consolidation settlement require knowledge of (1) the hydraulic
boundary conditions, (2) the thickness, (3) the compressibility, and (4) the hydraulic
conductivity of the compressible soil layers.
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5-2.2 Initial Vertical Stress State.
Consolidation analysis focuses on volume change caused by changes in vertical stress.

The initial geostatic vertical total stress ( σ z 0 ), pore water pressure ( u0 ), and vertical
effective stress ( σ 'z 0 ) should be calculated using methods discussed in Chapter 4.
Figure 5-1(a) illustrates initial vertical stress conditions for hydrostatic water conditions.
Artesian water conditions are associated with confined aquifers in which higher pore
water pressure is present in a more permeable soil layer below a confining stratum.
Artesian conditions affect the calculation of pore water pressure within the compressible
layer as well as the calculated vertical effective stress. As illustrated in Figure 5-1(b),
the hydrostatic pore water pressure is labeled u0 while the artesian pressure at any
depth is ∆u . The artesian pressure must also be subtracted from the total vertical
stress in order to obtain the correct initial effective stress.
Figure 5-1 Initial Vertical Stresses for a) Hydrostatic and b) Artesian Pore Water
Pressure Conditions
5-2.3 Stress History.
The stress history of a soil refers to the past stress states that the soil has experienced.
It will affect the structure and behavior of the soil under new loading. This is especially
true for fine-grained soils. Of particular importance is the highest vertical effective
stress to which the soil has been consolidated, which is known as the preconsolidation
stress or maximum past pressure, σ ' p . A method for determining the preconsolidation
stress is shown in Figure 3-12.
For some soil deposits, the existing vertical effective stress is the highest vertical
effective stress the soil has ever experienced and σ ' p is equal to σ 'z 0 . This type of soil
deposit is referred to as normally consolidated and sometimes abbreviated “NC.”
More commonly, soil deposits have been preloaded or preconsolidated at some point in
the past, and σ 'z 0 is less than σ ' p . This type of soil is referred to as overconsolidated
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and sometimes abbreviated “OC.” The overconsolidation ratio ( OCR ) is a helpful

measure of soil behavior and is found as:
σ 'p
OCR = (5-1)
σ 'z 0
where:
σ ' p = preconsolidation stress and
σ 'z 0 = current vertical effective stress.
In situ vertical stress profiles for steady state conditions are summarized in Figure 5-2.
In Figure 5-2(a), the clay layer is normally consolidated because the preconsolidation
stress is equal to the effective vertical stress. The clay in Figure 5-2(b) is slightly
overconsolidated as a result of a higher groundwater level compared to (a), which
reduces the current vertical effective stress. Similarly, overconsolidation as a result of
excavation (or erosion) and previous loading are depicted in Figure 5-2(c) and (d).
Figure 5-2 Vertical Stress History Examples
In some cases, a soil layer may be encountered that has not yet finished consolidating
as a result of a prior change in effective stress. This state is referred to as
underconsolidated. In this case, the pore water pressures in the soil have not yet
reached equilibrium following a change in stress and are higher than the hydrostatic
values. Two examples are provided in Figure 5-3. In the first example, the groundwater
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level has been lowered in the sand layers. The decrease in pore pressure (and
corresponding consolidation) of the low permeability clay layer will be time-dependent.
The second example is partial consolidation under a prior applied stress. In either case,
the initial pore pressure variation with depth within the clay layer must be measured or
estimated in order to calculate the variation of initial vertical effective stress within the
soil.
Figure 5-3 Vertical Stress Profile Cases – Transient
5-2.4 Evaluation of Existing Conditions.
For purposes of settlement calculations, the existing conditions at the start of

construction or the application of a new load must be evaluated. At a minimum, this
evaluation should include the following steps:
• Review the available site and geologic data. In particular, determine potential
sources of overconsolidation (e.g., glaciation, erosion, human activity,
groundwater fluctuations) and estimate the likely magnitude of preconsolidation
and/or OCR .
• Determine the variation of the preconsolidation stress with depth from laboratory
consolidation tests (see Chapter 3). Measurements of undrained shear strength
can also be used along with correlations to provide additional estimates of
preconsolidation stress. For example, undrained strength ( su ) is often related to
the in situ vertical stress and the OCR by:
su ≈ USRNCσ 'z 0 OCR m (5-2)
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where:
USRNC = the soil’s normally consolidated undrained strength ratio, and
m = an empirical coefficient (See Section 8-3).
Equation 5-2 can be rearranged to obtain:

1/ m
 su 
σ ' p ≈ σ 'z 0   (5-3)
 USRNCσ 'z 0 
• Compare estimates of preconsolidation stress to current vertical effective stress.

A helpful tool for this purpose is a plot showing the subsurface profile, the
laboratory test data, and the variation of effective vertical stress with depth, such
as that shown in Figure 5-4.
• If underconsolidation is expected or indicated, measurements of pore water
pressure with depth are required to identify the extent of the underconsolidation.
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Figure 5-4 Example Evaluation of Existing Conditions

5-2.5 Change in Vertical Stress.
The methods presented in Chapter 4 should be used to evaluate the change in vertical
stress at the required depths within the compressible soil layer. Surcharge loads of
wide lateral extent will result in constant value of ∆σ z . Most other loading conditions
will result in ∆σ z that varies with both depth and lateral location below the applied load.
It is the responsibility of the engineer to determine the critical locations at which
settlement will be evaluated.
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5-3 SETTLEMENT CALCULATIONS.
5-3.1 Basic Formulation.
At the most basic level, settlement at the ground surface is equal to the change in
thickness of the soil underlying a load. The change in thickness ( ∆H ) divided by the
initial thickness ( H ) is equal to the vertical strain (engineering strain). Thus, settlement
( s ) is the sum of the vertical strain ( ε z ) caused by ∆σ z for each compressible soil layer
multiplied by the initial thickness of each layer, or:
n
s=∆H =∑ ε z ,i H i
i =1
(5-4)
where:
H i = thickness of each layer in same units as s .
Most settlement calculations can be split into a component related to the vertical strain
and a component related to the initial layer thickness. This concept can be used
understand the calculation procedures at a deeper level.
Many of the settlement prediction methods in this chapter use foundation geometry to
define influence factors or to select the appropriate procedure. The shortest dimension
of the foundation or loaded area will be designated as B while the longest dimension is
L . The applied stress at the base of the foundation is indicated by q0 .
5-3.2 Soil Layers in Settlement Calculations.
Calculations of distortion settlement of fine-grained soils and total settlement of coarse-

grained soils often treat the soil as one layer. In this case, the effect of the variation in
strain with depth below the load is built into the calculation procedure and influence
factors. This approach is illustrated by Figure 5-5(a).
In contrast, consolidation settlement of fine-grained soils is typically calculated by

dividing the soil into multiple layers. The vertical strain is determined for each soil layer,
which allows the effects of load geometry and changing soil conditions to be considered
explicitly. Figure 5-5(b) shows a layer of compressible soil divided into many thin layers
of equal thickness. This method is flexible and theoretically sound but requires a large
number of calculations that may be tedious if not automated. Figure 5-5(c) illustrates an
approach in which the layer thickness increases with depth. This method recognizes
that conditions change most quickly near the load. Regardless of the method used to
define soil layers, soil properties within a given layer should be constant. Actual layer
boundaries in the subsurface profile must supersede the layer division suggestions in
Figure 5-5.
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Figure 5-5 Three Possible Methods to Define Layers for Homogeneous

Conditions
5-4 SETTLEMENT OF COARSE-GRAINED SOILS.
As indicated in Table 5-1, distortion and consolidation settlement occur in coarse-

grained soils in a relatively short time span. If considered, secondary compression of
these soils is typically estimated as a proportion of the calculated short-term settlement.
For this reason, it is common practice to combine the components of settlement for
coarse-grained soils. A variety of calculation methods are available. The soil properties
for most of the methods are based on the results of field tests, such as CPT or SPT,
due to the variability and difficulty of sampling coarse-grained soils.
5-4.1 Short Term Settlement of Coarse-Grained Soil.
5-4.1.1 Elastic Method.
If the soil supporting a load is idealized as an elastic medium with a modulus equal to
Es and Poisson’s ratio of ν , the resulting settlement ( s ) is:
q0
s= ( Bµ0 µ1 ) (5-5)
Es
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where:
qo = stress applied by load,
B = width of the applied load,
µ0 = influence factor associated with embedment of the load, and
µ1 = influence factor associated with the problem geometry and Poisson’s ratio.
Some sources report Equation 5-5 with a ( 1 −ν 2 ) term. This term is often combined with
the influence factors directly. When using charts and tables for µ1 , the engineer must
check carefully to determine whether or not the ( 1 −ν 2 ) term has been included and
what value of ν has been assumed, if appropriate.
The influence factors, µ1 and µ0 , can be found using Figure 5-6 for the ratios of L B ,
H B , and D B represented by the problem geometry. Note that H B ratios can
theoretically be very high, when a significant depth of soil is present below a loaded
area. However, based on the concept of critical depth (see Section 4-2.1.5), the zone
that contributes to settlement typically has a thickness of 4B to 5B below the loaded
area. The use of H B ratios greater than 4 to 5 may overestimate settlement.
The settlement predicted by Equation 5-5 will be directly related to the value of Es ,
which is a difficult parameter to measure or obtain. General guidance for the selection
of Es for coarse-grained soils is provided in Table 5-2. Most of the correlations
summarized in this table are based on the results of Standard Penetration Test (SPT)
blow counts. An average SPT value ( N ' ) is used to predict settlement in coarse-
grained soils. In most cases, N ' is equal the average N 60 value from the bottom of the
loaded area to a depth of B below the load. In dense, saturated silty sands, the value
of N 'SM is calculated as:
15 + 0.5 ( N '− 15 )
N 'SM = (5-6)
where:
N ' = average N 60 value.
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Figure 5-6 Elastic Influence Factors for ν = 0.5 for (a) µ1 (after Giroud 1972) and
(b) µ0 (after Burland 1970)
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Table 5-2 Approximate Modulus Values for Coarse-Grained Soil (after Bowles
1996, Duncan and Mokwa 2001)
Soil Conditions Normally Consolidated Preloaded or Compacted
Loose Sand D = 40% 200 to 400 ksf 400 to 800 ksf
r
Medium Dense Sand D = 60% 300 to 500 ksf 500 to 1000 ksf
r
Dense Sand D = 80% 400 to 600 ksf 600 to 1200 ksf

r
Lower:
= E s 11.5 ( N '+ 7.5 )
Dry or Moist Sand Es 20( N '+ 42 )
=
E s 15 ( N '+30 )
Upper:=
Lower:
= Es 5.6( N '+9 )
Clayey, Silty, or Saturated Sand Not available
Upper:
= Es 7.7( N '+15 )
Note: Correlations provide values of E s in ksf units.
5-4.1.2 Schmertmann Method.
The Schmertmann Method is a common approach for the calculation of settlement for
coarse-grained soils. This method uses typical patterns of vertical strain below a rigid
foundation along with estimates of modulus based on either CPT or SPT. The variation
of the strain influence factor ( I z ) with depth is based on observations from model scale
tests as well as numerical simulations. As shown in Figure 5-7, I z increases with depth
below the loaded area up to a peak value ( I zp ) and then decreases to zero at a depth of
2B for square footings and 4B for continuous footings. The magnitude of I zp is a
function of the applied load and the effective vertical stress ( σ 'zp ) at the depth of the
peak influence factor.
The compressibility of the coarse-grained soil is incorporated through layer moduli

estimated from CPT or SPT results. The soil profile immediately below the foundation is
divided into layers with relatively constant cone tip bearing resistance, qc (or SPT blow
count). Schmertmann et al. (1978) recommend that CPT qc values should be multiplied
by 2.5 to obtain Es for axisymmetric ( L = B ) conditions. Similarly, CPT qc values should
be multiplied by 3.5 to obtain Es for plane strain ( L B > 10 ) conditions.
Schmertmann (1970) provided multipliers to estimate qc from SPT blow count, N .

Robertson and Cabal (2014) found a similar correlation between CPT and SPT. Table
5-3 combines the SPT-CPT correlation with the qc - Es correlation to provide
approximate correlation between N 60 and Es . The general values in Table 5-3 can be
replaced by regional correlations that follow the principles provided in Schmertmann
(1970) and Robertson et al. (1983).
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Table 5-3 Estimates of E s based on SPT N 60 values.
Approximate E Value (ksf)

s
Soil Type
Axisymmetric Plane Strain
(L = B) (L/B > 10)
Silt, sandy silt, slightly cohesive silt-sand mixtures 10 N 60 14 N 60
Clean fine to medium sand, and slightly silty sand 17.5 N 60 24.5 N 60
Coarse sand and sand with little gravel 25 N 60 35 N 60
Sandy gravel 30 N 60 42 N 60
The settlement ( s ) is then calculated for n layers as:
n  I 
C1C2 ( q0 − σ 'z 0 ) ∑  z ,i
s= ⋅z (5-7)

i =1  Es ,i
 i

where:
C1 = coefficient to correct for the effects of embedment,
C2 = coefficient to correct for the effects of time,
q0 = applied foundation pressure,
σ 'z 0 = the existing vertical effective stress at the bottom foundation,
I z ,i = the average strain influence factor for the layer,
Es ,i = the layer modulus, and
zi = the layer thickness.
The correction for foundation embedment is found from:
 σ 'z 0 
1 − 0.5 
C1 =  ≥ 0.5 (5-8)
 q0 − σ 'z 0 
The correction for time ( t ) in years after initial loading is:
 t 
C2 = 1 + 0.2 log   (5-9)
 0.1yr 
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Figure 5-7 Influence Diagram and Modulus Correlation for Schmertmann CPT
Method (Schmertmann 1970, Schmertmann et al. 1978)
5-4.1.3 Empirical Methods.
The difficulty of obtaining representative measures of compressibility or modulus for

coarse-grained soils has led to the development of many different empirical methods.
These methods are based on measurements and observations of load-settlement
behavior from plate load tests as well as actual foundations. The underlying basis of
these methods remains the elastic theory presented in Equation 5-5 but the soil
modulus and influence factors are replaced with empirical correlations to SPT blow
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count and foundation dimensions. Three of the more popular relationships are provided
in Table 5-4.
Table 5-4 Empirical Equations for Settlement of Coarse-Grained Soils

Method Equation Comments
s in inches
2
B in ft
Terzaghi and 6 q0  B 
Peck (1967) s=   q0 in ksf
N '  B +1 
A range of constants
have been used.
2.5q0
s=
Meyerhof (1965), ( N '−1.5)CB
Duncan and s in inches
 1.0 for B < 4ft
Buchiagnani  q0 in ksf
(1987) C B = Interpolate for 4 to 12 ft
 0.8 for B ≥ 12ft

0.75  1.31  s in inches

Burland and s=B  1.4  q0 C s
Burbridge (1985), N'  B in ft
Terzaghi et al.  1.25( L B ) 
2 L in ft
(1996)A Cs =   , C s → 1.56 for strip load q0 in ksf
 ( L B )+ 0.25 
A Settlement for load applied at the ground surface to normally consolidated sand. See the
provided references for methods to correct for the effects of embedment and
overconsolidation.
5-4.1.4 Accuracy and Reliability.
Tan and Duncan (1991) provide a helpful perspective for assessing the usefulness of
the various settlement methods for coarse-grained soils. They evaluated the accuracy
and “reliability” 11 of 12 SPT-based methods and the Schmertmann CPT Method using
more than 90 case histories. The most accurate methods will make a reliably
conservative estimate of settlement (i.e., greater than or equal to the actual value) only
about half of the time. Likewise, more reliable methods, such as Terzaghi and Peck
(1967), tend to greatly over-predict settlement, which is well-documented. Figure 5-8
can be used to select an appropriate method for determining settlement for each project
based on considerations of accuracy and reliability. In cases where more accuracy is
required, one of the methods that plots to the lower left may be used. Where it is critical
not to exceed the calculated settlement, a method with higher reliability can be used.
11In this context, reliability was defined as the percentage of cases where the measured settlement was
less than the predicted settlement. Reliability is not used in a formal probabilistic sense in this case.
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Figure 5-8 Comparison of Settlement Calculation Methods for Coarse-Grained

Soils based on SPT Blow Count (after Tan and Duncan 1991)
5-4.2 Long-Term Settlement of Coarse-Grained Soil.
Creep or secondary compression of coarse-grained soil is sometimes considered by

multiplying the calculated short-term settlement by a time-dependent influence factor.
One suggested relationship is the creep factor ( C2 ) included in the Schmertmann
Method. This factor can also be applied to the results of other coarse-grained
settlement methods. Terzaghi et al. (1996) suggest a similar approach to calculate a
creep factor ( Ct ) which is related to the magnitude of the applied stress. The resulting
values of Ct for their approach are summarized in Figure 5-9 where the applied stress (
q0 ) is normalized by atmospheric pressure, Pa . For larger loads, this method predicts
lower values of Ct because creep movements are a smaller proportion of the overall
expected settlement. The creep factors for both methods have the same mathematical
form. The total settlement at time ( t ) is found by multiplying the short-term settlement
by the value of Ct .
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Figure 5-9 Creep Factors for Settlement of Coarse-Grained Soils

( t 0 = 0.1 year and q0 in same units as Pa )
(after Schmertmann 1970, Terzaghi et al. 1996)
5-5 SETTLEMENT OF FINE-GRAINED SOILS.
5-5.1 Immediate Settlement of Fine-Grained Soils.
Immediate settlement of fine-grained soil is the result of one of two mechanisms:

(1) “elastic” compression and volume change of the unsaturated soil or (2) distortion of
saturated soil without volume change. Immediate settlement may be a significant
proportion of settlement for unsaturated or heavily overconsolidated clay.
Similar to coarse-grained soil, immediate settlement of fine-grained soil can also be

calculated using Equation 5-5 and the influence factors provided in Figure 5-6. For
saturated clay, a Poisson’s ratio of 0.5, which is the value assumed in the construction
of the figure. In most cases, the undrained modulus ( Eu ) should be used. Values of Eu
can be measured in laboratory or field tests, such as the pressuremeter. Caution
should be used as laboratory tests may underestimate the magnitude of Eu . Similarly,
field tests typically load the soil horizontally rather than vertically, which may lead to
erroneous results. Empirical correlations, such as those in Figure 5-10, can be used for
comparison or in place of test values when appropriate. OCR can be estimated from
empirical correlations or measured using one-dimensional consolidation tests as
described in Chapter 3.
If the factor of safety against bearing capacity failure is less than about 3 (see DM 7.2),
then the immediate settlement should be modified to account for partial yield of the soil.
D’Appolonia et al. (1971) can be used to determine the appropriate adjustment for this
condition.
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Figure 5-10 Correlation of Normalized Undrained Modulus and Overconsolidation

Ratio (after Duncan and Buchignani 1987)
5-5.2 Primary Consolidation Settlement of Fine-Grained Soils.
Primary consolidation occurs as water flows from saturated soil and the excess pore
water pressures caused by loading are able to dissipate. The magnitude of primary
consolidation settlement can be predicted with reasonable accuracy when the soil’s
preconsolidation stress can be determined reliably and when the change in total stress
can be accurately predicted. Settlement calculations involving recompression of
overconsolidated soils tend to have the largest percent error. The amount of error for
overconsolidated soils is related heavily to the quality of the samples used for
consolidation tests, which affects the accuracy of the predicted recompression index
and preconsolidation stress.
5-5.2.1 Use of Consolidation Test Results.
One-dimensional consolidation tests are used to predict swell, recompression, and

virgin compression of soils. Specific details about the testing process are described in
further detail in Section 3-2.6. The results of consolidation tests are typically plotted in
terms of void ratio vs. vertical effective stress or vertical strain vs. vertical effective
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stress. 12 In either case, vertical effective stress ( σ 'z ) will be plotted on a logarithmic
scale.
Ideally, the soil behaves in the manner depicted in Figure 5-11 with log-linear segments
describing the volume change or vertical strain that occurs as the vertical effective
stress changes. The initial condition corresponds to initial vertical effective stress ( σ 'z 0 )
and either the initial void ratio ( e0 ) or to zero initial vertical strain, ε z = 0. From the initial
condition, the loading of overconsolidated soil results in relatively elastic recompression
until the preconsolidation stress is reached. Any increase in σ 'z that extends beyond
σ ' p results in plastic deformation or virgin compression. The slopes of these two lines
are defined by the recompression index ( Cr ) and the compression index ( Cc ). A
normally consolidated soil ( σ 'z 0 = σ ' p ) will experience virgin compression due to any
increase in σ 'z .
Figure 5-11 Consolidation Behavior based on (a) Void Ratio and

(b) Vertical Strain
12In old soil mechanics references, the water content was often used in lieu of void ratio. For saturated
soils, the water content is equal to the void ratio divided by the specific gravity.
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Interpretation of consolidation tests in terms of vertical strain is often a more practical

approach. This method emphasizes that prediction of strain is the intent of the
calculations and will be used in much of the discussion in this chapter. The modified
recompression index ( Cε r ) and the modified compression index ( Cε c ) are defined as:
Cr
Cε r = (5-10)
1 + e0
and
Cc
Cε c = (5-11)
1 + e0
where:
Cr = recompression index,
Cc = compression index, and
e0 = initial void ratio.
Methods for determining Cr , Cc , and σ ' p from laboratory tests are illustrated in Figure
3-12. Many useful correlations have been developed between Cc and soil index
properties. Table 5-5 provides a list of some of these correlations. Azzouz et al. (1976)
found that correlations to e0 produced the most accurate prediction of Cc . Nine of the
correlations are compared in Figure 5-12. Additional correlations can be found in
Chapter 8.
The recompression index is typically 5 to 10% of the magnitude of Cc . Typical values

for Cr fall in the range of 0.015 to 0.035 with nearly all results between 0.005 and 0.05
(Leonards 1976).
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Table 5-5 Correlations for Compression Indices
Basis Correlation Applicable soil type Source
=Cc 0.009( LL −10 ) Inorganic soils with sensitivity

Terzaghi and Peck (1967)
less than 4
LL =Cc 0.007( LL −7 ) Remolded clays Skempton (1944)
=Cc 0.006( LL −9 ) Predominantly lean to fat clay Azzouz et al. (1976)
=Cc 1.15( e0 −0.35 ) All clays Nishida (1956)
Cc 0.3( e0 −0.27 )
= Inorganic soil; silt; silty clay;
Hough (1957)
some clay
e0 =Cc 0.75( e0 −0.5 ) Very low plasticity soils Sowers (1970)
Cε c 0.156e0 + 0.017
= All clays Elnaggar and Krizek (1971)
Cc 0.4( e0 −0.25 )
= Predominantly lean to fat clay Azzouz et al. (1976)
Cc =0.01wn Osterberg (1972)
Chicago clays
(in Azzouz et al. 1976)
Cc =0.0115 wn Organic soils, peat Moran et al. (1958)
wn
Cε c =+(
0.1 0.006( wn − 25 ) ) Varved clays Prior NAVFAC DM 7.1
−5 2
Cc 17.66 × 10
= wn
−3 Chicago clays Peck and Reed (1954)
+ 5.93 × 10 wn − 0.135
=Cc 0.01wn −0.05 Predominantly lean to fat clay Azzouz et al. (1976)
Note: Liquid limit ( LL ) and natural water content ( wn ) are in percent.
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Figure 5-12 Common Compression Index Correlations

5-5.2.2 Magnitude of Primary Consolidation.
Primary consolidation settlement ( sc ) should be determined for each compressible

layer. Thick layers should be divided into a series of sublayers (e.g., Figure 5-5b and
c). For each layer, the appropriate equation for sc must be selected based on whether
the soil is normally or overconsolidated and the relative magnitude of the change in
vertical stress.
For normally consolidated soil ( σ 'v 0 ≈ σ ' p ), calculate sc as:
  σ ' + ∆σ z 
sc =  Cε c log  z 0   H (5-12)
  σ 'z 0 
where:
Cε c = modified compression index,
σ 'z 0 = initial vertical stress at the midpoint of the soil layer or sublayer,
∆σ z = change in vertical stress at layer midpoint, and
H = initial thickness of the soil layer or sublayer.
For overconsolidated soil layers in which the final stress is less than or equal to the
preconsolidation stress ( σ 'z 0 + ∆σ z ≤ σ ' p ), calculate sc as:
  σ ' + ∆σ z 
sc =  Cε r log  z 0   H (5-13)
  σ 'z 0 
where:
Cε r = modified recompression index.
For overconsolidated soil layers in which the final stress is greater than the
preconsolidation stress σ 'z 0 + ∆σ z ≤ σ ' p ), calculate sc as:
 σ'   σ ' + ∆σ z 
=sc  Cε r log  p  + Cε c log  z 0   H (5-14)
  σ' 
  σ 'z 0   p 
where:
σ ' p = preconsolidation stress.
Equations 5-12 to 5-14 all follow a consistent pattern in which the compression or
recompression index is multiplied by the logarithm of the change in stress to obtain the
vertical strain. The strain is then multiplied by the layer thickness to obtain the change
in thickness or expected settlement of the layer. In each of these equations, Cε c and
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Cε r can be replaced with Cc and Cr along with the initial void ratio, if desired, using
Equations 5-10 and 5-11. Example calculations are shown in Figure 5-13.
Figure 5-13 Primary Consolidation Example
5-5.2.3 Typical Construction Process.
An example of the primary consolidation caused by changes in stress associated with

typical construction processes is illustrated in Figure 5-14. In this example, the clay is
overconsolidated by past loading. The construction process involves lowering the
groundwater level, excavating for a basement level, and applying the structural load.
Some aspects of construction will cause increases in effective stress and settlement.
Other phases, such as excavation and groundwater rise, will result in swelling. As
noted, the amount of settlement or swell experienced during the first phases of
construction will depend on the rate of construction as well as the rate at which pore
water pressures dissipate in the clay layer.
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Figure 5-14 Vertical Movements during a Typical Construction Process
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5-5.2.4 Corrections to the Magnitude of Consolidation Settlements.
Settlement of overconsolidated clay layers may be overestimated using Equations 5-13

and 5-14 for loads of limited lateral extent. For cases where the width of the load is less
than four times the thickness of the clay layer, conditions deviate significantly from one-
dimensional consolidation. Leonards (1976) recommended that the corrected primary
consolidation settlement can be found by multiplying the calculated value (Equation 5-
13 or 5-14) by a correction factor, α . Values of α can be found using Figure 5-15.
Figure 5-15 Correction Factor for Overconsolidated Clays and Loads of Limited
Lateral Extent (after Leonards 1976)
5-5.3 Time Rate of Primary Consolidation.
The time rate of primary consolidation is considered for situations where predicted
settlement exceeds tolerable values. In these cases, treatment of the foundation soil,
such as acceleration of consolidation or placement of a surcharge to increase in
preconsolidation stress, may be considered. Knowledge of the settlement rate or
consolidation completed at a particular time is important for planning remedial measures
for structures damaged by settlement.
Time rate calculations can be performed with greater accuracy and flexibility using
numerical methods, such as the finite difference method. Various computer programs
are available for this purpose. The analytical methods presented in this section are
useful for understanding time rate of settlement concepts and checking the results of
numerical methods.
The time rate of consolidation is typically assessed starting with one-dimensional theory
applied to vertical drainage. The average degree of consolidation ( U z ) is the average
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percentage of the excess pore pressure that has dissipated at a particular time ( t )
following the addition of a load. The amount of settlement experienced at the ground
surface is typically assumed to be proportional to U z . The value of U z can be related to
a time factor ( T ) for a given set of conditions. The time factor for vertical drainage ( Tv )
is calculated as:
cv t
Tv = (5-15)
H dr2
where:
cv = coefficient of consolidation for the soil layer in the vertical direction,
t = time after application of load, and
H dr = drainage path length.
The relationship between Tv and U z is provided in Figure 5-16.
Figure 5-16 Degree of Consolidation for Instantaneous Uniform Loading and

One-Dimensional Flow
Two conditions are typically considered for vertical drainage. Single (or one-way)
drainage refers to conditions where water can flow one direction to leave the
consolidating soil layer as shown in the inset to Figure 5-16. Double (or two-way)
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drainage occurs when pervious layers lie above and below the consolidating layer, and
H dr is half of the layer thickness ( H ) in this case.
The degree of compression ( U z ) is the amount of pore water pressure that has been
dissipated at a particular depth within the soil layer. The degree of compression will
vary with time and depth within the consolidating layer as illustrated in Figure 5-17. It
can be used to estimate the remaining excess pore pressures at any depth and time
following application of a change in vertical stress. The upper half of this figure can be
used for single drainage conditions.
Figure 5-17 Degree of Compression and Excess Pore Pressure

(Contours Indicate the Time Factor)
5-5.3.1 Effect of Initial Excess Pore Pressure Distribution.
The rate of consolidation can be affected by the initial distribution of excess pore water
pressure, especially for single drainage conditions. The degrees of consolidation and
compression predicted in Figure 5-16 and Figure 5-17 are appropriate for a uniform
distribution of initial excess pore pressure, which is a reasonable assumption for
relatively wide loads, regardless of the type of drainage.
Other scenarios, such as foundation loading and consolidation of hydraulic fill, can
result in a distribution of initial excess pore pressure that is not constant with depth.
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Figure 5-16 is also appropriate for double drainage when the distribution of initial u x
varies linearly. Solutions for single drainage and linearly varying distributions of initial
u x can be found in Terzaghi et al. (1996). However, for these more complex loading
conditions, numerical solutions are preferred.
5-5.3.2 Accuracy of Time Rate Predictions.
The time rate of primary consolidation observed in field measurements is often faster
than that predicted by the methods described in this section. This discrepancy is the
result of a number of effects.
The theoretical conditions of one-dimensional consolidation and vertical drainage rarely

mimic the in situ conditions. In most cases, loading and subsequent drainage is actually
two or three-dimensional, which tends to increase the time rate of consolidation. As the
width of the loaded area becomes small with respect to the thickness of the
compressible layer, consolidation proceeds much more quickly. Figure 5-18 allows the
time required to reach a degree of consolidation of either 50% or 90% to be corrected
for two- and three-dimensional effects. The results are plotted for soil profiles with an
impermeable layer below the compressible soil and for those with an underlying
permeable layer. See Davis and Poulos (1972) for further details on these effects.
Numerical methods can be used to account for some of these differences; however,
many of the common programs used for consolidation calculations only consider one-
dimensional flow.
Many clay soils contain thin seams of sand and silt, which have significantly higher
permeability. These seams provide internal drainage boundaries, greatly reducing the
maximum vertical drainage path length. For example, the presence of a single
additional drainage layer can increase the time rate of settlement by a factor of four. In
addition, soil deposits tend to have a higher horizontal hydraulic conductivity, which
coupled with three-dimensional effects can increase the settlement rate. Figure 5-19
provides a simple means to account for effect of anisotropy in the coefficient of
consolidation on the time required to reach 50% consolidation.
Finally, disturbance of soil samples tends to decrease the coefficient of consolidation

measured in laboratory tests. Even high-quality samples and tests have some degree
of disturbance, leading to lower cv and slower time rate of settlement predictions.
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Figure 5-18 Effect of Load Geometry on Time Rate of Consolidation

(after Davis and Poulos 1972)
Figure 5-19 Effect of Anisotropy on Time Rate of Consolidation

(after Davis and Poulos 1972)
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5-5.3.3 Gradual Load Application.
The length of the construction period or the amount of time required to apply the load
can also affect the time rate of primary consolidation. Gradual load application modifies
the relationship between the degree of consolidation and the time factor. Figure 5-20
provides a method to account for gradual loading.
Figure 5-20 Degree of Consolidation for Gradual Load Application for Vertical
Drainage (after Olson 1977)
5-5.3.4 Coefficient of Consolidation.
The coefficient of consolidation can be estimated based on index properties (see

Chapter 8), calculated from volume change vs. time measurements in laboratory tests,
or inferred from field measurements of pore pressure dissipation.
5-5.3.4.1 Laboratory Measurement of Coefficient of Consolidation.
The vertical coefficient of consolidation, cv , is often found from data obtained using
incrementally loaded one-dimensional consolidation tests on vertically oriented
specimens. The coefficient of consolidation in other directions can be determined by
trimming and mounting specimens in other orientations in the testing equipment.
Regardless of the specimen orientation, the data is normally assessed using either the
Casagrande or the Taylor method, which are described in Figure 5-21. Volume change
measured during the consolidation phase of shear strength tests can also be used with
these procedures. Other methods exist for determining cv from incrementally loaded
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consolidation tests, and a single equation can be used to determine cv from constant
rate of strain consolidation tests.
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Figure 5-21 Determination of Coefficient of Consolidation from Laboratory Data

(Note: the y-axis can also be plotted as volume or void ratio)
5-5.3.4.2 Use of Field Measurements.
Field observations of excess pore pressure dissipation with time can be used to
measure the in situ coefficient of consolidation. At any given time after loading, the ratio
of excess pore pressure to change in vertical stress ( u x ∆σ v ) can be measured and
plotted at the appropriate normalized depth on Figure 5-17. The corresponding time
factor can be estimated from the contours, and cv can be calculated using Equation 5-
15.
An example of both laboratory and field determination of cv is provided in Figure 5-22.
5-5.3.5 Time Rate of Consolidation for Layered Profiles.
The consolidating soil may contain layers with varying values of cv . In this situation, the
behavior of a layered system can be approximated by conversion to an equivalent
single layer system using the following procedure:
1. Select any layer ( i ) with properties cv ,i and H i .

2. Transform the thickness of every other layer with properties cv ,n and H n to an
equivalent layer with the soil properties of layer i as follows:
cv ,i
H 'n = H n (5-16)
cv ,n
3. Calculate the total thickness ( H 't ) of the transformed system as:
H 't = ∑ H 'n (5-17)
4. Treat the system as a single layer of thickness ( H 't ) with coefficient of

consolidation ( cv ,i ),
5. Use the appropriate method, such as Figure 5-16, to determine the percent
consolidation at various times.
The complexity introduced by multi-layer problems is well-suited to the use of numerical

analysis for the calculation of time rate of settlement. An example of multi-layer time
rate of consolidation calculation is provided in Figure 5-23.
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Figure 5-22 Determination of cv from Lab and Field Data
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Figure 5-23 Multi-layer Consolidation Example
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5-5.4 Secondary Compression of Fine-Grained Soils.
Secondary compression settlement ( ss ) occurs as soil particles rearrange and the soil
structure creeps under constant vertical effective stress. For practical purposes,
secondary compression can be assumed to occur after the end of primary consolidation
and to follow a relatively linear trend with respect to time on a log scale such that:
 C  t    t 
=ss = α
log    H 0  Cεα log    H 0 (5-18)
 1 + e0  t   t 
  p    p 
where:
Cα = secondary compression index,
Cεα = modified secondary compression index,
e0 = initial void ratio,
t = time after loading,
t p = time required to finish primary consolidation, and
H 0 = initial layer thickness.
Values of Cα or Cεα can be obtained from the results of a one-dimensional

consolidation test that is allowed to creep for a significant length of time past the end of
primary consolidation. The magnitude of Cα is stress dependent for a particular soil and
should be determined at an effective stress similar to that expected in situ. For a variety
of clays, silts, and organic soils, Mesri (1973) showed that Cεα is approximately related
to the natural water content, wn , (in percent) by:
Cεα ≈ 10−4 wn (5-5-19)
The secondary compression index is closely linked to the compression index, Cc . For
this reason, an excellent method for estimating Cα is the use of the Cα Cc ratio. Typical
values of this ratio are found in Table 5-6. Recognizing that Cc is also stress-
dependent, these ratios should be used with the slope of the laboratory virgin
compression curve at the effective stress of interest (Mesri and Godlewski 1977, Mesri
and Castro 1987) rather than the field corrected value typically used in consolidation
calculations. An example of secondary compression settlement calculations is provided
in Figure 5-24.
Table 5-6 Typical Values of Cα C c (after Terzaghi et al. 1996)
Soil Type Cα Cc
Granular soils including rockfill 0.02 ± 0.01
Shale and mudstone 0.03 ± 0.01
Inorganic clays and silts 0.04 ± 0.01
Organic clays and silts 0.05 ± 0.01
Peat and muskeg 0.06 ± 0.01
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Figure 5-24 Calculation of Secondary Compression
In many cases, one-dimensional consolidation tests are presented using the void ratio
or vertical strain corresponding to the end of 24-hour load increments. Some amount of
secondary compression will have occurred during this time period and is thus included
in the laboratory consolidation curve. The engineer should be aware that consolidation
settlements calculated with such data will account for some degree of secondary
compression.
If secondary compression is expected to be important in the analysis, consolidation test

results should be plotted based on the void ratio or vertical strain at the end of primary
(EOP) consolidation for each load increment. Primary consolidation settlement
calculated using the EOP consolidation curve can be added directly to estimates of
secondary compression settlement.
5-5.5 Organic Soils and Peat.
Settlement of organic soils and peat can be calculated using the procedures for
inorganic fine-grained soils. Primary consolidation tends to occur more rapidly in these
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soils due to their relatively high hydraulic conductivity. Large secondary compression is
typically measured and contributes a significant portion of the total settlement.
5-6 DIFFERENTIAL AND TOLERABLE SETTLEMENT.
5-6.1 Differential Settlement.
For important structures, settlement should be calculated for a number of points across
the footprint of the structure in order to establish the expected settlement pattern. Figure
5-25 illustrates various types of settlement profiles.
Figure 5-25 Components of Settlement

(after Duncan and Buchignani 1987, Ricceri and Soranzo 1985)
Differences in settlement across a structure can be considered in multiple ways. The

simplest is to consider the maximum difference in predicted settlement across the
structure, which is often referred to as the differential settlement, δ max . However, it is
often necessary and more informative to consider the effects of the size and flexibility of
the structure when evaluating the impact of differential settlement. In some cases, a
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structure will tilt over an angle ( ω ) which causes difference in settlement between
various points but does not necessarily cause structural distress.
Differential settlement that results in non-uniform deflection or non-uniform tilt causes

bending and tensile strain in the structure. Differential movement that causes a
concave upward shape is referred to as sagging while a concave downward shape is
called hogging. This type of movement can be quantified by either the angular
distortion or the deflection ratio. Angular distortion ( δ l ) is the slope of the expected
settlement profile or the ratio of the settlement between two points to the distance ( l )
separating the points. The deflection ratio ( ∆ L ) is the maximum expected deviation
from uniform settlement divided by the overall length of the structure and is an
approximate measure of the curvature caused by settlement. These two measures
have been shown to provide the best indication of structural distress.
Natural variation in soil deposits causes settlement calculations to be highly uncertain.

For this reason, it may be sufficient to use approximate relationships to estimate
differential based on magnitude of total settlement. For example, Terzaghi et al. (1996)
suggest that the differential settlement for footings on sand will likely be 75% or less of
the predicted total settlement. For clays, the differential settlement can sometimes
approach the magnitude of the total settlement.
5-6.2 Tolerable Settlement.
5-6.2.1 Criteria.
Most of the guidance regarding acceptable settlement is based on experience with

measured settlement of real structures and observations of the associated damage.
Wahls (1981) summarizes four important points regarding settlement: (1) settlement
must be expected, (2) some form of differential settlement is the most important to
consider for structural distress, (3) structural design can reduce the level of differential
movement, and (4) many structures can tolerate large settlements and remain safe.
Table 5-7 provides guidance for the selection of tolerable angular distortion for various
types of structures. In some cases, the limits that are provided that include a margin of
safety to reduce or prevent cracking or damage. Most of these guidelines ignore the
size and stiffness of the structure, which control the magnitude of the strains in the
structure.
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Table 5-7 Angular Distortion Limits for Various Structures

(after Skempton and MacDonald 1956, Polshin and Tokar 1957,
Duncan and Buchignani 1987, and Day 1990)
Tolerable δ l
Type of Structure / Condition L H
Decimal Ratio
Small slab-on-grade structures – damage to structure --- 0.01 1/100
Steel frame with flexible siding ---

0.008 1/125
Circular steel tanks on flexible base with fixed top ---
Considerable cracking – brick and panel ---
Structural damage begins --- 0.0067 1/150
Safe limit – flexible brick walls (includes safety factor) >4
Tilting of high buildings becomes visible --- 0.004 1/250
Slab-on-grade structures – drywall cracking

--- 0.0033 1/300
Overhead cranes
Steel or reinforced concrete frame with insensitive finish such
---
as dry wall, glass or moveable panels 0.002 to 1/500 to
0.003 1/333
Circular steel tanks on flexible base with floating top ---
Tall slender structures, such as stacks, silos, and water tanks
---
with rigid mat foundations
0.002 1/500
Safe limit –cracking of buildings (includes safety factor) ---
Steel or reinforced concrete frame with brick, block, plaster, ≥5 0.002 1/500
or stucco finish ≤3 0.001 1/1000
Frames with diagonal bracing --- 0.00167 1/600
Machinery sensitive to settlement --- 0.0013 1/750

≥5 0.0008 1/1250
Load-bearing brick, tile, or concrete block walls
≤3 0.0004 1/2500
Wroth and Burland (1974) proposed a deflection ratio criterion that allow the properties
of the structure to be explicitly considered in terms of the controlling type of strain, the
length to height ratio of the structure, and the structure’s relative stiffness, E G . Figure
5-26 plots these criteria for three different types of bending, assuming the critical strain
for structural distress ( ε crit ) is 0.075%. Other values of E G and ε crit can be
considered using the equations provided on the figure.
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Figure 5-26 Allowable Deflection Ratios Related to Structural Proportions

(after Burland and Wroth 1974, Wahls 1981)
5-6.2.2 Reduction of Differential Settlement.
The reduction of total settlement (Section 5-7) is also the primary means of reducing
differential movement. Settlement that occurs early in the construction process doesn’t
generally contribute to structural distress. The sequence and rate at which the load is
applied can be compared to the expected rate of consolidation to estimate the total and
differential settlement that will be experienced by different parts of the structure. If the
consolidation rate is relatively fast, building finishes and other sensitive components
may be installed after much of the total settlement has occurred and thus will
experience less differential settlement than the superstructure. Estimates of this type of
effect are heavily dependent on the rate of construction and consolidation and must be
considered on a project-specific basis.
Some buildings, such as light steel frame structures, are very flexible and can tolerate
large settlements. In this case, limitation of damage to utilities and machinery housed in
the facility may control design.
5-6.3 Differential Settlement of Mat Foundations.
Settlement calculations for mat or raft foundations are often made based on changes in
stress calculated assuming uniform loading. The rigidity of the foundation structure will
affect the accuracy of this assumption and the distribution of settlement. The flexibility
of the foundation and the soil structure interaction can be difficult to assess. Predictions
of differential settlement are often less accurate than those for total or average
settlement for these reasons, especially for mat foundations.
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Mat behavior can be estimated using a mat stiffness factor ( K m ) which compares the
foundation to the stiffness of the underlying soil. As indicated in Table 5-8, mats with
low stiffness ratios can be considered completely flexible. Flexible mats will apply a
relatively uniform pressure distribution, and the center, edges, and corners will settle
differentially. Mats with high values of K m will act in a rigid manner and will tend to
settle uniformly. Influence factors for intermediate stiffness are provided by Brown
(1969) and Frazer and Wardle (1976).
Table 5-8 Relative Mat Stiffness and Behavior

(after Brown 1969, Frazer and Wardle 1976)
Mat Behavior
Foundation Mat Stiffness Factor, K m
Shape
Flexible Intermediate Rigid
2
Em (1−ν s ) tm3
Circular Km = 8 K m ≤ 0.08 0.08 ≤ K m ≤ 5 5≤ K m
Es B3
2
4 Em (1−ν s ) tm3
Rectangular Km = K m ≤0.05 0.05≤ K m ≤10 10≤ K m
3 Es (1−ν m )2 B3
Em = modulus of elasticity of mat, ν m = Poisson’s ratio of mat
Variables: E s = modulus of elasticity of soil, ν s = Poisson’s ratio of soil

t m = thickness of mat, B = diameter or width of mat (least dimension)
5-7 METHODS OF CONTROLLING SETTLEMENT.
Methods for reducing or accelerating settlement are summarized in Table 5-9. Further
details for some of the basic methods are discussed in the following sections. For a
more in-depth summary of ground modification techniques that can be used to remove
settlement potential, reduce excess settlement, or accelerate settlement, see FHWA’s
Ground Modification Vol. 1 and 2 (2017).
5-7.1 Removal or Displacement of Compressible Soils.
Excavation and replacement is a simple method of reducing or eliminating settlement

for cases were the compressible stratum is shallow and relatively thin. This method is
particularly useful for sites where extensive earthwork is already required.
Surficial soils with low shear strength and high compressibility, such as organic soils,
should be removed and replaced with engineered fill. The suitability of deeper soils to
support the planned fill or structure will depend on the shear strength and
compressibility of the underlying soils as judged by an appropriate subsurface
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exploration. A common example of removal and replacement is the removal of topsoil

which occurs prior to the placement of most fills.
In some cases, partial excavation of very soft foundation soils can be accompanied by
displacement caused by the weight of the new fill. The boundary between the new fill
and the displaced soil should be kept as vertical as possible. This method is most
applicable to the displacement of peat and muck deposits and has been used
successfully for soft soils up to 65 feet deep. Jetting and blasting methods can be used
in the fill and foundation to facilitate displacement of the soft soil. Caution should be
used with partial displacement in fibrous organic soils, as these materials tend to resist
displacement, which can result in trapped pockets and differential settlement.
Table 5-9 Methods to Reduce, Accelerate, or Prevent Excess Settlement

(after FHWA 2017)
Primary
Method / Technology Description / Comments
Purpose
Reduce Removal and replacement Full or partial removal of compressible soil reduces or
amount of eliminates settlement potential. Displacement methods
soft soil Partial displacement may include jetting or various types of blasting.
Column-supported embankments
Compressible soil is bypassed by much stiffer,
Reinforced load transfer platforms reinforcing elements, typically columns. The columns
transfer most of the load from the fill or structure through
Reinforce
Non-compressible columns the soft soil to deeper and stiffer materials. A reinforced
soft soil
load transfer platform is often required at the ground
Stone columns surface. In some cases, the reinforcing elements also
improve the surrounding soil.
Rammed aggregate piers
Deep mixing methods

Compaction and/or chemical modification of soft soil can
Improve soft reduce its compressibility and reduce settlement
Vibro-compaction
soil potential. Compaction methods are typically more
effective on coarse-grained materials.
Dynamic compaction
Surcharge
Increased gradients cause the consolidation process to
proceed more quickly. Final settlements can be equal to
Pumping or vacuum
or greater than the settlement caused by the design load.
Accelerate Often used along with vertical drains.
consolidation Electro-osmosis
of soft soil
Prefabricated vertical drains (PVDs) Vertical drainage elements are used to shorten the
drainage path. This allows pore pressures to dissipate
Aggregate columns or sand drains and consolidation to occur more quickly.
Balanced / compensated foundation

Reduction of the applied stress will reduce the
Reduce
magnitude of the consolidation settlement. This can be
applied load Lightweight granular fill
accomplished by permanent excavation or the use of
to soft soil
lightweight construction materials.
Geofoam or foamed concrete
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5-7.2 Balancing Load by Excavation.
The balanced load approach (a.k.a. compensated foundation or floating foundation) can
sometimes be used to support heavy structures over compressible strata. In this
approach, the weight of the structure is balanced, completely or partially, by soil that is
permanently excavated from the building footprint. The construction of a permanent
basement level is required to create stress relief. This method works particularly well
for situations where a stronger surface layer overlies a compressible stratum.
Excavation for the structure results in vertical stress relief and some amount of swelling
or heave. If the weight of the structure is equal to or less than the weight of the
excavated material, the total settlement experienced by the structure will be the result of
recompression of the compressible strata. The magnitude of swelling and subsequent
recompression will depend on construction and site factors, such as the amount of time
between excavation and loading, the construction sequence, and the subsurface
drainage conditions.
Dewatering may be required to facilitate the construction of a balanced foundation. If

the groundwater is significantly lowered, the amount of heave and subsequent
recompression will be reduced because of negative excess pore pressures in the soil.
Settlement for balanced foundations can be predicted using the methods in Sections 5-
3 to 5-5. The net vertical stress ( q0− net ) is found by subtracting the vertical stress
reduction ( σ z − red ) from the building load, q0 . The final groundwater conditions affect the
value of σ z − red . For cases where the groundwater is not lowered or is allowed to return
to its initial condition following construction, σ z − red is equal to the total vertical stress at
the foundation level prior to construction. If the groundwater is lowered permanently
below the foundation level, σ z − red is equal to the effective vertical stress at the
foundation level prior to construction.
5-7.3 Preconsolidation by Surcharge.
A surcharge causes some or all of the consolidation to occur prior to construction. This
method works well for fill beneath paved areas, for large floor loadings, and for
structures with relatively light column loads. For heavier structures, improvement by
surcharging may not be sufficient to provide adequate bearing resistance. In this case,
a portion of the surficial layer of the compressible soil may need to be replaced with a
more rigid compacted fill or reinforced fill.
A portion of the predicted consolidation and secondary compression can be removed by

surcharging as illustrated in Figure 5-27. The calculations in this figure assume that the
time rate of consolidation is the same under both the surcharge load and the final load.
This should be approximately true provided the coefficient of consolidation is not
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significantly different under the two loading conditions. In order to eliminate the
settlement due to primary consolidation under the final load, the degree of consolidation
required under the surcharge ( U f + s ) is:
 q 
log 1 + f 
U f +s =  σ 'z 0  (5-20)
 q  q 
log 1 + f 1 + s  

 σ 'z 0  q f  
where:
q f = final applied load,
qs = additional surcharge load, and
σ 'z 0 = initial vertical effective stress at midpoint of the consolidating layer.
If some amount of secondary compression must also be removed by surcharging, the

required degree of consolidation becomes:
  q f  Cα  t 
log 1 + + log 
t  
  σ 'z 0  Cc  p  
U f +s = (5-21)
 q  q 
log 1 + f 1 + s  
 
 σ 'z 0  q f  
where:
q f = final applied load,
qs = additional surcharge load,
σ 'z 0 = initial vertical effective stress at midpoint of the consolidating layer,
Cα Cc = ratio of secondary to primary compression indices,
t p = time required for primary consolidation, and
t = time after loading for which secondary compression is considered.
An example of the calculations for the surcharge approach is provided in Figure 5-28.
The major limitations of the surcharge method are time and cost. The time required for
consolidation to occur may not fit within the construction schedule. Use of vertical
drains as described in the following section can alleviate this difficulty. In soft soils,
shear failure can be induced at the edge of the surcharge. This should be considered
using slope stability methods (Chapter 7) or bearing capacity theory (DM 7.2).
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Figure 5-27 Surcharge Load and Consolidation Required to Eliminate Settlement

under Final Load
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Figure 5-28 Surcharge Loading Example
5-7.4 Vertical Drains.
Vertical drains are constructed or inserted vertically through compressible soil layers.
The drains intercept horizontal water flow. The water is then transmitted to a drainage
layer at the surface and/or to underlying coarse-grained soil, depending on the drainage
conditions. The drains are typically installed in triangular or square patterns as shown
in Figure 5-29 with spacing ranging from 3 to 6.5 feet. Drains typically shorten the
maximum drainage path to 8 feet or less. While vertical drains were constructed mostly
of sand prior to the 1980s, current practice is to use drains comprised of a plastic core
encased in geotextile fabric, typically referred to as prefabricated vertical drains (PVD).
For detailed information on PVD materials and drain construction practices, see FHWA
(2017).
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Figure 5-29 Vertical Drains – (a) Triangular Pattern, (b) Rectangular Pattern, and
(c) Equivalent Cylinder for Theoretical Solutions
In addition to shortening the drainage path, vertical drains take advantage of the higher
hydraulic conductivity and coefficient of consolidation often found in horizontal direction
in many soils. While the vertical coefficient of consolidation can be measured in
laboratory tests or estimated based on index tests (see Chapters 3 and 8), the
horizontal coefficient of consolidation ( ch ) is rarely measured directly. More often, ch is
assumed to be about 1.5 to 4 times higher than cv , depending on the amount of
horizontal layering present. Higher ratios of ch to cv are encountered when layers of silt
and sand are present. Asaoka (1978) presents a method for determining ch from field
measurements on a test fill.
Many analytical and numerical methods have been proposed for radial drainage theory,
most based on Barron (1948). The differences in the methods tend to have less effect
on predictions of consolidation rate than the uncertainty and variability in ch . Using the
calculation method presented by FHWA (2017), the time factor ( Tr ) for a desired degree
of radial consolidation ( U r ) is found by:
1 1 
T=
r ( Fn + Fs + Fr ) ln   (5-22)
8  1−U r 
where:
Fn = factor related to drain spacing,
Fs = factor related to soil disturbance (smear), and
Fr = factor related to well resistance in the drain.
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If smear and well resistance are ignored, Fs and Fr are set equal to zero.
Figure 5-29 illustrates the relationship between the drain configuration and the effective
drainage diameter, d c . The drain spacing is often expressed in terms of the ratio ( n )
between the drainage diameter and the diameter of the well, d w . The drainage factor (
Fn ) is equal to:
n2 3n 2 − 1
Fn
= ln ( n ) − ≈ ln ( n ) − 0.75 (5-23)
n2 − 1 4n 2
where:
n = ratio of d c to d w .
The error in the approximation is less than 10% for n greater than 4 and less than 1%
for n greater than 12.
Most PVDs have a rectangular cross section while the solutions to radial consolidation
problem assume a circular drain. Hansbo (1979) found that the equivalent drain
diameter can be approximated as:
2 (a + b)
dw = (5-24)
π
where:
a and b = PVD dimensions.
Values of d w for modern PVDs range from 1.5 to 5.5 inches. A diameter of 2 inches is
commonly used, which results in n values in the range of 20 to 50 for typical drain
spacing.
Given the uncertainties with the measurement or estimation of soil properties, the
effects of soil disturbance or smear around the drains and drain resistance are often
ignored (FHWA 2017). Smear tends to be important mostly for drains in high plasticity
clays or sensitive soils, or where ch has been directly and accurately measured. In
order to account for smear, the soil disturbance factor ( Fs ) can be calculated as
kh
Fs ≈ ln ( s ) (5-25)
ks
where:
kh = hydraulic conductivity of the soil layer,
k s = hydraulic conductivity of the disturbed zone, and
s = ratio of the diameter of the disturbed zone to the diameter of the drain.
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Resistance to water flow within the drain can also decrease the effectiveness of vertical
drains. If desired, this effect can be estimated by:
kh
=Fr π z ( 2 Lm − z ) (5-26)
qw
where:
z = depth along the drain,
Lm = maximum distance water must flow through the drain, and
qw = discharge capacity of the drain.
5-7.4.1 Combination of Vertical and Horizontal Drainage Effects.
Prediction of the degree of consolidation from vertical drainage can be combined with
the effects of horizontal drainage to vertical drains using the method proposed by
Carrillo (1942). The combined degree of consolidation ( U c ) in percent is:
U=c 100 −
(100 − U )(100 − U )
r z
(5-27)
100
where:
U r = degree of consolidation for radial drainage, and
U z = degree of consolidation for vertical drainage.
5-7.4.2 Vertical Drain Design.
Vertical drain design involves the selection of the appropriate drain type, drain spacing,
and construction procedures based on the time available for consolidation, design
degree of consolidation for that time period, and soil properties. The appropriate time
factor for radial drainage ( Tr ) can be determined from Equation 5-22 or from Figure 5-30
or Figure 5-31. Figure 5-30 plots the relationship between U r and Tr for a range of drain
spacing for three smear conditions. Figure 5-31 provides solutions for gradual loading
and radial drainage. From Tr and the time available for consolidation ( t ) the required
effective drain diameter ( d c ) can be calculated as:
cht
dc = (5-5-28)
Tr
where:
ch = coefficient of consolidation in the horizontal direction.
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Figure 5-30 Degree of Radial Consolidation

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Figure 5-31 Radial Consolidation with Gradual Loading (after Olson 1977)
An alternative approach is provided in the design chart in Figure 5-32. This chart allows
the spacing to be selected directly based on the other properties and variables. Other
drain design considerations include stability against foundation failure and provision of
adequate flow in the surface drainage blanket. An example of vertical drain design is
provided in Figure 5-33.
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Figure 5-32 Design Chart for Radial Drainage
5-7.4.3 Vertical Drains with Surcharge.
In some cases, a surcharge is used along with vertical drains to accelerate the rate of
settlement and reach the final settlement more quickly. Surcharges are especially
important for soil conditions in which a large amount of secondary compression is likely
to occur. The method presented in Figure 5-27 can also be used with vertical drains.
5-7.4.4 Construction Control Requirements.
Extensive discussion of specifications, quality assurance, site preparation, and

installation procedures for vertical drains can be found in FHWA (2017). Field
instrumentation should be installed to monitor performance of the drains, progress of
consolidation, and horizontal deformations. The type of instrumentation required
depends on the application. For cases where instability is not likely, such as a low
height fill of large lateral extent, the primary purpose of instrumentation is to monitor
progress of consolidation. Settlement plates are sufficient for this purpose. On the
other hand, stability and pore pressure dissipation is a concern for the construction of
large embankments over soft soil. Piezometers should be used to monitor the
dissipation of excess pore pressure during consolidation. Inclinometers should be
installed to monitor horizontal deformations in the foundation soil.
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Figure 5-33 Radial Consolidation Example
5-8 VOLUME EXPANSION.
5-8.1 Mechanics of Volume Expansion.
Positive volume change or volume expansion of soil is controlled by a variety of factors.

For fine-grained soils, these factors include physical particle interactions, chemical
interactions, mineralogy, soil fabric or structure, stress history, temperature, and pore
water chemistry.
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Unloading or reduction in effective stress is the most important physical process that
causes swelling in non-frozen soils. This unloading can be the result of a decrease in
the total stress, such as that caused by an excavation, or as the result of increased
positive pore water pressure from a raised groundwater level.
Various theories have attempted to explain the chemical processes associated with
swelling, including osmotic pressure theory and water adsorption theory (Mitchell 1993).
These theories predict the swell pressure, which is the pressure exerted by the swelling
soil on an unyielding boundary. While such theories have a limited degree of accuracy
and are not typically useful for practical application, they provide insight into general
trends in soil behavior. The osmotic pressure concept shows that pore water with low
electrolyte concentration leads to higher swell pressures. Similarly, according to water
adsorption theory, the specific surface area of the clay particles is the most important
factor for determining the amount of water required for hydration. For this reason, clay
minerals with very thin particles with high surface area, such as montmorillonite,
smectite, and vermiculite, are the most susceptible to swelling. The liquid limit and clay
fraction are indicators of the amount of these swelling clay minerals present in a soil
(Terzaghi et al. 1996).
Soil and rock that has been consolidated to a relatively low void ratio is the most
susceptible to swelling. Low void ratios are the result of either high normal stresses or
high levels of matric suction (ψ ) under unsaturated conditions. The deformation
experienced during consolidation stores energy in the soil particles. In addition, the
water content in these soils may be lower than that required to fully hydrate the clay
particles. If either the normal stress or the suction is lowered, the clay will tend to swell
to release the stored energy and to hydrate the clay minerals. Some clay minerals,
such as kaolinite, exhibit swelling mostly at the low void ratio associated with heavy
overconsolidation. Clays containing more active minerals, such as sodium
montmorillonite, experience similar amounts of swell regardless of void ratio because
the swelling behavior is dominated by hydration.
Clay shales and shales containing pyrite (iron sulfide) or anhydrite (calcium sulphate)
are also susceptible to swelling when exposed to water and/or air. The oxidation and
hydration of pyrite and anhydrite can cause a volumetric expansion of ten times the
original volume.
Soils can also experience volume expansion caused by freezing and the formation of
ice lenses within the soil. Silts, silty sands, and fine sands are particularly susceptible to
frost-related swell. These soils have moderate hydraulic conductivity which allows
water to flow easily and a small enough void space to permit a significant level of
capillary rise. For design of foundations in frost-susceptible soils, see ASCE (2001).
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5-8.2 Effects of Volume Expansion.
Below an excavation, the total vertical stress is reduced, which initially causes a
reduction in pore water pressure. As the pore water pressures return to equilibrium, the
effective stress on the soil reduces and the soil swells. In most cases, excavation is
followed by the construction of a building and the application of a pressure ( q0 ) that
meets or exceeds the reduction in total stress. In this case, the heave caused by stress
reduction will be cancelled out by reloading. Movements during these stages of
construction are typically difficult to predict.
Negligible heave is observed for excavations in coarse-grained soils above the water
table. Soft clays will experience immediate distortion-related swell that can be predicted
using the method in Section 5-5.1. However, the required elastic modulus can be very
difficult to predict. Over time, clay soils will experience an increase in water content and
swell as a result of the change in effective stress.
In arid climates, changes in water content tend to vary depending on the location below
a structure. The soil below the middle of the structure interacts less with the
environment, and the water content in this zone tends to increase with time. This
results in swell below the middle of the structure. Around the edges, the soil
experiences more fluctuation in water content and can shrink, leading to perimeter
settlement. The combination of these two mechanisms can result in the hogging shape
of differential movement described earlier.
5-8.3 Estimates of Heave or Swell Pressure.
Many methods have been proposed to estimate one-dimensional heave (vertical

movement) or swell pressure. Some of these methods are empirical, based mostly on
index properties of the soil. Other methods use theory and oedometer testing to predict
swelling caused by both changes in total stress and changes in suction.
Most methods predict a swell percentage or vertical strain of the soil, which must be
multiplied by the thickness of soil that experiences swell. The thickness of soil that
swells as a result of a change in total stress, such as an excavation, is related to the
size of the excavation. This depth of influence can be estimated using the methods
presented in Chapter 4. The thickness of soil that swells due to changes in suction is
related to the depth of the groundwater table and the depth of seasonal water content
fluctuations, which is commonly in the range of 3 to 15 feet depending on climate.
In some cases, it is helpful to consider the swell pressure that an expansive soil or rock
can develop against an unyielding support or structure under a certain set of initial
conditions. ASTM D4546 provides three different methods to measure the swell
pressure in a one-dimensional consolidation apparatus. The swell pressure mobilized
in situ is often less than that measured in the laboratory (Terzaghi et al. 1996).
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5-8.3.1 Empirical Relationships.
Some of the available empirical correlations related to swelling are summarized in Table
5-10. These methods are mostly based on the Atterberg limits, clay fraction, and initial
soil state as described by unit weight and water content. A few of the correlations
require the soil’s specific gravity or initial matric suction.
Table 5-10 Empirical Correlations to 1D Heave and Swell Pressure and Required
Input Parameters (after Rao et al. 2011, Vanapalli and Lu 2012)
Grain-Size Distribution
Initial State (γ, w)

Atterberg Limits
Specific Gravity
Initial Suction
Empirical
Method for Source
Prediction of:
Seed et al. (1962) X X

Van der Merwe (1964) X X
Ranganatham and Satyanarayan (1965) X X
Vijayvergiya and Ghazzally (1973) X X
McCormack and Wilding (1975) X X
O'Neil and Ghazzally (1977) X X
1D Heave
Johnson and Snethen (1978) X X
Weston (1980) X X
Bandyopadhyay (1981) X X
Chen (1975) X
Cokca (2002) X X X
TXDOT (2014) X X X
Komornik and David (1969) X X
Nayak and Christensen (1971) X X X
Schneider and Poor (1974) X X
McCormack and Wilding (1975) X X
Johnson (1978) X
Swell Pressure
Nayak (1979) X
Erzin and Erol (2004) X X
Sridharan and Gurtug (2004) X X
Thakur and Singh (2005) X
Erzin and Erol (2007) X X X
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Correlations, such as those listed in Table 5-10, are useful because of their simplicity
and the ready availability of input information. However, the correlations are based on
limited data sets and are most appropriate for application in similar soils. In addition,
correlations that are only based on Atterberg limits and clay fraction will not be able to
account for differences caused by initial soil conditions. None of the correlations
presented in Table 5-10 is able to account for the contribution of change in stress to
heave or the effects of total normal stress on swelling.
5-8.3.2 Stress-Strain-Suction Relationships.
More rigorous predictions of one-dimensional swelling consider the vertical strain that
results both from changes in total stress and from changes in suction. These methods
require measurement of the soil’s stress-strain relationship, often at different levels of
controlled suction. Suction-controlled oedometer tests and coefficient of linear
extensibility tests are two means to obtain this data.
A variety of methods have been developed to predict the swelling strain caused by
changes in total stress and changes in suction (or water content). Terzaghi et al. (1996)
describes how the swelling process can be described in a manner analogous to
consolidation. Vanapalli and Lu (2012) summarize many different methods to account
for the stress-strain-suction relationship in the calculation of vertical strain based on the
results of one-dimensional swell pressure measurements, suction-controlled oedometer
tests, and coefficient of linear extensibility tests. Some of the methods also require
measurement of matric suction as a function of water content. Terzaghi et al. (1996)
also emphasizes that the deterioration of the clay structure during swelling often leads
to a significant amount of time-dependent secondary swelling.
While theoretically sound, predictions of swell based on stress-strain-suction

relationships are usually impractical. The amount of swell predicted by these methods
is heavily dependent on soil moisture conditions at the start of construction, which
cannot be accurately predicted during the design phase. In addition, the advanced soil
testing required to use these methods is typically unavailable.
5-8.4 Design in Expansive Soils.
In many cases, heave and swell pressure estimates are used mostly to make decisions
regarding remedial treatment of expansive soils and rock because of the uncertainties
inherent in these estimates. Design of structures and pavements focuses on efforts to
eliminate or reduce the effects of shrink-swell behavior of expansive soils and rock that
are deemed to be problematic. Table 5-11 summarizes common approaches to
foundation design in expansive soil and rock.
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Table 5-11 Foundation Design Approaches in Expansive Soil

(after Bowles 1996)
Approach Comments
Possible admixtures include lime, cement, or kiln dust. The admixtures

reduce the hydration demands of the clay minerals and also increase bonding
Alter the soil
that resists swelling. Limited to depths for which it is practical to mix and
recompact the soil.
Compaction on the “wet” side of the line of optimums results in degree of

saturation greater than 80 to 85%. High initial saturation reduces the potential
Wet compaction for swelling but will increase shrinkage potential in areas exposed to the
atmosphere. Wet compaction results in lower dry unit weight that may have
lower shear strength and stability.
Construct void zones within the foundation system, such as waffle slabs. If
Control direction
the soil has a tendency to swell, it will first swell into the voids prior to affecting
of swelling
the structure.
Environmentally driven changes in water content cause most problematic

swelling. Swelling can be bypassed by extending the structure below the zone
Eliminate changes
of active water content change. If the structure cannot be constructed at this
in water content
depth, the excavation can be filled with soils that are not susceptible to
swelling and have low hydraulic conductivity.
Use a capillary In cases where the source of water is migration from deeper soils, a capillary
break break of coarse-grained material or geomembrane may be useful.
Use a sealing Asphaltic sealing compounds can be used on expansive shale to reduce or
compound prevent water from reaching the rock.
The foundations can be extended below the active zone depth and designed
Design against with sufficient uplift capacity to resist the forces applied to the structure by
swelling swelling soil. Drilled piers with a bond break along the side of the shaft in the
active zone are one common approach using this method.
For some structures, the structural loads can be concentrated to increase the
bearing pressure to levels that meet or exceed the swell pressure. This
Balance swell
approach is not practical for most low-rise buildings because the structural
pressure
loads are too light. In addition, the bearing pressure required to resist swell
may exceed the allowable bearing capacity of the soil.
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5-8.5 Construction Practices in Expansive Soils.
Good construction and maintenance practices in expansive soils and rock are primarily
related to limiting exposure to the atmosphere and changes in water content. For
swelling caused mostly by a temporary reduction in vertical stress, swelling can be
reduced by collecting and removing surface water, pumping down groundwater, and
placing a concrete mudmat immediately after excavation. Inert shale should be
protected from wetting both during construction and long-term through the elimination of
underdrainage, use of impervious backfill, and placement of appropriate surface
drainage features. Excavations in shale should not be completed to final grade until
foundation concrete is ready for placement. Some level of temporary and permanent
protection can be provided through asphaltic coatings.
Some structures extend below future surface water or the groundwater table. In this
case, access to water is impossible to avoid. The methods described in Table 5-11 can
be used to reduce or restrict swelling. Where rock is shallow, rock bolts can be used
with appropriately reinforced slabs and foundations to resist swelling.
For buildings in semi-arid and arid climates, changes in suction within the soil are the
major cause of shrinking and swelling. Efforts should be made to collect surface and
rain water near structures to prevent wetting of the soils during wet periods. During dry
periods, evaporation and transpiration remove water from soil and increase suction
leading to shrinkage. These effects can be limited through the use of pavement and
avoiding placement of vegetation around structures.
Engineered fill constructed from high plasticity clays will tend to shrink and swell in
response to the climate. In addition to deformation, this volume change results in a
time-dependent reduction in shear strength that must be accounted for in slope design.
If possible, high plasticity clays should not be used for the portions of embankments
exposed to fluctuation in water content. Swelling of high plasticity clay fill can be
avoided by compaction at high water content (i.e., wet of the line of optimums). Fill
compacted wet will have lower dry unit weight, lower shear strength, and higher
compressibility. High plasticity clay can be placed wet for structural fill below lightly
loaded buildings provided the lower bearing capacity is considered. Consistent relative
compaction is important to avoid differential settlement. Admixtures, such as cement or
lime, can also be mixed with high plasticity clay fill during construction to reduce
swelling potential.
5-9 HYDROCOMPRESSION.
Hydrocompression refers to the volume change of compacted soil when wetted

following construction. This phenomenon is especially problematic in deep fills
constructed from sandy clays and clayey sands (Brandon et al. 1990). Significant
settlement can occur in the regions of deep fill while net swelling may result in areas of
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relatively shallow fill depth. Damage to structures by hydrocompression tends to be

worst over locations where the fill depth is changing rapidly and strains are extensional
at the surface. The damaging effects of hydrocompression can be reduced by
increasing the relative compaction of the fill and/or increasing the compaction water
content.
The magnitude of hydrocompression can be predicted using the procedure described in

Brandon et al. (1990). Specimens of the fill material can be compacted and loaded
incrementally in one-dimensional consolidation to a range of total vertical stresses
corresponding to those present in the planned or existing fill. After the intended total
vertical stress is applied, the specimen is inundated with water and the volumetric strain
caused by wetting is measured. In this manner, the relationship between volumetric
strain caused by wetting and confining stress can be estimated. The expected
hydrocompression can be found by dividing the fill depth into thin sublayers (see Figure
5-5c) and determining the change in thickness from the corresponding strain.

Topic Reference
Duncan, J. M. and Buchiagnani, A. L. (1987). Engineering Manual for

Settlement Calculations Settlement Studies, CGPR #2, Center for Geotechnical Practice and
Research, Virginia Tech, 94 pp.
Federal Highway Administration (FHWA) (2017). Ground Modification

Vertical Drains Methods Reference Manual – Volume I, FHWA-NHI-16-027, Geotechnical
Engineering Circular 13, Washington D.C.
Vanapalli, S. K. and Lu, L. (2012). “A state-of-the-art review of 1-D heave

Volume Expansion prediction methods for expansive soils.” International Journal of Geotechnical
Engineering, 6, 15-41.
5-11 NOTATION.
Symbol Description
B Shortest dimension of foundation or loaded area
C1 Schmertmann coefficient to correct for the effects of embedment
C2 Schmertmann coefficient to correct for the effects of time (creep)
Ccε Modified compression index
ch Coefficient of consolidation in horizontal direction
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Symbol Description
Cr ε Modified recompression index
Ct Creep factor for coarse-grained settlement methods
cv Coefficient of consolidation in vertical direction
Cα Secondary compression index
Cαε Modified secondary compression index
dc Effective drainage diameter
dw Equivalent diameter of well or PVD
e0 Initial void ratio
Em Modulus of elasticity of mat
Es Modulus of elasticity of soil
Eu Undrained modulus
E G Relative stiffness for structures
Fn Radial drainage factor related to drain spacing
Fr Radial drainage factor related to well resistance
Fs Radial drainage factor related to soil disturbance (smear)
H Initial thickness in settlement calculations
H dr Drainage path length
Hi Thickness of each soil layer (may be listed without subscript)
H 't Total thickness of transformed soil system
Iz Schmertmann strain influence factor
I zp Schmertmann peak influence factor
kh Hydraulic conductivity in horizontal direction
Km Mat stiffness factor
ks Hydraulic conductivity of the disturbed zone
kv Hydraulic conductivity in vertical direction
l Distance between two points along a structure
L Longest dimension of a foundation or loaded area
LL Liquid limit
Lm maximum distance water must flow through a vertical drain
m Empirical exponent used to relate undrained shear strength to OCR
N 60 Standard Penetration Test corrected blow count
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Symbol Description
n Vertical drain spacing ratio
N Standard Penetration Test blow count
N' Average Standard Penetration Test value
N 'silty Standard Penetration Test blow count for saturated silty sands
OCR Overconsolidation ratio
qc Cone tip bearing resistance
q0 Applied stress at the base of the foundation or structure
q0− net Net vertical stress applied by the structure
qf Applied stress following removal of surcharge
qs Surcharge load
qw Discharge capacity of the drain
s Ratio of the disturbed zone diameter to the diameter of the drain
s Settlement
sc Primary consolidation settlement
ss Secondary compression settlement
t Time after start of consolidation
T Time factor
tm thickness of mat
tp Time required to finish primary consolidation
Tr Time factor for radial consolidation
Tv Time factor for vertical drainage
u0 Initial pore water pressure
Uc Combined degree of consolidation
U f +s Degree of consolidation following surcharge application
Ur Degree of radial consolidation
USRNC Undrained strength ratio for normally consolidated conditions
ux Excess pore water pressure
Uz Degree of compression
Uz Average degree of consolidation
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Symbol Description
wn Natural water content
z Depth along vertical drain
zi Layer thickness for settlement calculations
α Settlement correction factor
γw unit weight of water
δ l Angular distortion
δ max Differential settlement
∆H Change in layer thickness
∆ L Deflection ratio
∆σ z Change in total vertical stress
ε crit Critical strain for structural distress
εz Vertical strain
µ0 Influence factor associated with embedment of the load
µ1 Influence factor associated with problem geometry and Poisson’s ratio
νm Poisson’s ratio of mat
νs Poisson’s ratio of soil
σ 'p Preconsolidation stress
σ 'z Vertical effective stress
σ 'z 0 Initial or in situ vertical effective stress
σ 'zp Initial vertical effective stress at depth of Schmertmann peak influence factor
σ z − red Vertical stress reduction
ψ Matric suction
ω Tilt angle due to differential settlement
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SEEPAGE AND DRAINAGE
6-1 INTRODUCTION.
6-1.1 Scope.
This chapter discusses methods for analyzing seepage in soils and bedrock and
provides design guidance for drainage elements in structures and foundations. The
chapter provides a summary of available methods for analyzing the seepage regime,
descriptions and analysis methods for internal erosion mechanisms, and discussion of
seepage and internal erosion mitigation methods.
6-1.2 Background.
Seepage is the flow of water through interstitial voids of soil or rock. Seepage is driven
by differential potential energy of water (i.e., hydraulic head) acting across the soil or
rock mass, resulting in the flow of water from higher to lower potential energy.
Seepage is a common phenomenon in geotechnical engineering applications and can

occur as natural groundwater flow, seepage through dams and levees or their
foundations, or flow into excavations extending below the groundwater surface. While
the movement of water occurs in unsaturated soils above the groundwater table, this
chapter deals only with seepage that occurs under saturated conditions.
Undesirable consequences of seepage can include internal erosion, excessive water

loss or accumulation, and excessive pore water pressures. Under the right conditions,
seepage can result in erosion of soil or rock, or internal erosion. Several different
mechanisms of internal erosion have been identified that can occur by one of several
mechanisms. In cases where seepage quantities are large, problems can occur,
including: excessive water loss from reservoirs, flooding of excavations, and unstable
ground due to excess moisture. Seepage may also result in excess water pressures
under structures leading to instability and uplift forces.
This chapter also discusses a number of strategies and methods for mitigating the
undesirable effects of seepage discussed in the previous paragraph. Each method
utilizes one or a combination of three basic strategies: 1) blocking or lengthening the
seepage pathway, 2) draining the excess water pressures in a controlled manner, and
3) filtering the seepage to block the transportation of soil particles.
6-2 SEEPAGE ANALYSES.
6-2.1 Hydraulic Head.
Hydraulic head is a measure of the energy of the water acting on or within geologic
media, expressed in terms of length units and referenced to a consistent datum. The
total hydraulic head ( ht ) at any given point is composed of three components: the
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pressure head ( hp ), the elevation head ( hz ), and the velocity head ( hv ). In geologic
media, the velocity head is typically negligible and the total head is expressed as:
u
ht = hp + hz = +z (6-1)
γw
where:
u = the water pressure at the point of interest,
γ w = the unit weight of water, and
z = the elevation of the point of interest above the elevation datum.
The total hydraulic head ( ht ) at a point is the height above the elevation datum that
water would rise in a piezometer if the tip of that piezometer were located at the point of
interest as illustrated in Figure 6-1a. The total hydraulic head will vary within a flow
regime unless conditions are completely static (i.e., no flow is occurring).
As an example, consider the pressurized tank in Figure 6-1b. Piezometers have been
set at two points in the side of the tank and the water rises in the piezometer above the
elevation of the tank due to the pressure in the tank. The pressure head ( hp ) is the
height that the water rises above the point of interest and the elevation head ( hz ) is the
height of the point of interest above the datum that has been set below the tank. The
total head ( ht ) is the combined heights of hp and hz and is the total height that the
water rises above the set datum. Since there is no flow within the tank, ht is constant
throughout the tank although hp and hz vary with elevation.
Figure 6-1 Example of the Components of Hydraulic Head

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6-2.2 Darcy’s Law and One-Dimensional Flow.
The principles of seepage mechanics and analysis of the seepage regime are best
illustrated by considering the example of one-dimensional flow illustrated schematically
in Figure 6-2. A soil-filled conduit with a cross section of area ( A ) and a length ( L ) is
attached to water reservoirs with different total heads. The difference in water height in
piezometers at each end of the conduit indicates the differential total head ( hL ) or head
loss across the soil. This head also represents the amount of energy that must be
dissipated as the water flows through the soil. The differential head creates a hydraulic
gradient ( i ), which is defined as:
hL
i= (6-2)
L
where:
i = the hydraulic gradient,
hL = the differential total head (or head loss), and
L = the length over which hL occurs.
The hydraulic gradient forces the water to flow through the soil at a volumetric flow rate
( q ) that is sufficient to create the head loss associated with hL . The volumetric flow
rate is defined as the flow volume that passes through the soil per unit of time.
Figure 6-2 One-Dimensional Flow through Soil
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One-dimensional flow is governed by Darcy’s Law:
q = kiA (6-3)
where:
q = the volumetric flow rate through the soil,
k = the hydraulic conductivity of the soil,
i = the hydraulic gradient across the flow region, and
A = the cross sectional area of the flow region perpendicular to the flow direction.
If the flow region has a constant height and an extended width (perpendicular to the
page), the flow area ( A ) can be defined by the height of the flow region times a unit
width. In this case, the flow rate per unit length of the model is:
q = kiy (6-4)
where:
q = the flow rate per unit length of the model,
k = the hydraulic conductivity of the soil,
i = the hydraulic gradient across the flow region, and
y = the height of the flow region.
The discharge velocity ( vd ) can be calculated by dividing the volumetric flow rate by the
cross-sectional area:
q
v=
d = ki (6-5)
A
It should be noted that vd is not a true particle velocity but rather the velocity based on
the total area of the flow region. Since the water only flows through the pore space of a
soil or rock, a water particle actually flows faster through than vd . The seepage velocity
( vs ) which measures how fast a water particle moves as a result of the hydraulic
gradient, is calculated as:
vd 1+ e
v=
s = vd (6-6)
n e
where:
n = the porosity of the soil and
e = void ratio.
Darcy’s law is valid for conditions where the seepage flow is laminar, which includes
most cases of seepage through soils. High velocity flows through coarse, open-graded
gravels may fall in the transition between laminar and turbulent flow. Turbulent flow
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results in more resistance to seepage and thus a lower volumetric flow rate than
predicted by Darcy’s law.
6-2.3 Two-Dimensional Seepage.
Analysis of seepage in a two-dimensional regime requires expansion of Darcy’s law.

The governing equation for steady state, two-dimensional flow is the LaPlace equation:
∂2h ∂2h
+ 0
= (6-7)
∂x 2 ∂y 2
where:
h = total hydraulic head and
x and y = coordinate directions.
The LaPlace equation is derived by applying the conservation of mass principle to an

element of soil, thereby using equilibrium to spatially link changes in total head within
the flow region. The first term in Equation (6-7) represents the change in hydraulic
gradient in the x direction through an element of soil while the second term represents
the change in gradient in the y direction in the same element. Derivation and further
discussion of the LaPlace equation can be found in books on groundwater, such as
Bear (1979) or Freeze and Cherry (1979).
Solutions to the LaPlace equation can be performed through (1) graphical solutions
such as flow nets, (2) closed-form equations such as the method of fragments or the
U.S. Army Corps of Engineers blanket theory equations (USACE 2000), or (3)
numerical solutions such as finite element analyses.
6-2.4 Flow Nets.
Flow nets are a relatively quick graphical solution tool for analyzing two-dimensional
flow regimes using few resources; namely a pencil, paper, and a good eraser. The act
of drawing of a flow net also helps the engineer to visualize and understand the
behavior of seepage flows. The understanding gained by drawing a flow net is often
deeper than that gained by numerical analyses.
A flow net for seepage in an isotropic soil layer beneath an impermeable dam is
presented in Figure 6-3. The flow region is divided into flow elements, most of which
resemble squares or are as close to square as possible. The long-short dashed lines
represent equipotential lines or contours of constant total head within the soil. Note that
the level upstream ground surface is an equipotential line since the reservoir level
applies a constant total head along this boundary. The level downstream ground
surface is also an equipotential line since the pressure is constant on the surface (equal
to zero pressure head) and the elevation is constant (constant elevation head). If the
downstream exit face were sloped and not submerged, the total head would not be
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constant and it would not be an equipotential line. The short dashed lines are flow lines,
which represent average paths that water particles will follow while flowing through the
soil. The bottom impervious boundary and the bottom of the impervious structure are
also flow lines.
Figure 6-3 Flow Net for Seepage Through an Isotropic Soil Layer Beneath an
Impermeable Dam
6-2.4.1 Drawing Flow Nets.
Flow nets for seepage through soil with isotropic permeability must comply with the
following rules in order to be correct:
a. Flow lines and equipotential lines should intersect at right angles.

b. The flow elements formed between the flow lines and equipotential should
resemble curvilinear squares. A circle can be inscribed in a curvilinear square
and touch all four boundaries of the flow element. More guidance on the shape
of admissible flow elements can be found in USACE (1986).
c. An impermeable boundary will act as a flow line. Common examples are the top
of an underlying layer and the bottom of an impermeable dam or levee.
d. Submerged inflow or outflow boundaries through which seepage passes are
equipotential lines with head equal to the water level elevation.
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e. Where the flow is unconfined (such as through an embankment) the top flow line
will be the phreatic surface. Points on this line will have pressure equal to zero
and, consequently, the total head is equal to the elevation of the line (see
Equation 6-1).
f. Along a phreatic surface for unconfined flow, equipotential lines will intercept the
phreatic surface at equal vertical intervals.
Flow nets constructed according to the above rules will have the following
characteristics:
a. Each flow channel, bounded by two adjacent flow lines, will convey the same
amount of flow as the other flow channels in the flow net.
b. Each total head drop, bounded by two adjacent equipotential lines, represents
the same decrease in total head as the other head drops in the flow net.
c. The flow elements can be subdivided into regions representing partial head
drops and partial flow channels.
Flow nets can be drawn for flow through non-homogenous soil profiles and soil with
anisotropic hydraulic conductivity. In stratified soil profiles, the flow will be dominated by
the permeable layers. If the ratio of a layer’s hydraulic conductivity compared to that of
the most permeable layer exceeds 10 to 100, the layer can be considered impermeable.
If this ratio is less than 10, the flow will be through both layers. However, the flow lines
and equipotential lines will be deflected at the interface as illustrated in Figure 6-4.
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Figure 6-4 Deflection of Flow at a Boundary with Changed Permeability
Flow through soil with anisotropic permeability can be transformed into an equivalent
isotropic region using the transformation factor ( a ):
kmax
a= (6-8)
kmin
where:
a = isotropic transformation factor,
kmax = the maximum hydraulic conductivity in anisotropic soil, and
kmin = the minimum hydraulic conductivity in anisotropic soil.
The dimensions of the flow region are transformed by dividing all of the dimensions of
parallel to the direction of kmax by a . A flow net is drawn for the transformed system
following the rules for isotropic hydraulic conductivity. For example, if k is largest in the
horizontal direction, then all of the x-coordinates will be divided by a in the transformed
system. If needed, the flow region and flow net can be transformed back to “real space”
by multiplying the dimensions parallel to the direction of kmax by a .
6-2.4.2 Interpreting Flow Nets.
Once drawn, flow nets can be used to calculate a number of properties including:
seepage quantities, pore pressures, uplift forces, and hydraulic gradients. The
volumetric flow rate through a flow net section can be calculated as:
N 
q =k ⋅ hL ⋅  f  ⋅ W (6-9)
 Nd 
where:
hL = the total differential head or head loss across the flow net,
k = the isotropic hydraulic conductivity (use kmax kmin for transformed flow nets),
N f = the number of flow channels in the flow net,
N d = the total number of equipotential (head) drops in the flow net, and
W = the width of the system perpendicular to the page, often taken as a unit width.
The ratio N f N d from the flow net is sometimes referred to as the shape factor ( SF ).
The shape factor incorporates the influence of geometry into the calculation of flow.
Two flow nets with a different number of flow lines but the same value of SF will predict
the same flow rates, total heads, and pore pressures.
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The flow net divides the total head loss across the system into N d equal head drops.
The head loss associated with each head drop is:
h
∆hL =L (6-10)
Nd
where:
∆hL = the total differential head or head loss across one head drop,
hL = the total differential head or head loss across the flow net, and
N d = the total number of head drops in the flow net.
The total head at any point within the flow net can be calculated by reference to the
known total head at either the upstream or downstream boundary. The change in head
from the boundary for any point is equal to the number of head drops from the boundary
multiplied by ∆hL . By knowing the total head and elevation at a point of interest, the
pore water pressure at any point can be calculated from the flow net as:
u hpγ=
= w ( ht − hz ) γ w (6-11)
where:
hp = the pressure head at the point in question,
ht = the total head at the point in question,
hz = the elevation head at the point in question, and
γ w = the unit weight of water.
The hydraulic gradient can be calculated between any two points in the flow region by
dividing the change in total head that occurs between two points by the distance over
which the head loss occurs. When calculating gradients, it may be useful to subdivide
flow net sections for more precision.
Figure 6-5 presents an example flow net with example calculations for discharge, uplift
pressure, and hydraulic gradient.
6-2.5 Closed-Form Equations.
The method of fragments and blanket theory are two closed-form solutions for
calculation of seepage flow below water retaining structures. The method of fragments
(Pavlovsky 1956, Harr 1977) subdivides the flow region into fragments of standard
geometry. Based on the geometry, the SF for each fragment is determined along with
the overall SF for the problem. The overall flow rate and pore pressures at particular
points can be determined from the results. The blanket theory equations are based on
the method of fragments and are particularly useful for seepage analyses of levees.
These equations are specifically derived for a levee foundation condition consisting of a
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low-permeability “blanket” layer overlying a high-permeability “foundation” layer. The

potential for soil heave occurring on the landside of the river can be readily assessed.
Both methods can be implemented in a spreadsheet application. For further information

on the method of fragments and blanket theory equations, see Holtz et al. (2011) and
Appendix B of USACE (1986), respectively.
Other common solutions have been developed for: (1) flow, heave potential, and exit
gradients into excavations, (2) relief well design, and (3) dewatering well design. These
specific solutions are presented in the later sections of this chapter.
Figure 6-5 Flow Net Example Calculations
6-2.6 Numerical Seepage Analysis.
Numerical analysis, such as finite element or finite difference, is the appropriate tool for
most seepage analysis problems. These methods are user-friendly and allow easy
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input of soil properties and complex geometric configurations, while providing rich
graphical output. Numerical analyses also can be extended to three-dimensions and
used to model unsaturated soils and transient flow conditions. However, those topics
are beyond the scope of this chapter. The graphical and analytical methods discussed
in Sections 6-2.4 and 6-2.5 provide an important means of validating numerical seepage
models.
Finite element analysis (FEA) is the most common numerical approach used for
seepage analysis, and this section is written from the perspective of FEA. Other
numerical approaches, such as finite difference, will also provide suitable results but
may use slightly different terminology.
6-2.6.1 General Concepts of Finite Element Seepage Analysis.
In finite element analysis, the flow region is divided into areas or volumes (referred to as
elements) within which the flow of water can be easily defined. Elements are formed by
connecting points in space (referred to as nodes) with lines. Two-dimensional elements
are often three-node triangles or four-node quadrilaterals. Within each element the flow
is defined with a system of equations that relate the hydraulic head at each node with
the hydraulic gradient and flow within the element. These equations are described in an
element matrix by linking the values of the common nodes. The equations (matrices)
for each of the elements are assembled into a global matrix that represents a set of
equations that define the flow through the entire system. For each node there is one
equation and one unknown value for each node.
In the simplest form of element, the direction and magnitude of the hydraulic gradient
throughout the element are constant. This results in the hydraulic head varying linearly
along the element boundaries and within the elements themselves. In more advanced
element types, the head varies according to a polynomial equation. Before solving the
problem, the one unknown for each node is either (1) the total head at the node or (2)
the total flow into and out of the system associated with the node. In general, nodes
within the body of the problem and along no-flow boundaries have unknown head and
total flow of zero (i.e., flow in equals flow out). At boundaries where flow enters the
system, the flow is unknown but the head is generally specified.
6-2.6.2 Boundary Conditions.
Boundary conditions describe the head, pressure, and flow conditions at the boundaries
of the model. Table 6-1 describes the most common boundary conditions used in basic
finite element models. Table 6-2 illustrates the application of boundary conditions to
finite element models.
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6-2.6.3 Interpreting Output.
The primary mathematical result of a finite element analysis are values of total hydraulic
head and nodal flow for each node in the finite element mesh. By post-processing the
total head and flow results from the FEA, the hydraulic gradient, hydraulic velocity,
volumetric flow rate, pressure head, and uplift force along a boundary segment can be
obtained. Most of the commercially available finite element seepage analysis software
have post processors that will calculate these values through interpolation algorithms.
Several of these interpretations are discussed below.
Table 6-1 Common Finite Element Analysis Boundary Conditions

Boundary General
Specific Description Common Usage
Condition Description
Flow allowed to Submerged boundaries (total head =
enter or exit the Total head held at a water surface elevation)
Constant
system along the constant value along the Horizontal seepage exits (total head =
Specify head
Head
boundary. boundary ground surface elevation)
Nodal heads are Sides of numerical models
known.
The pressure head held at
Constant Nodal flow is Seepage exits (pressure head = 0)
a constant value along the
Pressure calculated during Internal drains (pressure head = 0)
boundary.
the analysis.
No-Flow Total nodal flow is Boundaries with impermeable soil,
Flow is not allowed across
(impermeable specified. May be rock, or structures
the boundary.
boundary) specified as zero Sides of numerical models
Specify flow
for no-flow The inflow or outflow rate

boundaries from the system is
Nodal Flow Nodal heads are Injection or extraction well
specified for a node (often
unknown. an internal node).
Nodal heads are
calculated during Inflow along an external Rainwater or other infiltration along a
Infiltration
the analysis. boundary is specified. surface
The boundary is set to be either a no-flow or Seepage exit areas where the phreatic
Unknown or
constant pressure boundary, depending on the surface is unknown, such as the
Variable
results of the analysis. downstream slope of a dam or levee
6-2.6.3.1 Pore Water Pressure.
The pore water pressure is calculated from the total head and the elevation using
Equation 6-11. Generally, commercial software will calculate the hydraulic pressure for
each node and provide a contour map of the pressure throughout the flow region.
6-2.6.3.2 Uplift Force.
Uplift force due to hydraulic pressure on a structure or mass of soil is calculated by

integrating the hydraulic pressures along a specified set of the boundary segments in
the model. A simple estimate of the uplift force can be calculated by plotting the pore
water pressure contours and assigning a representative area along the boundary
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segment to each contour interval. The total uplift force can be calculated by summing
product of the interval length and the representative pressure for each interval.
Table 6-2 Examples of Boundary Condition Usage

Schematic of Seepage Regime Model and Boundary Conditions
Impermeable Dam
Coffer Dam
Permeable Dam or Levee
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6-2.6.3.3 Hydraulic Gradients.
Hydraulic gradients are calculated for each element based on the hydraulic heads at the
element nodes. Strictly speaking for simple elements, the hydraulic gradient in a given
direction should be constant within each element. However, commercial software will
often interpolate contours of hydraulic gradient based on the calculated head at each
node. For higher order elements, the hydraulic gradient in a given direction may vary
within the element. In these cases, the calculated gradients may be interpreted by
plotting its variation within the flow region.
As a note of caution, the calculated hydraulic gradient will vary depending on the size of
the element used near sharp corners in the flow region (often termed singularities). In
this case, the calculated hydraulic gradients will increase as the element size
decreases. As the element dimension approaches zero, the calculated hydraulic
gradient may approach infinity. At these locations, the hydraulic gradients should be
calculated over a distance consistent with the mechanisms of concern and known
ground conditions. For example, the hydraulic gradient at the toe of a levee resting on a
blanket layer of low k soil should be calculated across the thickness of the blanket.
6-2.6.3.4 Discharge Velocity.
Discharge velocity is calculated from the hydraulic gradient and the hydraulic
conductivity assigned to the element using Equation 6-5.
6-2.6.3.5 Seepage Flow Volume.
Seepage volumes are calculated for elements using Darcy’s law (Equation 6-3). For
flow across a model boundary, the flow rate through each element is calculated, using
the vector component of hydraulic gradient perpendicular to the boundary, the hydraulic
conductivity assigned to the element, and the area of the element boundary along the
model boundary. The flow rate across a portion of an FEA model can be assessed by
calculating the flow across a line using a similar methodology applied to the area
associated with the elements intersected by the line.
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6-3 HYDRAULIC CONDUCTIVITY (COEFFICIENT OF PERMEABILTY).
One of the most varied soil properties in geotechnical engineering is hydraulic

conductivity, which can be defined as the discharge velocity of water through a unit area
under a unit hydraulic gradient. Common hydraulic conductivity values can range from
less 10-8 cm/sec for high plasticity clay to in excess of 1 cm/sec for open graded
gravels; a range of over 8 orders of magnitude. Small changes in soil gradation,
especially changes in the fines content, can result in significant variation in hydraulic
conductivity.
Terminology regarding hydraulic conductivity is varied across the profession. The term
coefficient of permeability is used as a synonym for hydraulic conductivity in literature.
The practicing geotechnical community commonly uses the term “permeability”
interchangeably with hydraulic conductivity. However, this is technically incorrect as
permeability is a property of the porous media alone and does not consider the viscosity
of the permeant fluid. Hydraulic conductivity is the preferred term, but both hydraulic
conductivity and permeability are used in this manual.
Hydraulic conductivity is most often required for in situ soil conditions. Hydraulic
conductivity can be assessed for these conditions by several strategies: 1) laboratory
testing, 2) field testing, and 3) empirical correlations including equations, charts, and
tables. Each of these are discussed in the following subsections.
6-3.1 Laboratory Testing.
Laboratory tests can be performed on intact samples or reconstituted samples. Details

of laboratory testing procedures are presented in Section 3-2.7. While laboratory tests
can measure the hydraulic conductivity of a wide range of soils, the limitations of
laboratory testing must be acknowledged. First, laboratory tests use a small sample of
soil or rock and usually test the vertical permeability because the samples are obtained
from boreholes. As a result, laboratory test results may not be representative of the
large-scale properties of a soil deposit if the layering and structure of the deposit is not
considered. Natural soils usually exhibit anisotropy with the horizontal permeability
being larger than the vertical. Thus, laboratory tests are likely to result in lower values
than are appropriate for many types of analyses. Finally, intact samples of coarse-
grained soils cannot be obtained using normal sampling procedures. Laboratory tests
on reconstituted specimens of these materials are likely no more reliable than
correlations. With the above in mind, laboratory permeability testing is most
appropriately reserved for reconstituted samples for applications such as fill materials,
cutoff wall backfills, pond liners, and filter materials.
6-3.2 Field Testing.
Two classes of field hydraulic conductivity testing are borehole tests and field pumping
tests. Details of field testing procedures are presented in Chapter 2. Borehole tests are
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effective in measuring the permeability of the soil in the general area of the borehole.
To measure permeability characteristics over a broader area, a pumping test can be
performed.
6-3.3 Empirical Relationships for Hydraulic Conductivity.
Numerous empirical and semi-empirical relationships have been developed for

correlating hydraulic conductivity with other soil properties (predominantly grain size and
gradation). The simplest of these relationships relate soil type to typical values of
ranges of k . Figure 6-6 correlates k with soil classification types for various unit
systems.
The United States Bureau of Reclamation (USBR 2014) has determined typical values
for the horizontal hydraulic conductivity of natural soil and rock deposits based on field
testing as indicated in Table 6-3 and the vertical hydraulic conductivity of embankment
fill materials based on laboratory testing as summarized in Table 6-4.
cm3/cm2/sec (cm/sec)
101 1 10- 1 10- 2 10- 3 10- 4 10- 5 10- 6 10- 7 10- 8
ft3/ft2/day (ft/day)
105 104 103 102 101 1 10- 1 10- 2 10- 3 10- 4 10- 5
3 2
ft /ft /min (ft/min)
101 1 10- 1 10- 2 10- 3 10- 4 10- 5 10- 6 10- 7 10- 8
2 2
gal/ft /day (gal/ft /day)
105 104 103 102 101 1 10- 1 10- 2 10- 3 10- 4
3 2
meter /meter /day (m/day)
104 103 102 101 1 10- 1 10- 2 10- 3 10- 4 10- 5
relative permeability
Very high High Moderate Low Very low
Soil Clean gravel Clean sand, clean sand Fine sand, silty sand Silt, clay, and sand-silt- Massive clay, no
types (GP) and gravel mixes (GW, and gravel mixes (SP, SM, clay mixes, organic silts, soil joints or
GP, SW, SP, SM) GM, GW-GM, GP-GM, organic clays (GM, GC, SM, other macropores
SW-SM, SP-SM) SC, MH, ML, ML-CL, OL, OH, (CL, CH)
GW-GC, GC-GM, SW-SC,
SP-SC, SC-SM)
Figure 6-6 Variation of Hydraulic Conductivity with Soil Type for Various Unit
Systems (after Freeze and Cherry 1979)
Table 6-3 Typical Ranges of Horizontal Hydraulic Conductivity for Natural Soil
and Unfractured Rock Deposits (after USBR 2014)
Soil Type kh (cm/sec) Rock Type kh (cm/sec)
Gravel, open-work >2 Sandstone, medium 1x10-4 to 2x10-1

Gravel (GP) 2x10-1 to 2. Sandstone, silty < 5x10-3
Gravel (GW) 1x10-2 to 1 Limestone < 1.5x10-2
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Sand, coarse (SP) 1x10-2 to 5x10-1 Granite, weathered 2x10-4 to 1x10-5

Sand, medium (SP) 1x10-3 to 1x10-1 Schist < 2x10-3
Sand, fine (SP) 5x10-4 to 5x10-2 Tuff < 1x10-3
Sand (SW) 1x10-4 to 5x10-2 Gabbro, weathered 5x10-5 to 5x10-4
Sand, silty (SM) 1x10-4 to 1x10-2 Basalt < 5x10-5
Sand, clayey (SC) 1x10-6 to 1x10-3 Dolomite < 5x10-6
Silt (ML) 1x10-6 to 1x10-3 Gneiss < 2x10-6
Clay (CL) < 3x10-6
Note: Materials with no lower bound indicated can range from practically impervious to the upper limit
indicated in the table.
Table 6-4 Typical Range of Vertical Hydraulic Conductivity for Compacted Soil
in Embankments (after USBR 2014)
Embankment Core Materials Embankment Shell Materials
Unified Soil Classification k h (cm/sec) Unified Soil Classification k h (cm/sec)

GM-SM < 1x10-5 GP 2x10-3 to 1.0
GM or GC < 1x10-5 GW 1x10-3 to 1x10-1
SP-SM < 1x10-5 GP-SP 1x10-3 to 5x10-1
SM < 1x10-5 GW-SW 5x10-4 to 5x10-3
SM-SC < 3x10-6 GM 1x10-5 to 5x10-4
SC < 3x10-6 SP (medium to coarse) 1x10-2 to 2x10-2
ML < 1x10-5 SP (fine to medium) 5x10-3 to 1x10-2
ML-CL < 1x10-6 SP (very fine to fine) 5x10-4 to 5x10-3
CL < 1x10-6 SW 3x10-4 to 5x10-3
MH < 1x10-7 SP-SM 1x10-5 to 1x10-3
SM 1x10-5 to 5x10-4
Note: Materials with no lower bound indicated can range from practically impervious to the upper limit indicated in the
table.
The size of the pore spaces is one of the most important factors controlling the hydraulic
conductivity of a soil, especially for coarse-grained materials. The effective diameter or
effective grain size ( Dα ) is the grain size that has the primary influence on the average
pore size of the soil. In terms of the effects on hydraulic conductivity, the α refers to a
particular percent passing on the grain-size distribution and typically has been assigned
a value of 5, 10, 15, or 20. In other words, the effects of pore size on hydraulic
conductivity has been found to correlate best with particle diameters corresponding to
the 5 to 20% passing size. Correlations that relate hydraulic conductivity to an effective
grain size or grain-size distribution can be expressed as (Kenney et al. 1984):
k = βα Dαx (6-12)
where:
k = hydraulic conductivity,
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βα = empirical or semi-empirical coefficient,

Dα = effective grain size,
α = percent passing corresponding to effective grain size, and
x = exponent - theoretically equal to 2 and empirically slightly above 2.
Most of the published correlations for the hydraulic conductivity of coarse-grained soils
can be expressed in terms of Equation 6-12. These correlations are summarized in
Table 6-5. Some of the relationships also account for the effect of void ratio on k .
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Table 6-5 Estimating Hydraulic Conductivity based on Effective Grain Size
Source Reported Application α β α (cm/sec/mm2) x
Kenney et al. Sand and fine gravel,

5 1 2
(1984) Cu = 1 to 12
Loose Sands with D10
Hazen (1892, Varies by source from 0.01 to 10
between 0.01 and 0.3 10 2
1911) Often taken to equal to 1
cm
Slichter (1905) Sands with D10
 9.3071e 
and between 0.01 and 0.5 10 0.01047 exp  2
(McCook 2010)  1+ e 
cm
0.7825
 e3 
Chapuis (2004) Sand 10 2.4622 1.565
 1+ e 
 
Sand, assuming uniform  e3 
5.52
Carrier (2003)
spheres
10  1+ e  2
 
Sherard et al. Sand and gravel with
15 Average = 0.35, range = 0.2 to 0.6 2
(1984) low fines content
Notes: k is estimated in cm/sec, Cu = coefficient of uniformity = D60 D10 , e = void ratio
The Kozeny-Carman equation (Carrier 2003) can be used to account for the effect of
the entire grain-size distribution and the particle shape:
2
 
cm  100%   1   e3 
2
k 1.99 ×10
=    2   (6-13)
s ⋅ mm 2  fi   S  1+ e 
 ∑ Dli0.404 + Dsi0.596 
where:
k = hydraulic conductivity (cm/sec),
fi = fraction of particles (by mass) between two adjacent sieve sizes,
Dli = the particle size of the coarser sieve (mm),
Dsi = the particle size of the finer sieve (mm),
S = surface area factor ranging from 6 for spheres to 8.5 for angular particles, and
e = void ratio.
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For a given soil, the ratio of hydraulic conductivities under two different void ratios can
be considered by (Kozeny 1927):
k1 1 + e2 e13
= (6-14)
k2 1 + e1 e23
where:
k1 = hydraulic conductivity for void ratio, e1 , and
k2 = hydraulic conductivity for void ratio, e2 .
For fine-grained soils, the hydraulic conductivity would be expected to decrease as

liquid limit, plasticity index, or fines content increase. Figure 6-7 shows that hydraulic
conductivity decreases as the fine content (percent passing the No. 200 sieve)
increases. For a given fines content, the range of kv is 2.5 to 3 orders of magnitude
with sandy and silty soils having higher kv than clayey soils.
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Figure 6-7 Variation of Hydraulic Conductivity with Fines Content

(after California Department of Water Resources 2013)
6-3.4 Anisotropy.
More often than not soils exhibit anisotropy with respect to hydraulic conductivity. In
natural soils this is usually a consequence of depositional layering of the soil and results
in horizontal hydraulic conductivity greater than in the vertical direction. However, in
some cases, often related to the formation and filling of vertical cracks, the vertical
hydraulic conductivity can be greater than the horizontal. In engineered fills, anisotropy
can form as a consequence of soil variation between lifts, differential compaction
between the top and bottom of lifts, and planes that form between lifts. Table 6-6 and
Table 6-7 present typical values for anisotropy in natural soil deposits and compacted
fill.
Anisotropy can greatly affect the seepage behavior in a soil deposit and can have a
significant effect on the calculated pressures, flows, and hydraulic gradients in the
seepage regime.
Table 6-6 Typical Values of Anisotropy in Natural Soils (after USBR 2014)
k h kv
Formation Ratio depends on:
Lower Upper
Stratified Deposits 10 1000 Range of k for laminations
Intact Soil or Rock 1 3 Particle shape and orientation
Fractured Bedrock 0.1 10 Arrangement and orientation of apertures and joints

Orientation of fissures and cracks that form during
Loess 0.02 2
consolidation and desiccation
Table 6-7 Typical Values of Anisotropy in Engineered Fill (after USBR 2014)
k h kv
Fill Zone or Method
Lower Upper
Core Zone, USBR Compaction Procedures 4 9
Core Zone, Standard Compaction Procedures 9 36
Hydraulic Fill 64 225
Embankment Shell, USBR Compaction Procedures 4 9
Embankment Drains, USBR Compaction Procedures 1 4
Note: USBR compaction procedures are based on Standard Proctor (ASTM D698).
Requirements for compaction water content and relative compaction and/or relative
density vary based on grain-size distribution and embankment height.
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6-4 INTERNAL EROSION.
About half of dam failures and accidents can be attributed to internal erosion through
the foundation and/or through the embankment or along a penetration through the
embankment (e.g., Foster et al. 2000). Internal erosion is a generic term that describes
erosion of particles caused by water seeping through a body of soil or rock. The water
may be seeping through the interstitial voids of a soil or rock mass, or may be flowing
along pathways of preferential flow (cracks or other defects).
Terminology describing the various mechanisms of internal erosion has evolved in

recent years as understanding of the mechanisms of erosion have developed.
Nomenclature for these mechanisms has been and continues to be inconsistent in
practice and in the literature due to this rapid evolution of understanding and
nomenclature. For example, the terms “piping” and “seepage-related erosion” have
been used as generic terms for internal erosion, and the term “internal erosion” has
been used to denote a specific internal erosion mechanism.
In 2014, the International Commission on Large Dams (ICOLD) adopted a system of

nomenclature describing the various mechanisms of internal erosion. The ICOLD
nomenclature is summarized in the following sections along with additions to the
nomenclature where necessary.
6-4.1 Heave.
Effective stress heave (a.k.a., quick condition) is the uplift of a mass of coarse-grained
soil due to a high hydraulic gradient acting on soil particles at an unprotected exit.
Seepage forces developed through viscous drag tend to lift the soil mass, resulting in
uplift or a quick condition, as illustrated in Figure 6-8a.
Coarse-grained soils with high vertical hydraulic exit gradients are susceptible to
effective stress heave. These gradients may occur at the base of deep excavations into
sand and at the toe of hydraulic structures founded on sand.
Total stress heave (a.k.a., blowout) is the uplift of a mass of low-permeability soil due to
high hydraulic pressure in an underlying aquifer as shown in Figure 6-8b. When the
pressure beneath a layer with low hydraulic conductivity exceeds the total weight of the
layer, uplift occurs and often results in cracking of the upper layer. Total stress heave
occurs in fine-grained soils with underlying aquifers, such as deep excavations into clay
and fine-grained blankets.
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Figure 6-8 Heave – (a) Effective Stress and (b) Total Stress
6-4.2 Erosion and Stoping.
Backward erosion piping (a.k.a., classical piping) is the successive removal of soil
particles at an unprotected exit resulting in the formation of an open pathway or pipe
that progresses toward the source of the seepage. A stable roof prevents collapse of
the pipe and allows its progression, as illustrated in Figure 6-9a. The toes of dams and
levees are especially susceptible to backward erosion piping, along with unprotected
exits, such the ground surface, internal voids or pathways, or defects in a conduit or
outlet.
General backward erosion (a.k.a., progressive sloughing or internal migration) is the

successive removal of soil particles at an unprotected exit resulting in progressive
sloughing of a slope. This type of erosion initiates similar to backward erosion piping,
but the lack of a “roof” prevents the progression of a pipe (Figure 6-9b). Slopes and
embankments consisting of coarse-grained soils with high seepage flows are
susceptible to general backward erosion.
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Stoping (a.k.a., sinkhole) is the near-vertical progression of a void caused by

successive collapse into a cavity as shown in Figure 6-9c. The cavity is often the result
of another internal erosion mechanism. Stopes often manifest as sinkholes at the
ground surface and occur in embankments with moderate to low cohesive strength.
Concentrated leak erosion (a.k.a., scour) is erosion that occurs along a concentrated
flow path and is caused by shear forces imposed by the flowing water (Figure 6-9d).
Concentrated leaks may be cracks within the soil or rock, gaps between soil and a
conduit or structure, or other pathways of low flow-resistance capable of carrying
eroded soil particles. Conditions that are susceptible to concentrated leak erosion
include cohesive embankments, outlet pipes and structures in embankments, and dam
and levee fill placed along steep rock abutments.
Figure 6-9 Erosion and Stoping Mechanisms

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Contact erosion (a.k.a., scour) is erosion that occurs along a contact between a highly
permeable material and an erodible soil. Contact erosion is caused by shear forces
imposed by high seepage velocities in the highly permeable material as illustrated in
Figure 6-9e. The contacts between erodible soil and open-graded gravel or open joints
in bedrock are locations susceptible to this type of erosion.
6-4.3 Internal Instability.
Soils can be internally unstable such that fine-grained particles erode from within a
framework or “skeleton” of coarse-grained particles as illustrated in Figure 6-10. The
process is called suffusion if the coarse-grained particles are in contact with each other
before erosion. Thus, no volume change results from suffusion. The process is called
suffosion if the coarse-grained particles are not in contact with each other before
erosion. Suffosion results in volume change or collapse. Well-graded, gap-graded, and
glacial till soils can be susceptible to either suffusion or suffusion.
Figure 6-10 Internal Instability – (a) Suffusion and (b) Suffosion
6-5 SEEPAGE AND INTERNAL EROSION MITIGATION METHODS.
6-5.1 Problems and General Strategies.
Seepage occurs through all earthen dam and levee embankments and their
foundations, into excavations below the water table, and below other structures
subjected to differential water pressure conditions. In many cases, the quantity of
seepage is such that it poses no adverse consequences or risk to the structure.
However, if seepage is excessive or the pressures and forces associated with the
seepage are too great, mitigation of the seepage issues may be required. Problems
associated with seepage can be classified into three categories:
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a. Excess volumetric flow rate of seepage

Undesirable consequences of excess flow include the loss of valuable water from
a reservoir, flooded or soft ground in excavations, and wet ground conditions
below a dam or levee that prevent activities or usage of the land.
b. High pore water pressures

High pore water pressures can result in unacceptable uplift forces beneath or
within dams and levees that lead to instability with respect to sliding or
overturning. Excessive water pressure forces can also act on buildings, retaining
walls, and other appurtenant structures.
c. Internal erosion potential

Some seepage conditions may create a condition where internal erosion is likely
through one of the mechanisms described in Section 6-4. Internal erosion
problems are often tied to high pore water pressures. Assessment of internal
erosion potential should be paired with an evaluation of the severity of seepage.
Table 6-8 provides a means to evaluate seepage severity and assess if further
investigation is required. This approach recognizes that the flow rate must be
sufficiently high for problematic internal erosion to occur
Table 6-8 Seepage Severity Categories (after Duncan et al. 2011, USACE 1956)
q hL W Severity of
Seepage Remediation Needed
(cfs per foot of head per foot of levee) Seepage
> 2.2x10-4 Heavy Yes
1.1x10-4 to 2.2x10-4 Medium Possible
2.2x10-5 to 1.1x10-4 Light Marginal
<2.2x10-5 Negligible Not Needed
Notes: q = estimated volumetric flow rate, hL = head loss across the structure, W = width
1 cfs/ft of head/ft. of levee = 44,883 gpm/ft of head/100 ft. of levee
If one or more of the problems described above must be mitigated, the mitigation
typically employs one of three general strategies:
a. Seepage barriers
This strategy consists of constructing an element that either directly blocks the
seepage pathway or lengthens the seepage pathway. Blocking seepage can
result in increased hydraulic pressures and hydraulic gradients in some locations
within an embankment or foundation. When methods to block seepage are used,
the engineer must assess whether the resulting pressures and gradients will be
detrimental to structural stability and the internal erosion potential.
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b. Providing controlled drainage

In locations where high pore water pressures result in either high hydraulic
gradients or uplift forces, the pressures can be reduced by providing a controlled
drainage system. The control in such a system involves assuring that the water
can be drained without causing internal erosion or resulting in excessive water
losses. Controlled drainage typically increases the volumetric flow rate.
c. Providing Filtration
In locations where there is potential for internal erosion, the progression of
erosion can be halted by providing adequate filtration of the eroding soil particles.
The most reliable and long-lasting filters are constructed using sands and gravels
graded to specifications that provide both soil particle retention and adequate
seepage flow capacity.
Specific methods for seepage and internal erosion mitigation options are provided in the
following sections.
6-5.2 Seepage Barriers.
Seepage barriers include a range of options for (1) blocking flow through a high-
permeability layer of soil or rock or (2) extending the seepage pathway to reduce
seepage volume and hydraulic gradients.
6-5.2.1 Vertical Barriers.
Vertical barriers or seepage cutoff walls are zones with low hydraulic conductivity that
are constructed (1) through permeable dam and levee embankments, (2) through dam
and levee foundations with permeable layers, (3) surrounding excavations below the
ground water table, or (4) blocking aquifers to prevent the spread of groundwater
contamination. Vertical barriers are generally classified based on the excavation or
construction method and the backfill type.
The various construction methods, their application, characteristics, and requirements

are summarized in Table 6-9. Continuous slurry trenches are generally backfilled using
one of three material types: Soil-Bentonite (SB) backfill, Cement-Bentonite (CB) backfill,
and Soil-Cement-Bentonite (SCB) backfill. Element slurry walls are generally backfilled
with concrete or plastic concrete using the Tremie method that fills the element from the
bottom up while displacing the slurry. The characteristics of vertical barrier backfill
materials are presented in Table 6-10.
312
Table 6-9 Construction Methods for Vertical Seepage Barriers (Cutoff Walls)
Construction
Description and Applicability Characteristics and Requirements
Method/Type
Interlocking steel sheets are typically driven into the
Leakage occurs only through interlocks, making performance reliant
ground with a vibratory hammer with rapid installation
on maintaining interlock integrity. Predrilling in soil with gravel and
Steel Sheet in mixed soils with limited amounts of gravel and
cobble improves chances of interlock integrity. Seepage resistance
Piles cobbles. Interlocking of steel sheets becomes difficult
tends to increase with age due to clogging and oxidation along the
with increased driving depth, increased soil density,
interlocks. Corrosion may be a concern for structural integrity.
and increased gravel and cobbles.
Interlocking vinyl sheets are pushed into very soft Leakage occurs only through interlocks, making performance reliant
Vinyl Sheet
ground or installed in excavated trenches. Depth is on interlock integrity. Corrosion is not a concern although chemical
Piles
limited by excavation stability (see slurry walls). stability in harsh environments should be assessed.
Continuous construction avoids construction joints but is susceptible
Wall constructed in a continuous trench that is
to “windows” in the wall due to partial trench wall collapse or sand
excavated with a long-reach excavator and/or
Continuous settling from slurry. Walls are generally ductile but of low strength
clamshell. Trench is stabilized with slurry consisting of
Slurry Trench and high erodibility. Backfill compressibility can result in vertical and
either a mixture of bentonite and water or a polymeric
Wall lateral consolidation leading to distress to overlying and adjacent
slurry. Wall material can vary from SB and SCB
structures. Stability of long excavations can be a concern.
backfill to self-hardening CB slurry (see Table 6-10).
Generally limited to soil and soft rock.
Continuous wall constructed by sequentially
The length of elements is determined by considering trench and
overlapping vertical elements. Slurry-supported
embankment stability, backfill procedures, and other construction
rectangular elements are excavated using a hydraulic
considerations. Care should be taken to ensure good connection
Element clamshell in soils and soft rock and using a
with construction joints. Excavation stability is less of a concern
Slurry Wall hydrocutter in moderate to hard rock. Circular
because of the limited duration of excavation. Elements can be
elements can overlap to form a secant wall. Backfill is
excavated into most soils and rock using a variety of excavation
usually concrete or plastic concrete placed from the
equipment.
bottom up using the tremmie method.
Insertion of a geotextile membrane into an excavated
trench forms a very low hydraulic conductivity barrier.
Membrane creates a very low hydraulic conductivity continuous
Vertical Membranes are often placed in slurry trench
seepage barrier. Membranes are often used in environmental
Membrane excavations. Interlocking elements are glued or
applications where very small leakage volumes are critical.
welded to edges of membrane sheets to form
interlocks with adjacent sheets.
313
Table 6-9 Construction Methods for Vertical Seepage Barriers (Cutoff Walls)
Construction
Description and Applicability Characteristics and Requirements
Method/Type
In situ soils are mixed in-place with a slurry consisting Sets of three or more augers are drilled at one time and overlapped
primarily of bentonite, cement, and water to produce a with adjacent sets to provide continuity of the wall. This method
soil-cement material that has reduced permeability often results in layering in the wall as the augers encounter different
and increased strength. The soil mixing can be soil types with depth. Cutter soil mixing uses a continuous cutter
Deep Mixing
performed using the multi-axis mixing or vertically- (resembling a very large chain saw) to mix the entire column of soils
Method
mixing cutter soil mixing techniques. Multi-axis mixing simultaneously in a vertical column. This method results in more
uses overlapping soil augers that are drilled into the uniform wall properties with depth. Strength and permeability of the
ground as the slurry is pumped through the tips of the “soil-cement” can be adjusted by the components and dosing of the
augers. slurry. Deep mixing is limited to soil and very soft rock.
Jet grout columns are constructed with a probe that
uses high-pressure jets to simultaneously erode soil
and fill the column with a grout mixture. Soil-cement
is formed as varying amounts of eroded soil are mixed
Because the column is formed by jets, walls can be constructed that
with water and grout. The procedure can be
seal against irregular rock or concrete surfaces that are otherwise
performed with single-, double-, and triple-fluid
difficult with rigid excavation. Equipment is adaptable for
methods. The single-fluid method injects the cement
Jet Grout construction in limited-space and low-overhead conditions. Due to
grout out of a single nozzle that simultaneously erodes
Walls very high pressures, the risk of hydrofracturing embankments is
the soil and provides the grout. The double-fluid
high, limiting the applicability in dams and levees. The jet grout
method uses a double nozzle that shoots a stream of
method is typically more expensive than other methods, thus limiting
grout through a shroud of air, increasing the range of
its use to limited access and special needs projects.
the grout jet so that a larger diameter column can be
produced. The triple-fluid method uses a jet of water
shrouded in air to cut the column followed by a jet of
grout to fill the column.
314
Table 6-10 Backfill Material Description and Characteristics for Vertical Seepage Barriers (Cutoff Walls)
Backfill Type Description and Applicability Characteristics and Requirements
SB backfill is a low-strength, high-ductility, and low-hydraulic
SB backfill consists of a mixture of the excavated soils
conductivity material. While the material is highly
and bentonite and is used primarily for continuous slurry
compressible, friction on the sidewalls may reduce vertical
Soil-Bentonite trenches. The SB mixture is often created by adding the
stress on the backfill, causing it to remain in an
(SB) Backfill trench slurry to the trench spoils on the ground adjacent
underconsolidated state in the trench. The SB may be
to the trench. The mixing is often done using a bulldozer
susceptible to vertical and horizontal consolidation and
or a soil-mixing machine.
hydraulic fracture.
CB backfill is a self-hardening slurry that is used to
Cement- support continuous slurry trenches and then hardens into CB is less prone to consolidation than SB. The strength of
Bentonite (CB) a consistency similar to stiff clay. The slurry is mixed in a CB can be adjusted but generally has the consistency of very
Backfill batch plant and may contain additives to increase stiff to hard clay.
strength or retard setup.
SCB is less prone to consolidation than SB. The strength of
SCB backfill is often blended and placed similar to SB SCB can be adjusted. It should be noted that some SCB
Soil-Cement- backfill but with the addition of cement to give strength to walls have been found to have discontinuities or windows
Bentonite the backfill for a more robust element within the near the bottom of the wall. These windows are thought to
(SCB) Backfill embankment. Primarily used as backfill for continuous be the result of premature setting of the backfill that breaks
slurry trenches. into blocks due to the slumping that occurs during normal
placement.
Soil cement is a term often used to describe the product The strength and permeability of soil cement will vary
of deep soil mixing. In the wet method, a slurry depending on the amount of cement and bentonite in the
consisting primarily of bentonite, cement, and water is slurry and the type of soil it is mixed with. Gravelly or clean
Soil Cement
blended with the in situ soil (see deep mix method in coarse-grained soils will tend to have higher strength and
Table 6-9). The less common dry method injects cement higher permeability than sands and silts that are mixed with
and bentonite powders directly into the soil for mixing. the same amount of an identical slurry.
Plastic concrete is conventional concrete with bentonite The bentonite results in reduced strength and erosion
Plastic added to increase ductility with the intent of making the resistance of the concrete. Thus, a balance of robustness
Concrete wall more compliant with the surrounding soils, and compliance should be considered when choosing
decreasing cracking of the wall and surrounding soils. between conventional and plastic concrete.
Conventional concrete adds a robust element to an The rigidity of the wall may cause stress concentrations in
Conventional
excavation or dam/levee embankment that is high- other elements of the embankment, resulting in cracking of
Concrete
strength, stiff, and highly erosion resistant. the wall or surrounding soil.
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6-5.2.2 Required Penetration for Cutoff Walls in Supported Excavations.
To prevent instability of the base of supported excavations due to heave (i.e., quick
condition or blowout), the vertical cutoffs must extend deep enough to reduce hydraulic
gradients or uplift pressures to acceptable levels. The first option to consider for
embedment is to extend the cutoff to a low-permeability layer having considerable
thickness. However, this is not always feasible.
In uniform pervious sands, critical hydraulic gradients may develop at the base of the
excavation. The required wall penetration to provide a factor of safety against effective
stress heave (quick condition) and piping in homogenous, isotropic sands can be
calculated using Figure 6-11. For homogenous, anisotropic sands, the penetration
depth can be reduced by the transformation factor, a (see Section 6-2.4.1 Equation 6-
8). For clean sand, exit gradients greater than about 0.5 to 0.75 will cause unstable
conditions for men and equipment operating on the subgrade. To avoid instability,
provide sheeting penetration for a safety factor of 1.5 to 2 against effective stress heave
as calculated in Figure 6-8.
In layered sands, variation in permeability results in a change of seepage conditions

from that assumed in Figure 6-11. Figure 6-12 presents guidance for situations with
layered sands. In layered soils with layers of very fine sand, silty or clayey sand, or silt
and clay, the risk of bottom heave (total stress heave) must be considered. Figure 6-13
presents guidance for avoiding bottom heave in excavations. Alternatively, the
conditions can be assessed using flow nets or finite element analyses, and Figure 6-11
through Figure 6-13 can be used to confirm the results.
The relationships presented in Figure 6-11 through Figure 6-13 were developed based
on the results of laboratory modeling performed by Marsland (1953). The results were
reported for loose and dense sands. Loose sands were placed “with a water jet” while
the dense sands were placed “with an electrically vibrated” hammer. No relative
densities were reported although porosities of 42 percent and 37 to 38 percent were
reported for the loose and dense sands, respectively.
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Figure 6-11 Required Depth of Penetration of Cutoff Wall-Supported Excavations

in Homogenous Isotropic Sand (after Marsland 1953)
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Supported Excavations for Stratified Sand (after Marsland 1953)
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Supported Excavations in Sand Containing Fine-Grained Layers
(after Marsland 1953)
6-5.2.3 Seepage Blankets and Berms.
Upstream seepage blankets are constructed upstream of a dam or levee to increase the
seepage pathway as depicted in Figure 6-14a. Upstream blankets are constructed of
compacted fill with low hydraulic conductivity or a geomembrane protected with a
coarse-grained soil cover. The increased seepage pathway will tend to decrease the
amount of seepage, decrease uplift pressures below the dam or levee, and decrease
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exit gradients. The primary design concern of seepage blankets is providing a uniform
foundation for the blanket. Differential settlement of the blanket under reservoir loading
can result in tearing or cracking of the blanket and concentrated leakage, decreasing
the effectiveness of the blanket.
Figure 6-14 Seepage Blankets and Berms
Downstream seepage berms are designed to prevent excessive hydraulic pressures on

the downstream side of a small dam or levee. Often these pressures are the result of a
low hydraulic conductivity soil layer (blanket layer) that blocks the exit of seepage water.
Seepage berms are usually constructed using one of three designs: (1) an impermeable
berm, (2) a permeable berm, or (3) a composite berm.
The concept of an impermeable berm used to resist uplift pressures on a low-

permeability blanket layer is presented in Figure 6-14b. The berm is constructed of
generally low-permeability soils and is designed to increase the weight of the blanket
layer thereby resisting total-stress heave. For blanket layers that allow some seepage,
the impermeable berm will increase the resistance to seepage flow through the blanket
layer, resulting in higher pressures below the levee and downstream toe. This also
requires the berm to extend long distances from the levee in order to produce low uplift
pressures beyond the berm.
A permeable seepage berm is illustrated in Figure 6-14c. The permeable berm is

constructed of permeable materials (generally sand) that increase the total weight on
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the blanket without significantly increasing seepage resistance. Where the blanket layer
is sufficiently permeable allow seepage on the landside, permeable berms can be much
shorter than impermeable berms. A disadvantage of this berm type is that appropriate
high-permeability soils can be very expensive in some locales.
A composite seepage berm consists of an upper layer of undifferentiated fill underlain

by a drain and filter layer (see Figure 6-14d). Thus, the berm acts similar to the
permeable berm but is less expensive because less of the filter material is required.
The drainage layer must be designed with adequate flow capacity to prevent the buildup
of water pressure in the layer. In some cases, pipes have been included in the drainage
layer to increase flow capacity. However, such pipes increase the amount of
maintenance needed for the berm.
6-5.2.4 Other Types of Seepage Barriers.
6-5.2.4.1 Cutoff Trench.
Cutoff trenches are constructed in the foundations of dams and levees as part of the
original construction. The cutoffs are generally a broad trench with sloped sidewalls
that is backfilled with low hydraulic conductivity material. The trench extends partially or
fully through permeable soil and rock layers in the foundation. In addition to the
seepage control offered, cutoff trenches provide an opportunity to visually inspect the
subsurface conditions beneath the dam or levee. Care should be taken to ensure that
the downstream side of the berm is properly filtered (see Section 6-5.3) to prevent
internal erosion into permeable foundation soil or rock.
6-5.2.4.2 Foundation Grouting.
Foundation grouting is performed by pumping stabilized cement grout into boreholes to

fill joints and voids in bedrock and reduce the effective hydraulic conductivity of the
formation. Foundation grouting can be performed prior to dam construction or after
construction as a remedial measure. Grout is pumped into discrete depth intervals of
the boring by sealing off an interval using inflatable bladders called packers. Grout lines
consisting of evenly spaced boreholes are typically aligned parallel to the axis of the
dam although other configurations may be applicable for special circumstances. In
many cases, multiple grout lines (grout curtains) are installed along the dam axis with
the boreholes inclined in opposite directions to increase the chances of encountering all
of the joints and voids in the rock.
6-5.2.4.3 Liners.
Leakage from ponds and reservoirs can be reduced by installing a liner in the base of
the pond or reservoir. Such liners can consist of compacted clay, synthetic
geomembrane liners, or native soils supplemented with bentonite or other materials to
reduce hydraulic conductivity. Similar to upstream blankets, the foundation for the liner
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must be sufficiently uniform to prevent differential settlements under the reservoir
loading, which could tear or crack the liner.
6-5.3 Filters and Drains.
Drainage systems provide relief of hydraulic pressures in foundations, intercept paths of

concentrated seepage, and control the release of the drained water to prevent internal
erosion. Properly designed drainage elements will have adequate flow capacity to
reduce the pressures for which they are designed while being filter compatible with the
surrounding soils to prevent internal erosion through the drainage elements.
Filters and drains are essential components for the drainage systems of a dam, levee,
excavation or other geotechnical construction. A filter prevents the migration of the
base soil into another soil layer or zone, into a drain, or out of the system through
surface flow. The base soil is the material from which the seepage flow is exiting. A
drain removes collected water from the collection point to a suitable discharge location.
Drains can consist solely of coarse-grained soil that allows rapid seepage through its
interstitial voids or can include drainage pipes that collect water from the surrounding
coarse-grained drain material.
6-5.3.1 Mineral Filter Criteria.
In order to satisfy its purpose, the filter must (1) have a gradation fine enough to prevent
the migration of the base soil into the filter, (2) have high enough hydraulic conductivity
to not restrict flow from the base soil, and (3) have the ability to collapse so that cracks
in the base soil are not propagated through the filter. Mineral filters consist of sand and
gravel that are specifically graded to meet these criteria. Design procedures for mineral
filters are have been developed by most U.S. agencies involved in dam design. These
criteria have been summarized by FEMA (2011). A brief summary of the design
procedure is provided in this section.
In the design procedure, the soil being filtered is the base soil and characteristics of the
base soil gradation will be followed by a B (e.g. the grain size of which 85 percent of
the base soil is passing will be denoted: D85 B ). Similarly, the filter gradation will be
followed by an F (e.g. D15 F ). Examples of the grain sizes associated with filter design
are provided in Figure 6-15.
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Figure 6-15 Example Base Soil and Filter Gradations
Based on FEMA (2011), filter design is completed using the following steps:
Step 1. Plot the grain-size distributions for the base soil and assess whether the base
soils contain dispersive soils (e.g., soils that disaggregate when in contact with water).
ASTM D6572 and D4647 can be used to identify dispersive soils.
Step 2. Assess if the base soil has particles larger than the No. 4 sieve or is gap
graded. Gap graded soil may be susceptible to internal instability. Special
consideration should be given to a gap graded soil to insure the fine portion of the base
soil is protected from erosion.
Step 3. If the base soil contains particles larger than the No. 4 sieve, regrade the grain-
size distribution to include only the portion of the distribution finer than the No. 4 sieve.
FEMA (2011) provides details on the regrading procedure.
Step 4. Determine the base soil category using Table 6-11.
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Table 6-11 Base Soil Categories for Mineral Filter Design (after FEMA 2011)
Percent Finer Than No. 200 Sieve (0.075-
Base Soil
mm) Base Soil Description
Category
(after re-grading where applicable)
1 > 85 Fine silt and clays
2 40 –85 Sands, silts, clays, and silty sands
3 15 –39 Silty and clayey sands and gravels
4 < 15 Sands and gravels
Step 5. Determine the filter criteria for base soil retention by calculating the maximum
D15 F based on the criteria in Table 6-12. The retention criteria are based on the
principle that the larger particles in the base soil (represented by D85 B ) must be
retained by the voids in the filter (controlled by D15 F ). By providing a maximum D15 F ,
the criteria ensure that the filter voids are sufficiently small. If the base soil has a range
of grain-size distributions, the smallest value of D85 B should be used for the retention
criteria.
Table 6-12 Restraint Criteria for Mineral Filter Design (after FEMA 2011)
Base Soil Filtering Criteria–Maximum D15 F

Category Non-Dispersive Soil Dispersive Soil
Max D15 F ≤ 9 ⋅ D85 B Max D15 F ≤ 6.5 ⋅ D85 B
1
Max D15 F ≥ 0.2 mm Max D15 F ≥ 0.2 mm
2 Max D15 F ≤ 0.7 mm Max D15 F ≤ 0.5 mm
 40− A   40− A 
Max D15 F ≤  [ B −C ] + C Max D15 F ≤  [ B −C ] + C
 25   25 
3 where: where:
A = % passing No. 200 sieve, A = % passing No. 200 sieve,
B = 4 × D85 B ≥ 0.7 mm, and B = 4 × D85 B ≥ 0.5 mm, and
C = 0.7 mm C = 0.5 mm
4 Max D15 F ≤ 4 ⋅ D85 B Max D15 F ≤ 4 ⋅ D85 B
Note: D85 B and percent passing No. 200 sieve are determined after regrading.
Step 6. Determine the flow criteria for the filter by calculating the minimum D15 F based
on the criteria in Table 6-13. The permeability criteria are based on the principle that
the filter’s hydraulic conductivity is strongly related to D15 F . By providing a minimum
D15 F , the filter will have sufficient flow capacity in comparison to the base soil. If the
base soil has a range of grain-size distributions, the largest value of D15 B should be
used for the permeability criteria. Plot the minimum and maximum D15 F on a gradation
plot and adjust the minimum D15 F to ensure that the ratio between the minimum and
maximum D15 F is not greater than 5 (i.e., Min D15 F Max D15 F ≤ 5 ).
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Table 6-13 Flow Criteria for Mineral Filter Design (after FEMA 2011)
US Agency Filter Permeability Criteria
USBR Min D15 F ≥ 5 ⋅ D15 B
USACE Min D15 F ≥ ( 3 to 5) ⋅ D15 B

NRCS Min D15 F ≥ ( 4 to 5) ⋅ D15 B
All Min D15 F ≥ 0.1 mm
Note: D15 B is determined prior to regrading.
Step 7. Filters should also have a coefficient of uniformity ( Cu = D60 F D10 F ) is between
2 and 6.
Step 8. Limit the amount of oversized material and fines in the filter. Most agencies
require that the maximum particle size of the filter ( D100 F ) be less than 2 inches (51
mm) although the USACE allows particles up to 3 inches (76 mm). The minimum
particle size associated with 5% passing for the filter ( D5 F ) is the No. 200 sieve or
0.075 mm. Any fines present in the filter soil should be non-plastic (i.e., PI = 0).
Step 9. Limit segregation potential of the filter by determining the maximum D90 F from
Table 6-14. Segregation occurs more easily if a wide range of particle sizes is present.
Table 6-14 Segregation Criteria for Mineral Filter Design (after FEMA 2011)
Minimum D10 F (mm) Maximum D90 F (mm)

< 0.5 20
0.5 – 1.0 25
1.0 – 2.0 30
2.0 – 5.0 40
5.0 – 10 50
10 – 50 60
Step 10. Plot the criteria for minimum D5 F , minimum and maximum D15 F , maximum
D90 F , and maximum D100 F on a gradation plot to create an acceptable gradation band
for the filter. Compare candidate filters with the criteria and gradation band.
Design for Drainpipe Perforations. If the filter is to contain a drainage pipe, the filter soil
around the pipe is referred to as the envelope material. The smallest value of D50 E ( E
stands for envelope) allowed by the gradation should be larger than the size of the
maximum pipe perforation.
An example of a filter design for a Type 4 base soil is presented in Figure 6-16.
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Figure 6-16 Example Filter Design
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6-5.3.2 Geotextile Filter Criteria.
In some cases, geotextiles can be used for soil filtration and drainage in lieu of mineral
filters and gravel drains. In general, the FEMA (2011) guidelines do not recommend the
use of geotextiles in critical areas of dams. The term geotextiles refers to a wide variety
of synthetic, fabric-like products that vary depending on the types of fibers used in the
manufacturing and how these fibers are interconnected. The most common forms
include woven, nonwoven, and knitted fabrics. Woven geotextiles are manufactured by
weaving two perpendicular sets of synthetic fibers to form a fabric. Nonwoven
geotextiles have a random orientation often resembling a felt fabric and are
manufactured by either needle punching, spun bonding, or resin bonding. Knitted
geotextiles consist of interlocking series of loops of fiber yarns that form a fabric.
A common drainage and filtration application for geotextile filter fabrics is a subsurface
drain such as presented in Figure 6-17. In this application the drainage is provided by
the poorly-graded gravel drainage rock and the slotted outlet pipe. The role of the
geotextile is to prevent migration of the surrounding soils into the drainage rock and
pipe. In this way, the geotextile prevents erosion of the surrounding soil and/or clogging
of the drain. Other applications include drains within retaining wall backfill and drainage
blankets below embankments.
Soil Cover
Geotextile
Fabric
Excavated
Trench Drain Rock
Slotted Drain
Pipe
Figure 6-17 Subsurface Drain Constructed of Filter Fabric, Drainage Rock, and a
Slotted Pipe
For the geotextile fabric in Figure 6-17 or other applications to perform adequately, the
geotextile should be designed to have the ability to: (1) retain particles of the base soil
to prevent migration through the fabric, (2) allow the passage of water without significant
buildup of water pressure, and (3) resist clogging due to accumulation of fine soil
particles, chemical precipitates, and biological precipitates.
Criteria for retention of soil particles was proposed by Giroud (2010) based on D20 B and
the soil’s linear coefficient of uniformity ( C 'u ). As shown in Figure 6-18, a log-linear
approximation is used to represent the base soil’s gradation curve. The linear particle
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size ( D 'x ) is the obtained the log-linear approximation. The values of D '100 and D '0 are
defined by the ends of a straight line drawn through the middle part of the gradation plot
(i.e., through D10 and D60 ).
Figure 6-18 Linear Coefficient of Uniformity (after Giroud 2010)
Geotextiles are designed for retention based on the geotextile opening size ( O95 ), which
indicates the size at which 95 percent of openings are smaller. According to Luettich
(1992), geotextile retention criteria for soils with less than 10% fines can be selected
using the base soil gradation and Table 6-15.
For soils with more than 10% fines and PI greater than 5, the geotextile should have
O95 less than 0.21 mm, provided the soil is non-dispersive. For dispersive soils with
more than 20% clay sized particles, a filter of 75 to 100 mm of sand should be placed
between the base soil and the geotextile, and the geotextile should be designed to
retain the filter sand. Retention of non-plastic soils with more than 10% fines can be
designed using Table 6-15.
Geotextile filters also must have a much higher hydraulic conductivity than the base soil.
A filter cake of trapped particles often forms on the face of a geotextile. These particles
must not reduce the hydraulic conductivity of the geotextile to the point that it restricts
seepage flow out of the base soil. However, clogging is difficult to quantify. Some
clogging will occur when the geotextile is put into service. The geotextile must remain
sufficiently open so that accumulation of particles and chemical and biological
precipitates will not reduce the hydraulic conductivity to the point where the filter
cake/geotextile system becomes less permeable than the base soil. The USBR (2014)
recommends the following considerations be taken to assess clogging potential:
• Use the largest opening size that satisfies the retention criterion.
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• Do not use geotextile filters in environments where precipitates are likely to form.
Avoid high alkalinity groundwater, which can form calcium, sodium, or
magnesium precipitates. Also avoid acidic seepage, which can form iron and
aluminum hydroxide precipitates.
• Avoid use of geotextiles with internally unstable ( Cu > 20) or dispersive soils.
• Avoid organic-rich environments such as agricultural runoff, landfill leachates,
and sites known to form iron bacteria.
• Make sure that the geotextile filter makes intimate contact with the soil.
• Do not place geotextile filters against cohesive soils containing voids.
Table 6-15 Geotextile Opening Size Criteria for Soils with Less than 10% Fines
(after Luettich et al. 1992)
Base Soil Description
Primary Relative Density Geotextile Retention
Secondary
Category Gradation Description Criteria
Characteristic ( Dr )
Description
Loose 9 '
O95 < D50 B
Obtain C 'u Dr < 35% C 'u
from straight Widely
Stable Soil Medium dense 13.5 '
Graded O95 < D50 B
(1 < Cc < 3) line drawn 35% < Dr < 65%
(C 'u > 3) C 'u
through D60
Less than Dense 18 '
10% fines and D30 O95 < D50 B
and less Dr > 65% C 'u
than 90% Loose
gravel '
Obtain C 'u O95 < C 'u D50 B
Dr < 35%
from straight Uniformly
Unstable Soil Medium dense
Graded O95 < 1.5 ⋅ C 'u D50 B
'
(Cc < 1, Cc > 3) line drawn 35% < Dr < 65%
(C 'u < 3)
through D30 Dense '
and D10 O95 < 2 ⋅ C 'u D50 B
Dr > 65%
The permeability requirement for a geotextile can be stated as:
kg ψ g ⋅ tg
FS=
g = (6-16)
ks ks
where:
FS g = factor of safety for geotextile permeability,
k g = hydraulic conductivity of the geotextile across the plane of the fabric,
k s = hydraulic conductivity of the base soil,
ψ g = permittivity of the geotextile, provided by manufacturers or from testing (ASTM
D4491), and
t g = geotextile thickness.
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Giroud (2010) suggests that an FS g of 10 to 20 is appropriate to maintain adequate filter
permeability. Others (e.g., Loudiere et al. 1983, Christopher and Fischer 1991)
recommend that the FS g value be between 10 and 100.
Over the past several decades geotextile filter fabrics have been developed to provide
filtration between base soil and coarse-grained drain materials. While these fabrics are
inexpensive compared to mineral filters, they are considered by many to be far less
reliable than mineral filters due to their propensity to clog and the potential to be
damaged during construction. For this reason, all major dam owning and regulating
agencies in the U.S. (U.S. Army Corps of Engineers, U.S. Bureau of Reclamation,
FEMA, and FERC) do not allow the use of filter fabrics within critical areas of dam
embankments or in high-hazard structures.
6-5.3.3 Surface and Subsurface Drainage.
Pavements and other surface treatments can be destabilized by water ponded at the
ground surface or shallow phreatic surfaces. Near-surface groundwater may be
collected by intercepting drains prior or collected in the pavement base material as it
exits the subgrade. Accumulation of surface water in wide flat areas, due to rainfall or
other surface sources, can be mitigated using trench drains connected to deeper
drainage systems.
6-5.3.3.1 Intercepting Drains.
Intercepting drains (a.k.a., stability trenches or stability drains) can be used in locations
where water seeping from a hillside has a detrimental effect on slope stability or the
performance of roadways. These drains consist of shallow trenches with collector pipes
surrounded by drainage material, placed to intercept seepage moving horizontally in an
upper pervious stratum as illustrated in Figure 6-19. The trench backfill can consist of a
filter material compatible with the surrounding soil as shown in Figure 6-20 or a
drainage aggregate wrapped in filter fabric (Figure 6-17). The type of trench backfill and
filtering mechanism used will depend on the criticality of the drain. Drain designs should
aim to be as simple and constructible as possible while still providing suitable filtering.
The effect of intercepting drains on seepage patterns can be evaluated using flow nets
or modeled by finite element analyses. Such analyses should also assess the
volumetric flow rate into the drains so that the drain pipes and outlets can be properly
designed.
6-5.3.3.2 Surface Blanket Drains.
Surface blanket drains are used to intercept ground water beneath pavements and
structures to mitigate the buildup of water pressure or destabilization of subgrades. In
many cases, the aggregate base layer of pavements may be assessed for its
effectiveness as a blanket drain. Design of surface blanket drains should consist of: (1)
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assessment of the flow rate of drainage into the blanket to calculate the needed spacing
for outlets and (2) assessment of the tolerable uplift pressure in the blanket to prevent
damage to the overlying pavement, embankment, or structure.
Figure 6-19 Use of Subsurface Interceptor Drains and Blanket Drains for
Roadway Drainage
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Figure 6-20 Subsurface Drain with a Two-Stage Filter and a Slotted Pipe
The thickness of aggregate base course needed to provide effective drainage can be
calculated using Figure 6-21. For a range of slopes, the degree of drainage is related to
the porosity and hydraulic conductivity of the aggregate base, the thickness of the base
layer, the subgrade slope, the drain spacing, and the time allowed for drainage. Figure
6-21 can be used to select drain spacing or evaluate suitability of base material.
Effective porosity (also called specific yield) is the quantity of water per unit volume that
is not retained in the soil by capillarity during discharge (Barber 1959). It ranges from
25 percent for a uniform material, such as medium to coarse sand, to 15 percent for a
well-graded sand-gravel mixture.
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Figure 6-21 Drainage of an Aggregate Base Course (after Barber 1959)
6-5.3.3.3 Drainage of Ponded Areas.
During times of heavy rainfall or runoff from adjacent areas, vertical drainage trenches
can be used to mitigate ponding water in level areas or enclosed basins. The flow rate
of seepage into parallel trenches can be estimated using Figure 6-23 for an underlying
zone that is either completely pervious or completely impervious in comparison to the
surface layer. The trench spacing ( 2 × S ) can be designed such that the calculated flow
rate into the trenches meets or exceeds the required surface infiltration for an area of
1× 2 × S . The trench must also be designed to sustain the required flow rate and may
include collector pipes. If sufficient drainage capacity cannot be provided using
trenches, surface drainage facilities are required to prevent ponding.
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6-5.3.1 Retaining Wall Drainage.
Water imposes additional destabilizing forces on retaining and basement walls that can
lead structural distress or collapse. Drainage systems should be designed to prevent
the buildup of water behind walls. Figure 6-22 presents several alternatives for
providing retaining wall drainage.
Figure 6-22 Retaining Wall Drainage Alternatives
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Figure 6-23 Seepage into Drainage Trenches Used for Draining Ponded Areas
(after Kirkham 1950 and 1960)
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6-5.3.2 Filters and Drains for Embankments.
Drains and filters are constructed within dams, levees, and other embankments or
slopes to drain excess pressures and intercept pathways for internal erosion. The
following paragraphs discuss the different types of embankment drains and their
applications. In general, embankment drains act to lower the phreatic surface and
piezometric pressures in an embankment or slope, resulting in increased slope stability
and decreased internal erosion potential. All drain systems should be designed to filter
the base soil and provide adequate drainage capacity for the designed purpose.
Embankment subdrains are located below embankment fills to intercept seepage from
native soil and bedrock and reduce seepage into the embankment as illustrated in
Figure 6-24. Subdrains consist of filter soil or drainage aggregate wrapped in filter
fabric. The drains often contain a slotted or perforated collection or outlet pipe.
The other major types of embankment drains are used for intercepting and controlling
seepage through or below water retaining structures.
Toe drains consist of a filter and drain located at or near the toe of a dam to collect
exiting seepage through the embankment (Figure 6-25a). Toe drains are common in
small homogenous dams and levee embankments as well as below the toe of small
zoned earth dams. Toe drains do not effectively intercept cracks or defects in dam or
levee embankments or shells.
Figure 6-24 Embankment Fill Subdrains
Blanket drains are located at the downstream base of a dam or levee and are designed
to collect seepage exiting the embankment (Figure 6-25b). Blanket drains are common
in small to moderate sized dam and levee embankments. They are not an effective
means of intercepting cracks or defects in dam or levee embankments or shells.
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Chimney drains are usually located near the centerline of a dam or levee and are
designed collect seepage through the embankment (Figure 6-25c). Chimney drains are
a standard component of all modern, large embankment dams. Chimney drains are
designed to intercept cracks, defects, or permeable layers in dam or levee
embankments or cores. They typically are connected to a blanket drain below the
downstream or landside portion of the embankment.
Figure 6-25 Toe, Blanket, and Chimney Drains
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An outlet filter collar is a zone of filter material located along an outlet or other
penetration through an embankment as shown in Figure 6-26. The outlet filter collar
prevents concentrated leak erosion along the penetration. These filters are generally
not connected to a drain.
Figure 6-26 Outlet Filter Collars
6-5.3.3 Foundation Drainage.
Under certain conditions, excessive water pressure is present in the soil or rock below
or adjacent to a dam or levee. High water pressures can lead to risk of heave, internal
erosion, or instability of slopes or structures. These conditions often occur when a low
hydraulic conductivity layer prevents drainage from an underlying layer with higher k.
Drainage methods for mitigating such pressures include relief trenches and wells.
6-5.3.3.1 Relief Trenches.
A relief trench penetrates the upper layer and allows high pressures in the deeper
pervious layer to dissipate by upward seepage as shown in Figure 6-27. The relief
trench is filled with an appropriate filter material. The width of the trench should be
sufficient to reduce the upward pressures beneath the low hydraulic conductivity layer
while limiting upward hydraulic gradients in the relief trench to 0.4. Commonly, relief
trench design is performed using two-dimensional finite element analysis.
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Figure 6-27 Relief Trench Used to Relieve Pressure from Beneath a Blanket
Layer with Low Hydraulic Conductivity (not to scale)
Relief trenches reduce the drainage path and increase the flow rate below a hydraulic
structure. Seepage at the ground surface must be managed when relief trenches are
used. A typical seepage water management system includes collection trenches
leading to a sump where the water can be pumped away.
6-5.3.3.2 Relief Wells.
A relief well is a large diameter well that extends into a layer of high hydraulic
conductivity. Rows of relief wells are installed in areas of excess foundation pressure,
often along the downstream side of a levee or dam. Similar to trenches, relief wells
reduce water pressure by providing a vertical seepage pathway and reducing the length
of the flow path as indicated in Figure 6-28.
Relief wells are constructed in large diameter shafts, often 0.5 to 1.0 meters in diameter,
cased with a 15 to 30 cm diameter casing. Details of a typical relief well installation are
presented in Figure 6-29. The portion of the relief well that extends into the pervious
foundation soil is cased with slotted or screened casing and packed with a filter material
designed to filter the surrounding base soil. Most modern casings consist of PVC solid
pipes with either slotted PVC or stainless steel screens. However, some wood-stave
screen wells are still in operation. The upper portion of the well consists of a solid
casing with an annular space sealed with concrete and bentonite to prevent flow and
erosion of the upper blanket material.
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Figure 6-28 Relief Well Used to Relieve Pressure from Beneath a Blanket Layer
with Low Hydraulic Conductivity (not to scale)
Figure 6-29 Typical Relief Well Construction
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The design of relief wells is a three-dimensional problem and can be visualized in terms
of the lowering of the piezometric surface that occurs at the wells. The piezometric
surface is lowered the most at the well locations as shown in Figure 6-30. Along the
line of the wells, the piezometric surface is typically highest at the midpoint between any
two wells. Figure 6-30 can be used to estimate the relief well discharge in terms of the
well spacing, the well radius, geometric variables, and an extra length factor that
empirically accounts for well resistance. The appropriate extra length factor can be
determined for wells that penetrate 25, 50, and 100 percent of the pervious foundation
layer thickness using the lower diagrams. Alternatively, relief wells can be designed
using three-dimensional finite element analyses.
Relief wells require a periodic program of well cleaning and maintenance to prevent
clogging of the well screen and filter. Relief well flows should be documented
throughout their life span and flows compared with river or reservoir levels to monitor
any changes that may be occurring in well efficiency.
6-6 DEWATERING.
Sometimes it is necessary to lower the groundwater level to allow for subsurface

construction or to prevent flooding of existing facilities. Dewatering options include: (1)
a collection system at the base of the excavation leading to a sump, (2) a well point
system, and (3) deep extraction wells. The selection of a dewatering system depends
on the expected inflows into the excavation, the potential for heave (see Section 6-4.1),
and the required depth of groundwater lowering. There are engineering firms and
consultants that specialize in dewatering.
6-6.1 Collection and Sump.
When the volume of inflow into an excavation is low (i.e., excavation into low hydraulic
conductivity soils) and the potential for heave is not a concern, a collection and sump
system may be sufficient. The collection system may consist of a series of trenches on
the bottom of the excavation or a system of relief trenches (see Section 6-5.3.3.1) at the
base of the excavation. In some temporary cases, a layer of poorly-graded gravel can
be placed over the bottom of the excavation that allows water to flow toward the sump.
The collection system leads to a sump with a pump to remove the inflow. Sump pumps
can be either electric or gasoline-powered (i.e., “trash” pumps).
6-6.2 Wellpoint Systems.
Wellpoints are 1-1/2- or 2-inch diameter pipes that are connected to a suction system to
remove the groundwater. The lower section of each wellpoint is slotted or perforated
and screened to prevent removal of the surrounding soil into the pipe. The pipes are
pushed or jetted in place, or installed in predrilled holes. The wellpoints are each
connected to a header leading to suction pumps.
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Figure 6-30 Calculation of Relief Well Discharge and Spacing (after USACE 1952)
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Wellpoints are generally effective when the soil has D10 greater than 0.05 mm or if the
soil has a structure, such as varves or laminations, for conducting groundwater
horizontally. The maximum differential pressure that can be developed in wellpoints is
limited by atmospheric pressure. Thus, the drawdown of the groundwater level using
wellpoints is ordinarily limited to 15 to 18 feet below the center of the suction header. If
a greater amount of drawdown is required, wellpoints can be installed in successive
tiers or stages as excavation proceeds as illustrated in Figure 6-31.
Figure 6-31 Staged Installation of Wellpoints to Lower the Groundwater Table for
a Deep Excavation
The discharge capacity for each wellpoint is generally 15 to 30 gpm (3 to 6 m3/hour).

Wellpoint spacing is typically between 3 and 10 feet (1 and 3 m). Closer spacing should
be used for soils with lower hydraulic conductivity. In such soils, the effectiveness of
wellpoints can be increased by predrilling the locations and backfilling with sand around
the wellpoint.
Due to the close spacing of wellpoints, a two-dimensional analysis is generally

sufficient. The drawdown and flow into the line of wellpoints can be analyzed either with
a flow net or two-dimensional finite element analysis. For soil with high k (clean fine
sand or coarser), the quantity of water to be removed controls wellpoint layout. For silty
soils, the flow rate is relatively small, and the number and spacing of wellpoints will be
influenced by the time available to accomplish dewatering.
6-6.3 Extraction Wells.
Extraction wells consist of a bored hole containing a well casing with a screened section
in the aquifer, a filter pack, and a pipe column. A turbine-type pump with a motor at the
surface can be used, or a submersible pump may be placed within the well casing.
Extraction wells are used if (a) the dewatering system must be kept outside the
excavation area, (b) large quantities of water must be pumped for a long period of time,
(c) pumping must commence before excavation to obtain the necessary time for
drawdown, or (d) pressures must be lowered in a confined aquifer that is below a low-
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permeability layer underlying an excavation. Extraction wells may be used for soils with
classifications ranging from gravel to silty fine sand, and for water bearing rocks.
Bored shallow wells with suction pumps can be used to replace wellpoints where
pumping is required for several months or in silty soils where correct filtering is critical.
Ejector or eductor pumps may be utilized within wellpoints for lifts up to about 60 feet.
The ejector pump has a nozzle arrangement at the bottom of two small diameter riser
pipes which remove water by the Venturi principle. They are used in lieu of a multistage
wellpoint system and if the large pumping capacity provided by extraction wells is not
required. Their primary application is for sands, but with proper control they can also be
used in silty sands and sandy silts.
Figure 6-32 presents equations for analysis of drawdown and pumping quantities for
single wells or a group of wells in a circular pattern. The radius of influence ( R ) is often
defined as the radius beyond which the well has no influence. R is a function of the
discharge ( q ) and thus changes depending on the rate of pumping. The equations
presented allow the calculation of R from data in a single observation well. Once R is
known, drawdown at other locations can be calculated.
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Figure 6-32 Drawdown and Pumping Quantities for Single Extraction Wells and
Groups of Extraction Wells (after USACE 1952)
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Topic Reference
Cedergren, H. R. 1977. Seepage, Drainage, and Flow Nets, John Wiley and
Sons, Inc., New York.
Groundwater Flow
Freeze, R. A. and Cherry, J.A.1979. Groundwater Prentice-Hall, Englewood
Cliffs, NJ.
Federal Emergency Management Agency (FEMA), 2011. Filters for
Filter Design
Embankment Dams, Best Practices for Design and Construction.
Bradley, N. and VandenBerge, D. R. 2015. Beginner’s Guide for Geotechnical
Finite Element Analyses, Center for Geotechnical Practice and Research,
Numerical Seepage Analysis Virginia Tech.
Potts, D. M. and Zdravkovic, L. 2001. Finite Element Analysis in Geotechnical
Engineering: Theory and Application ICE Publishing.
6-8 NOTATION.
Symbol Description
a Isotropic transformation factor for flow nets
A Cross sectional area of the flow region perpendicular to the flow direction
C 'u Linear coefficient of uniformity (geotextile design)
Dli Particle size of the coarser sieve (Kozeny-Carman equation)
Dsi Particle size of the finer sieve (cm)
Effective grain size - α is the percent of soil particles smaller than the stated size, values of 5,
Dα
10, 15, and 20 are commonly used for α
Dx Particle size for which X% of the soil is finer
D 'x Particle size for which X% of the soil is finer for linearized particle distribution (geotextile design)
Dx B Particle size for which X% of the soil is finer for a base soil
Dx F Particle size for which X% of the soil is finer for a filter material
e Void ratio of the soil
fi Fraction of particles between two adjacent sieve sizes (Kozeny-Carman equation)
FS g Factor of safety for geotextile permeability
hL Head loss across flow region
hp Pressure head
ht Total hydraulic head
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Symbol Description
hv Velocity head
hz Elevation head
i Hydraulic gradient
k Hydraulic conductivity of soil (various subscripts)
kg Hydraulic conductivity of geotextile across plane of fabric
L Length of flow path
n Porosity of the soil
Nd Number of equipotential (head) drops in the flow net
Nf Number of flow channels in the flow net
O95 Geotextile apparent opening size
q Volumetric flow rate
R Radius of influence in well design
S Surface area factor for grain shape (Kozeny-Carman equation)
tg Geotextile thickness
u Water pressure at the point of interest
vd Discharge velocity
vs Seepage velocity
x Exponent on effective grain size for hydraulic conductivity correlations
y Height of the flow region
z Elevation of a point of interest above the elevation datum
βα Empirical or semi-empirical coefficient relating k to Dα

∆hL Total head loss for one equipotential drop on a flow net
ψg Geotextile permittivity, provided by manufacturers or from testing (ASTM D4491)
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SLOPE STABILITY
7-1 INTRODUCTION.
Slope stability analysis is a common category of analyses in geotechnical engineering

practice. The analysis contains elements of statics, rock or soil mechanics, and
numerical methods. The techniques used range from simple chart solutions to
complicated numerical computer solutions. Regardless of the solution method
employed, the most important element in slope stability analysis is the shear strength of
the soil or rock.
Although rock slope stability and soil slope stability rely on the same basic mechanics,
the modes of failure can be very different. In this chapter, rock slope stability will be
discussed in a separate stand-alone section.
7-2 TYPES OF SLOPES AND MODES OF FAILURE.
There are many different categories of slopes. Natural slopes are those that are
ungraded and the slope geometry is controlled by nature. Figure 7-1 shows some
general cross-sections and failure conditions for natural slopes. If the slope is
steepened by grading or excavation, it is called a cut, a cut slope or an excavated slope.
Dam abutments can be natural slopes or cut slopes.
Embankments constructed of compacted soil form another category of slopes. These

embankments can be highway embankments, dams and levees, fill slopes,
Mechanically Stabilized Earth (MSE) slopes, and others. Figure 7-2 shows cross-
sections through example embankment slopes and details about the failure conditions.
Principal modes of failure in soil or rock are (i) rotation on a curved slip surface
approximated by a circular arc, (ii) translation on a planar surface whose length is large
compared to depth below ground, and (iii) displacement of a wedge-shaped mass along
one or more planes of weakness. Other modes of failure include toppling of rock
slopes, falls, block slides, lateral spreading, earth and mud flow in clayey and silty soils,
and debris flows in coarse-grained soils. Figure 7-1 and 7-2 show examples of potential
slope failure problems in both natural and man-made slopes.
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Figure 7-1 Failure Conditions for Different Cross-sections through Natural
Slopes
Figure 7-2 Failure Conditions in Embankment Foundations and Cut Slopes
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7-3 DEFINITION OF FACTOR OF SAFETY.
The stability of slopes is characterized by the factor of safety, F . Although there have
been various methods of defining the factor of safety found in the engineering literature,
the most common is shown below:
s
F= (7-1)
τ
where:
s = shear strength,
τ = shear stress required for equilibrium.
If the factor of safety is equal to unity, the slope is in a condition of barely stable
equilibrium, right at the point of failure. As the factor of safety increases above unity,
the stability of the slope increases. Slopes having a factor of safety less than one are
considered unstable.
The value of s used in the calculation of factor of safety depends on the strength model
used to characterize the soil, which is often associated with the drainage conditions
assumed as well as the soil type. Table 7-1 shows different strength models that can
be used in the analyses for different soil types and drainage conditions.
The shear stress required for equilibrium ( τ ) is calculated by statics along with
assumptions regarding the conditions for equilibrium and other factors. The number of
unknowns exceeds the number of equilibrium equations in most forms of slope stability
analysis, so assumptions must be made. A major difference in the various methods
used to perform slope stability analyses is the conditions of equilibrium that are
satisfied.
7-4 METHODS OF ANALYSIS OF SOIL SLOPES.
Slope stability analysis in geotechnical engineering practice has evolved over the past
100 years. Initial methods of assessing the stability of slopes involved mapping areas
of instability and determining slope angles from surveys to create landslide hazard
maps. These types of maps are still produced and can be useful in screening potential
stability issues with natural slopes and examining regional landslide risk. Beginning in
the 1920s, assessment procedures were developed that provided a more quantitative
basis for determining the stability of slopes.
Three numerical procedures are used, with varying degrees of popularity, to assess the
stability of slopes: (1) limit equilibrium analysis, (2) finite element and finite difference
analysis, and (3) plasticity analysis.
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Table 7-1 Strength Models for Different Soil Types and Drainage Conditions
Drainage
Soil type s , strength Comments
conditions
Drained conditions are often assumed for coarse-
Coarse-grained grained soils, such as sands and gravels under
Drained =s σ ' ff ⋅ tan(φ ')
soils static loading conditions. Non-linear envelopes can
be used as well.
Used for dynamic loading and certain conditions of
static loading. Can be used for clays and silts as
Coarse-grained well for dynamic or cyclic loading.
Undrained s = s su
soils For normal undrained analysis, the effective stress
strength parameters should be used for coarse-
grained soils.
A non-linear envelope can be used for this case as
Overconsolidated
Drained c ' σ ' ff ⋅ tan(φ ')
s =+ well. Some engineers do not like to use effective
fine-grained soils
stress cohesion for any soil.
Normally The effective stress friction angle should correspond
consolidated Drained =s σ ' ff ⋅ tan(φ ') to the soil in a normally consolidated state. This is
fine-grained soils equal to the fully softened friction angle.
The undrained shear strength can be determined
Fine-grained
Undrained s = su using a variety of laboratory or in situ tests. The
soils (saturated)
magnitude of su can vary with depth.
This strength model is used for partially saturated
soils like compacted clays. The shear strength
Fine-grained
c σ
s =+ ⋅ tan(φ ) parameters should be measured using
soils (partially Undrained ff Unconsolidated Undrained triaxial tests. Only use a
saturated)
linear envelope for the range of stress where it
appears to be appropriate.
Note: σ ' ff = effective stress on failure plane, φ ' = effective stress friction angle, c' = effective cohesion,
σ ff = total normal stress on failure plane, φ = total stress friction angle, c = total stress cohesion,
ssu = undrained steady state shear strength, and su = undrained shear strength
7-4.1 Limit Equilibrium Analysis.
Limit Equilibrium Analysis is the most popular method of analysis to quantify the stability
of soil slopes. The procedure involves dividing the sliding mass into one or more free
bodies, and determining the forces acting on the free bodies using equations for force
and/or moment equilibrium. The shear stress required for a condition of barely-stable
equilibrium ( τ ) is determined for each free body from the analysis of the system of free
bodies. These shear stresses are used with Equation 7-1 to calculate the factor of
safety of the slope.
Figure 7-3 shows the division of slopes into one, three, and multiple free bodies for
analysis using the limit equilibrium method. For these analyses, the failure surface can
be linear, circular, or a combination of linear and arc segments. For most limit
equilibrium analyses, assumptions must be made so that the equilibrium equations can
be satisfied. In many cases, iteration must be performed to obtain a solution, and it
takes a computer program to efficiently apply the analysis method.
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Figure 7-3 Examples of Limit Equilibrium Analysis

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In general, the most common methods used in practice assume that the failure surface
is circular or a surface comprised of many line segments. The methods that solve for all
conditions of equilibrium (force and moment) provide the most accurate answers.
These include Spencer’s Method (Spencer 1967) and the Morgenstern and Price
Method (Morgenstern and Price 1963). Most of the commercially available computer
programs can perform these two methods.
In some cases, the potential critical failure surface may be easily identified by a feature
such as a thin weak seam. In most other cases, it is necessary to search for the critical
failure surface. The common slope stability programs have very robust search routines
for circular and noncircular failure surfaces.
Figure 7-4 shows the formulas and calculations for a slope stability analysis using
Bishop’s Simplified Method. In this method, the soil mass is divided into vertical slices.
The free body diagram for a slice is shown on the figure. This method uses circular
failure surfaces, and the side forces are assumed to be horizontal. Moment equilibrium
is satisfied, but only equations for vertical force equilibrium are used. Horizontal force
equilibrium is not satisfied.
7-4.2 Finite Element Analysis of Slopes.
Use of finite element analysis to analyze slope stability is becoming more popular in
geotechnical practice. A thorough assessment of the use of the finite element method
is presented by Griffiths and Lane (1999). The finite difference method is an alternative
to the finite element method, and both of these assess the slope in a similar fashion.
Instead of calculating a factor of safety as defined by Equation 7-1, the finite element
and finite difference solutions use a strength reduction factor (SRF). The SRF is a
factor by which the cohesion ( c ) and the tangent of the friction angle ( φ ) are reduced to
a point where the solution no longer converges. At the critical SRF, the displacements
increase rapidly and the equations for equilibrium can no longer be solved.
The finite element method has a few advantages over the more common limit
equilibrium method. The failure surface does not have to be identified prior to the
analysis. The FE method actively seeks out the critical failure surface automatically as
part of the analysis procedure. In addition, the equations for equilibrium are satisfied.
7-4.3 Limit Analysis.
Limit analysis is based on the upper and lower bound theorems for the theory of
plasticity. The process involves section of a potential failure surface, and analyzing the
failure mass based on a kinematically admissible velocity field and a statically
admissible stress field. The use of limit analysis for slope stability calculations has been
around for over 40 years, but it has found limited use in geotechnical engineering
practice. Commercially available programs that use limit analysis have recently become
available, and the popularity may increase.
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Figure 7-4 Example Slope Stability Analysis using Bishop’s Simplified Method
for an Effective Stress Analysis
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Figure 7-4 (cont.) Example Slope Stability Analysis using Bishop’s Simplified
Method for an Effective Stress Analysis
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7-5 WATER PRESSURE EFFECTS.
Water pressures have a profound effect on the stability of slopes. As an example, many
failures that occur in natural and constructed slopes occur during periods of heavy
precipitation. There are two categories of water pressures in slope stability analysis: (1)
internal water pressures and (2) external water pressures. Examples of these are
Figure 7-5 Examples of Internal and External Water Pressures in Slope Stability
Analyses
Internal water pressures are the same as pore water pressures. These are the static or
dynamic water pressures that act on the failure plane within the soil mass. Internal
water pressures must be included in slope stability analyses where the soil is
characterized by effective stress or drained shear strength parameters ( c ' and φ ' ).
Internal water pressures are not included in the analysis where the soil is characterized
by total stress or undrained shear strength parameters ( su or c and φ ).
External water pressures are the pressures applied to the free body where standing
water is in contact with the soil mass at locations other than the failure surface.
Examples of external water pressures are the pressures applied by the reservoir on the
upstream slope of a dam or water-filled tension cracks. External water pressures are
included in both effective stress and total stress slope stability analyses.
7-5.1 Incorporating Water Pressures in Computer Analyses.
It is important that the engineer fully understand the way that water pressures are
specified in computer programs that they use, and how to verify that the correct water
pressures are being used. The nomenclature for inputting water pressures is not
consistent between the many different programs available for slope stability analysis.
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The water table or groundwater table is normally defined as a line that connects points
where the pore water pressure ( u ) is equal to zero. Water table is often used
synonymously with phreatic surface. The piezometric surface is a surface connecting
the height or elevation that water would rise in a series of standpipe piezometers. For
hydrostatic conditions with a horizontal water table, the water table, phreatic surface,
and piezometric surface are the same. For hydrodynamic cases where water is flowing,
the piezometric surface can be at a higher elevation than the water table (upward flow
of water) or at a lower elevation than the water table (downward flow of water). An
example of this is shown in Figure 7-6. At Point A, the vertical distance to the phreatic
surface is indicated on the drawing. However, the pore water pressure is controlled by
the distance to the piezometric line, which is also shown on the figure.
Figure 7-6 Approximate Flow Net for Seepage into a Drain Showing the
Difference between the Piezometric Surface and the Phreatic Surface
Internal water pressures can be accommodated in slope stability software in one or

more of the following methods:
(1) Water table or phreatic surface – can be corrected to approximate the

piezometric surface based on the slope of the surface above the specific point.
(2) Piezometric surface
(3) Finite element seepage analysis
(4) Point-by-point entry in x-y coordinates with subsequent interpolation between
points
(5) Pore pressure coefficient, ru
Of these different methods, the finite element seepage analysis is probably the best. It
allows a large density of pore water pressure points to be calculated and interpolation
between these points is sufficiently accurate with most interpolation methods. Use of
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the pore pressure coefficient is perhaps the least accurate of the methods listed and is
normally only used to verify results from historic analyses.
External water pressures can be automatically applied by most programs by the use of
the water table or phreatic surface. As an alternative, external water pressures can be
applied by the use of triangular and rectangular distributed loads. A third method is to
assign the water properties of a soil material, having the unit weight of water but no
shear strength. External water pressures in tension cracks are normally handled
automatically by the computer program, but triangular distributed loads can also be
used to model water-filled tension cracks.
7-5.2 Seepage Forces.
The flow of water through a slope can serve to destabilize slopes. As indicated in
Figure 7-1, the flow of water parallel to the surface of a slope can reduce the factor of
safety to about half of the factor of safety without flowing water. As water flows through
a soil, a seepage force ( S ) is imparted to the soil from the viscous resistance to the flow
of water. The seepage force can be calculated from:
S = i ⋅γ w (7-2)
where:
i = hydraulic gradient, and
γ w = unit weight of water.
The seepage force equation provides a force per volume for the volume where the
hydraulic gradient or head loss occurs. For slope stability analyses, the effect of flowing
water can be handled by (1) calculating the seepage forces for each slice or free-body
and (2) using the buoyant unit weight of the soil below the phreatic surface.
Unfortunately, using this method requires that the calculations be largely done by hand.
The computer programs that are commercially available do not perform these
calculations automatically. The alternative method, which is accommodated by current
computer programs, correctly calculates the factor of safety of slopes where flow is
occurring by using total unit weights of soils below the water table along with boundary
water pressures as discussed above. This method is currently used in geotechnical
7-6 STRENGTH MODELS AND ANALYSIS CASES.
Slope stability analysis cases are often categorized as undrained (or total stress) or
drained (effective stress) analyses. These are often called short-term and long-term
analyses because their respective use is related to the amount of time required for the
soil to consolidate under the changed loading. For this reason, both effective stress and
undrained (total stress) strength parameters can be assigned to the different soils in a
cross section depending on their drainage condition during the duration of loading.
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The general analysis cases are listed below with guidance on strength models to be
employed with each.
7-6.1 End of Construction (Short Term).
End of construction or post-construction analyses are performed to examine the stability

after construction is completed and prior to any dissipation of pore water pressures in
fine-grained soils. An example of a valid end of construction analysis is construction of
a compacted clay embankment over saturated fine-grained soils. A cross section of this
is shown in Figure 7-7a. For this case, the compacted clay embankment would be
partially saturated, and the strength would be represented as a c − φ soil, with the
strength parameters determined from UU triaxial tests. The in situ fine-grained soils
would be essentially saturated, and the strength would be characterized as φ = 0 ,
su = c . The shear strength of the in situ soils could be measured by UU triaxial tests,
DSS tests, laboratory miniature vane shear tests, field vane shear tests, and under
some conditions, cone penetration tests.
7-6.2 Cut Slope in Clay.
The cross section of a cut slope in a stiff clay is shown in Figure 7-7b. The critical time
in the performance of the cut slope occurs long after the cut is made, when the phreatic
surface has reached a steady-state condition. For this type of scenario, effective stress
or drained shear strengths should be used for the clay. If the clay has a relatively low
plasticity ( LL < 40 and PI < 20), it would be appropriate to use the peak drained
strength parameters ( c ' and φ ' ) determined from CU triaxial tests or CD direct shear
tests. The use of a non-linear effective stress envelope for the clay would also be
appropriate (Duncan et al. 2014). If the clay has a high plasticity, contains fissures,
and/or is heavily overconsolidated, then the fully softened shear strength should be
used to account for changes in shear strength that will likely occur over time. The fully
softened shear strength can be measured using remolded test specimens in a direct
shear apparatus. Again, the use of a non-linear envelope would be appropriate.
7-6.3 Steady State Seepage in Dams.
One of the stability analyses required for earth and rockfill dams is the evaluation of the
factor of safety for the condition of steady state seepage (Figure 7-7c). This case
assumes that the reservoir has been at a relatively constant elevation for long enough
that a steady-state seepage pattern has developed. The pore water pressures in the
dam and foundation can be calculated using finite element analysis. The pore
pressures above the phreatic surface are normally assumed to be equal to zero.
Effective stress strength parameters ( c ' and φ ' ) are used in the dam and foundation
soils. The effect stress strength parameters for the dam materials can be measured
with CD direct shear, CD triaxial, or CU triaxial tests on compacted test specimens. In
the example, the strength parameters for the silty sand can be determined using in situ
tests.
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Figure 7-7 Analysis Cases for (a) End of Construction for Embankment on Clay,
(b) Cut Slope in Clay, and (c) Levee or Dam in a Condition of Steady State
Seepage
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7-6.4 Stabilizing Berm for Failed Slope.
Berms are often used to stabilize failed slopes. For example, Figure 7-8 is a cross
section of a slope that failed in a fat clay. A stability berm has been constructed at the
toe of the slope to increase the factor of safety. It is presumed that the failure surface is
known based on inclinometer readings or borings. If sufficient displacement has
occurred on the failure surface, then the appropriate shear strength to use is the
residual shear strength. The residual friction angle ( φ 'r ) or a nonlinear residual
effective stress envelope is best determined by ring shear tests conducted on remolded
test specimens of the fat clay. This should be an effective stress analysis with the
appropriate phreatic surface used in the analysis.
Stability Berm
Fat Clay (CH)
Failure Surface
Figure 7-8 Stabilizing Berm Used to Increase the Factor of Safety of a Failed
Slope
7-6.5 Other Analysis Cases.
Many other analysis cases are analyzed in geotechnical engineering practice that can
be significantly more complex than the simple examples provided. For earth and rockfill
dams, a critical case for the upstream slope is rapid drawdown. This occurs when the
steady state seepage condition is changed by lowering the reservoir. If the reservoir is
lowered rapidly (meaning days or weeks), the upstream slope no longer has the
stabilizing support of the external water pressure. The shear strengths in the dam are
based on the effective stresses prior to drawdown, and the lowering of the reservoir can
cause an undrained failure. This is normally performed as a three-stage analysis, and it
uses a strength model that is more complex than those used in the other cases
described above (Duncan et al. 2014). These types of require considerable technical
ability and they should be performed by engineers who have experience with rapid
drawdown analysis. Effective stress rapid drawdown methods, such as those based on
uncoupled transient seepage analyses, should be avoided.
The stability of slopes is also analyzed for cases of earthquake loading. These
analyses can range from simple pseudostatic methods, where a horizontal force is
applied to the free body of the limit equilibrium analysis, to very complex numerical
analyses. Earthquake analyses are another category that require considerable
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judgment and experience to obtain meaningful results and should be conducted by
engineers skilled in this branch of geotechnical engineering.
7-6.6 Back-Analysis of Slopes.
Failed slopes offer a unique opportunity to develop a model of the shear strengths and
ground water conditions at the time of failure. This type of model can be very useful in
designing slope stabilization or in analyzing nearby slopes. When performing a forward
analysis, the most important unknown is the factor of safety for a specific failure surface.
When conducting a back-analysis of a slope, the factor of safety is known ( F = 1), so a
different unknown can be calculated. Normally, the shear strength of a soil layer is the
desired property to be determined from back-analysis. If the back-analysis is performed
on a slope that failed in an undrained condition, then the average undrained shear
strength of a soil layer can be determined from the analysis. If the back-analysis is
performed on a slope that failed in a drained condition, then only one of the effective
stress strength parameters ( c ' or φ ' ) can be determined. Often, the effective stress
cohesion is assumed to be equal to zero, and the friction angle is calculated.
It is important that the other parameters used in a forward analysis be known with
confidence for back-analysis. These include the slope geometry, soil stratigraphy,
shear strength parameters for layers where strengths are not back-calculated, unit
weights, etc. For back-analysis of drained failures, it is important that the pore pressure
conditions at the time of failure be known. The location of the failure surface should
also be known to obtain the most accurate results. Often, the failure surface location
may be known from the results of inclinometer readings. In other cases, the location of
the failure plane can be determined by careful drilling and sampling. If only the head
scarp and toe exit of the failure plane is known, then the failure surface can be
determined from the computer program’s search routine by only searching for surfaces
which go through those two points. If the position of the failure surface is not known, it
is important to search for the critical surface and not to assume the location. If the
location of the failure surface is assumed, then the resulting shear strength back-
calculated will be too low.
7-6.7 Evaluation of Slope Stability Results.
Limit equilibrium slope stability calculations involve thousands of calculations. Since the
adoption of computer programs to perform these calculations, project specifications
often required that hand calculations be used to verify the results of the computer
analyses. While this was possible when simpler methods of slope stability were used,
such as Bishop’s Simplified Method and the Ordinary Method of Slices, this requirement
became impractical for more complex methods, like’s Spencer’s Method and
Morgenstern and Price’s Method. The time required to perform the calculations by hand
for one failure surface for a cross section containing many slices and an advanced
shear strength model has become excessive. In recognition of this, Wright (2013)
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suggested that new project specifications should require that two different slope stability
computer programs should be used to verify analyses. After the minimum factor of
safety for the critical failure surface is determined using one program, a different
program should be used to analyze the same failure surface using the same method. If
the analyses are correct, the factors of safety calculated should be within about 1% of
each other.
It is important for the engineer to be able to examine the forces acting on each slice. In
particular, the forces at the boundary between slices should be in compression, and the
normal force at the base of the slice should be in compression. Many computer
programs allow the line of thrust to be plotted on the cross section. The line of thrust is
a line that is plotted at the point of application of each of the side forces. Ideally, the line
of thrust should be located within the soil mass defined by the slices. If portions of the
line of thrust plot outside of the free body being analyzed, it often indicates tensile
forces between slices or tensile forces at the base of slices. The addition of tension
cracks to the cross section can be used to prevent tensile forces between slices, and
these can be readily accommodated in commercial computer programs. Further
information regarding the utility of the line of thrust can be found in Whitman and Bailey
(1967).
The search routines for common commercial slope stability computer programs analyze
thousands to tens of thousands of failure surfaces. The stability methods that solve for
all conditions of equilibrium have iterative solutions, and the solutions do not always
converge. There can be various reasons why the solutions do not converge, including
geometry issues with invalid failure surfaces, problems with interpolation of advanced
strength models, numerical issues with the solution procedure, etc. Critical failure
surfaces can be missed as a result of non-convergence. The engineer should be able to
identify when convergence issues are present. The location of this information varies
depending on the program that is used, and is sometimes difficult to find. Prior to
performing any analyses for record, the engineer should find this information and
assess the validity of their results.
7-6.8 Slope Stability Charts.
Chart solutions for slope stability analyses have been available since the 1930s. Prior
to the introduction of computers into geotechnical engineering practice, slope stability
charts allowed an approximate solution to be quickly obtained. The charts are also
useful in estimating the critical failure circle, showing the mode of failure (toe circle vs.
deep circle), and other valuable information.
Charts have been developed for many different categories of slope stability analysis,
including drained analyses, undrained analyses, rapid drawdown analyses, infinite slope
analyses, surcharge loading at the crest, tension cracks, and other specialty cases. A
comprehensive set of chart solutions can be found in Duncan et al. (2014). An example
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of a chart for infinite slope analysis is shown in Figure 7-9. Infinite slope analysis is
useful for explaining sloughing failures for slopes having an effective stress cohesion
equal to zero, and to examine the effects of seepage on the factor of safety of slopes.
Charts have historically been used to obtain a quick solution to slope stability problems.
Even with the advent of computer-based slope stability analyses, chart solutions still
provided an approximate factor of safety in less time than required to run a computer
analysis. However, the modern computer programs allow slope stability problems to be
defined and solved very quickly, and the speed advantage of chart solutions has been
diminished. Chart solutions are best suited for slope stability problems that have simple
soil profiles and straightforward strength interpretations. As the soil profiles and
strength interpretations become more complex, the accuracy of chart solutions
decreases. Even so, chart solutions still offer a viable complementary analysis method
to computer solutions.
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Figure 7-9 Chart Solution for Infinite Slope Analysis (after Duncan et al. 2014)
7-7 SLOPE STABILIZATION.
It is often necessary to increase the factor of safety of existing slopes or to repair slopes
that are moving or have failed. Figure 7-10 shows different methods of stabilizing
slopes.
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Figure 7-10 Methods of Stabilizing Slopes
7-8 REQUIRED FACTOR OF SAFETY FOR SOIL SLOPES.
There are many different sources that specify the minimum factor of safety. Often, the
design values are determined by municipal or government organizations. For earth
dams, the U.S. Army Corps of Engineers’ EM 1110-2-1902 (2003) recommends the
values listed in Table 7-2. Other organizations dealing with earth dams, such as the
U.S. Bureau of Reclamation (USBR), have specified their own values. The values
recommended by USBR are given in Table 7-3.
Table 7-2 Factors of Safety for New Earth and Rockfill Dams (USACE 2003)
Analysis Condition Required Fmin Slope

End of Construction 1.3 Upstream and downstream
Steady state seepage (Long term) 1.5 Downstream
Maximum pool level 1.4 Downstream
Rapid drawdown 1.1 to 1.3 Upstream
Table 7-3 Factor of Safety for Dams using Spencer’s Method for Dams
(USBR 2011)
Minimum
Loading Shear Strength
Pore Pressure Characteristics factor of
Condition Parameters
safety
Generation of excess pore pressures in
embankment and foundation materials with
laboratory determination of pore pressure and 1.3
monitoring during construction.
Effective embankment and foundation materials and no field
End of 1.4
monitoring during construction and no laboratory
construction
determination
embankment only with or without field monitoring 1.3
during construction and no laboratory determination
Undrained Strength 1.3
Steady-state Steady-state seepage under active conservation
Effective 1.5
seepage pool
Effective or Steady-state seepage under maximum reservoir
1.2
undrained level (during a probably maximum flood)
Rapid drawdown from normal water surface to
Operational inactive water surface 1.3
conditions Effective or
undrained Rapid drawdown from maximum water surface to
active water surface (following a probable maximum 1.2
flood)
Effective or Drawdown at maximum outlet capacity (inoperable
1.2
undrained internal drainage; unusual drawdown)
Other Construction modifications (applies only to
Effective or
temporary excavation slopes and the resulting 1.3
undrained
overall stability during construction).
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The required minimum factor of safety is dependent on many different factors, including:
(1) type of structure, (2) type of analysis, (3) consequences of failure, (4) uncertainty
involved with design parameters, (5) frequency of specific loading event, and many
others. An important concept in arriving at a minimum factor of safety involves the
degree of uncertainty associated with the design parameters. Sometimes, the minimum
factor of safety depends on if the analysis has “well-defined conditions” or “poorly-
defined conditions.” Engineering judgment is required to classify a particular site or
project into one of these two categories. In some cases, the designation of “well-
defined conditions” can only be applied for sites that have already been built upon.
In general, well-defined conditions means that the site exploration and field or laboratory
testing program was thorough enough for the engineer to have confidence in the soil
stratigraphy and shear strength interpretation. A poorly-defined condition can occur
when the borings are spread far apart, few laboratory tests have been conducted,
and/or the soil stratigraphy is highly variable. An example of this is the geotechnical
guidelines for a large Washington DC suburb. The requirements for factors of safety for
two different soil formations are given as follows 13:
“For long-term stability, a minimum Factor of Safety (FS) of 1.25 is

required when supported with sufficient field and laboratory
characterization of the slope’s soils. Otherwise, a minimum FS of 1.5 is
required. In case of Critical slope or structure, a minimum FS of 1.5 is
required unless a laboratory measured residual strength test is obtained
and used in the analysis. In this case, a minimum FS of 1.25 is required
when supported with sufficient field and laboratory characterization of the
soils.”
“For long-term stability of the soil formations other than Potomac

Formation clay if slope stability analysis is deemed necessary by the
engineer or if it is required by the County, a minimum Factor of Safety
(FS) of 1.25 is only acceptable when the slope is not critical and the
analysis is supported with sufficient site-specific in situ or laboratory
strength tests of the encountered soils. Otherwise, a minimum factor of
safety of 1.5 must be used in the analysis.”
7-9 MECHANICALLY STABILIZED EARTH SLOPES.
Mechanically stabilized earth (MSE) is a term that refers to soil retention structures that
include both retaining walls and earth slopes. This section discusses the application of
MSE technology to slopes. The design of MSE retaining walls is discussed in DM 7.2.
13 This example is intentionally left uncited to maintain anonymity of the source.

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Design and analysis of MSE slopes is a specialty area in geotechnical engineering.
There are engineering consultants who specialize in MSE walls and slopes. In this
manual, the rudiments of design and analysis are presented in order that outside
designs can be evaluated and not for the purpose of completing a full MSE design.
7-9.1 Applications of MSE.
Reinforced earth slopes are fill structures in which discrete layers of geosynthetic or
steel elements are installed during construction at specified locations. A typical cross
section of an MSE slope is shown in Figure 7-11.
Figure 7-11 Typical Cross-Section of an MSE Slope
The reinforced soil zone is that portion of the slope in which layers reinforcement are
installed. The retained soil zone and foundation soil zone are located behind and below
the reinforced soil zone, respectively. The layers of primary reinforcement shown in
Figure 7-11 resist the development of failure planes through the reinforced soil zone.
The layers of secondary reinforcement prevent surficial failure at the slope face.
The primary limitations on the use of MSE slopes relate to constructability and utilities.
Constructability is not typically an issue if the slope is part of a larger fill area but may be
a concern when an existing hillside must be excavated to build the slope. Utilities can
be significant factors in the poor performance of MSE structures, particularly if the utility
is wet such as a storm sewer or water main. The malfunction of wet utilities may
contribute to about one-third of the failures of MSE retaining walls (Valentine 2013).
The distinction between an MSE retaining wall and an MSE slope has been defined by
the Federal Highway Administration (FHWA) based on the face angle, θ (FHWA
2009b). If θ < 70 , then the structure is a slope. If θ ≥ 70 or greater, then the structure
is a wall. The distinction between an MSE retaining wall and slope has important
implications for land use efficiency as illustrated in Figure 7-12. However, the increased
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land use efficiency of MSE walls and slopes is accompanied by complications related to
facing requirements and structural deformations, as well as increased cost.
Figure 7-12 Difference in Usable Land for Walls and Slopes
7-9.2 Reinforced Slope Materials.
The soil that is installed in the reinforced zone of an MSE slope is an important
structural component of the slope. The properties that are required of reinforced soil
should be based on the geometry of the slope and the structures that may depend on
the slope for support. FHWA’s recommendations for the properties of fill soil in the
reinforced zone of MSE slopes permit up to 50% fines that have a PI ≤ 20 (FHWA
2009b). The fill properties shown in Table 7-4 are recommended for relatively tall or
steep slopes; however, the recommendations are only applicable to slopes with a height
less than 70 feet.
Table 7-4 Recommendations for Reinforced Fill Soil in MSE Slopes

Based on Geometry
Slope Geometry Recommended Properties for Reinforced Fill

Sieve Size Percent Passing
Slope < 1.2H:1V 4 in 100
Gradation
(θ < 40° ) ASTM D6913
No. 4 20-100
No. 40 0-60
and No. 200 0-50
H < 70 ft Plasticity Index
PI ≤ 20
ASTM D4318
U. S. Standard Sieve Percent Passing
1.2H:1V ≤ Slope < 0.36H:1V 4 in 100
Gradation
( 40° ≤ θ ≤ 70° ) ASTM D6913
No. 4 20-100
No. 40 0-60
and No. 200 0-35
H < 70 ft Plasticity Index
PI ≤ 10
ASTM D4318
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Other important components of MSE walls and slopes include internal and external
drains, filters, separators and erosion control. The use of geosynthetics for these
applications is summarized in Table 7-5.
Table 7-5 Summary of Applications and Materials for Reinforced Soil Slopes
Application Component Material Purpose Comments
Polyester (PET) and
Polypropylene (PP) Provide tensile PET and HDPE are usually used
Primary and
Geotextile; PET, PP and strength and for primary reinforcement. PET,
Reinforcement Secondary
High-Density confinement to HDPE and PP can be used for
Reinforcement
Polyethylene (HDPE) fill soil. secondary reinforcement.
Geogrid
Soft Armor Consult manufacturers for
Face Rolled Erosion Control recommendations of RECP
Product (RECP) Prevent specifications based on θ and
(θ < 40° ) erosion service period.
caused by Typically, galvanized WWF 4x4-
surface water W4.0xW4.0 should be used.
Welded Wire Fabric runoff Hardware cloth required behind
(WWF)
Facing WWF to prevent spilling of
Hard Armor retained gravel fill.
Face
Nonwoven PP Separation Separate gravel fill at slope face
(θ ≥ 40° ) Geotextile and filtration from finer reinforced soil.
Fill soil
GP or GW with 1.0 in minimum
Gravel immediately
particle size.
behind WWF
Drainage Typically, ASTM C33 No. 57 or

Gravel
medium No. 67 stone.
Blanket Drain
Install above and below drainage
Nonwoven PP Separation
gravel to separate from adjacent
Geotextile and filtration
finer grain soil.
Internal
Typically, ASTM C33 No. 57 or
Drainage Drainage
Gravel No. 67 stone. Can be replaced
medium
by drainage composite.
Chimney Drain Use drainage composite with
geonet core to replace drainage
Drainage
Drainage Composite gravel. Space drainage
medium
composite to typically provide
33% to 75% coverage.
Line swale with TRM. Consult
manufacturer for specifications.
RECP in form of Turf Divert and Locate swale 5 ft. to 10 ft. behind
External Drainage
Reinforcement Mat control surface slope crest. Size swale based on
Drainage Swale
(TRM) water runoff hydraulic analyses. Use bench
with swale at mid-slope for H
over 25 to 30 ft.
7-9.3 Geosynthetic Reinforcement Strength.
Geosynthetics that are used in soil reinforcement applications are typically designed to
exhibit their maximum tensile strength in one direction. In this respect, such
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geosynthetics are said to be uniaxial and the design strength direction usually
corresponds to the material’s MD or roll-direction. Geosynthetics that exhibit significant
tensile strength in both the MD and cross machine direction (XMD) are said to be
biaxial.
The tensile strength of a geosynthetic that is used for the design of an MSE slope is
based on a minimum average roll value (MARV) that is reported by the manufacturer.
In the United States, the industry practice is to reduce the average value by two
standard deviations and to define the result as the minimum average roll value (MARV).
Reduction factors are then applied to the strength MARV to account for potential
degradation of strength as a result of creep, environmental conditions and installation
damage.
7-9.3.1 Long-Term Design Strength.
The assessment of a geosynthetic’s long-term tensile strength ( Tal ) for use in the design
of an MSE slope follows current FHWA procedures (FHWA 2009b):
TULT
Tal = (7-3)
RFCR × RFD × RFID
where:
TULT = ultimate tensile strength of the geosynthetic based on the MARV,
RFCR = reduction factor applied to account for creep under sustained tensile loading,
RFD = reduction factor applied to account for the degradation due to environment, and
RFID = reduction factor applied to account for damage during installation.
The reduction factors are discussed briefly in the following sections. Details regarding
the determination of RFCR , RFD , and RFID for a geosynthetic can be found at
Appendix B of FHWA (2009b).
7-9.3.2 Reduction Factor for Creep ( RFCR ).
The tendency of a geosynthetic to elongate under sustained tensile loading is called

creep and it is a property of materials that are manufactured using PET, HDPE, PP and
other polymers. If the magnitude of the load is sufficiently great and if it is maintained
for a sufficient period of time, the creep can induce rupture or result in such elongation
that the material’s performance is compromised.
In the United States, the standard of practice to determine a reduction factor for creep
( RFCR ) that is based on the sustained load that will induce creep rupture at the end of
the design service period. For permanent MSE slopes the design service period should
be no less than 75 years. A longer service period may be appropriate for structures that
support critical infrastructure.
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7-9.3.3 Reduction Factor for Durability ( RFD ).
Geosynthetics may degrade depending on their base polymer and if they are exposed
to certain environmental conditions. Polyester geosynthetics may degrade as a result of
hydrolysis. Geosynthetics manufactured with HDPE and PP are subject to degradation
by their reaction with oxygen, particularly in the presence of elevated temperatures.
Oxidation can also be initiated by exposure to UV light (i.e., UV-oxidation). The
resistance of these polymers to oxidation can be significantly increased by the addition
of antioxidants during the manufacturing process. Polyester is also susceptible UV-
oxidation but to a lesser degree than HDPE and PP (FHWA 2009a). Protection of PET
yarns is typically provided in the form of coatings. In the case of all geosynthetics, the
protective roll wraps should not be removed until the material is installed to minimize UV
light exposure.
The FHWA recommends a default reduction factor for durability ( RFD ) of 1.3 for PET,
HDPE, and PP geosynthetics provided certain criteria are satisfied. Also, a lower RFD
may be used if it is indicated by product specific testing. Further details are given in
FHWA (2009b).
7-9.3.4 Reduction Factor for Installation Damage ( RFID ).
The standard of practice in the United States is for the manufacturers of geosynthetic
reinforcement to assess the potential for installation damage through the performance
of full-scale tests. Such testing programs typically evaluate the damage induced by
compaction of coarse gravel, sandy gravel and silty or clayey sand.
The geosynthetic manufacturer should be consulted to obtain its recommendation for

the reduction factor for installation damage ( RFID ). The manufacturer should also be
consulted for its recommendations regarding measures to reduce installation damage.
Typical measures include the following:
• Tracked vehicles should not traffic directly on panels of geosynthetic

reinforcement. There should be no less than 8 inches of fill soil between the
tracks and the geosynthetic. Sharp turns by tracked vehicles on fill soil should be
avoided.
• Rubber tire vehicles may operate directly on the geosynthetic reinforcement at
speeds less than 10 miles per hour. Sudden braking and turning should be
avoided.
• Fill soil should not be dumped directly onto geosynthetic reinforcement. Rather,
it should be dumped onto fill soil that has already been spread and then bladed
onto the geosynthetic by a dozer.
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7-9.4 Soil-Geosynthetic Interaction.
The determination of geosynthetic reinforcement length in the design of an MSE slope

is based in part on the resistance of the reinforcement to pullout from between layers of
confining soil. In FHWA (2009b) the FHWA defines a geosynthetic’s resistance to
pullout as:
Pr = F ∗ ⋅ α ⋅ σ 'v ⋅ Le ⋅ C (7-4)
where:
Pr = geosynthetic reinforcement’s resistance to pullout,
F ∗ = pullout resistance factor,
α = scale correction factor to account for nonlinear stress reduction,
σ 'v = effective vertical stress at the soil-reinforcement interface,
Le = length of reinforcement embedded behind the trial failure surface, and
C = number of surfaces on which pullout resistance is mobilized (i.e. 2 for
geosynthetics).
The critical failure surface used to calculate Le should be that surface that exhibits the
minimum F deemed acceptable. Also, Le should be no less than 3 feet to assure
adequate pullout resistance.
Some manufacturers of geosynthetic reinforcement have characterized soil-

geosynthetic interaction in terms of a coefficient of interaction ( Ci ) based on pullout
tests with Ci defined as:
tan δ
Ci = (7-5)
tan φ '
where:
φ ' = the effective stress internal angle of friction, and
δ = the effective soil-geosynthetic interface friction angle.
The values of F ∗ and Ci are related by:
F ∗ = Ci tan(φ ) (7-6)
7-9.5 Analysis and Design of Reinforced Slopes.
The most technically challenging aspect of the design of an MSE slope is deciding the
required strength, length and vertical spacing of the layers of reinforcement. The critical
failure surface for a slope with layers of horizontally-oriented geosynthetic reinforcement
is frequently a combination of circular and linear segments, particularly if the critical
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failure surface is entirely outside of the reinforced soil zone (i.e., a global failure surface
rather than a compound failure surface).
Most of the software packages that are used for soil slopes can also be used for MSE
slopes. There are some software packages specifically written for MSE slopes, and
these are listed in Appendix B. Spencer’s Method (Spencer 1967) and the Morgenstern
and Price Method (Morgenstern and Price 1963) both lend themselves to the analyses
of noncircular failure surfaces, solve for all conditions of equilibrium (i.e. moment and
force), and provide the most accurate solutions. Either of these two methods is
preferred for the analysis of MSE slopes.
One of the most important considerations in the modeling of an MSE slope in a limit
equilibrium slope stability computer program is the definition of F given in Equation 7-1.
However, the presence of geosynthetic reinforcement requires that its strength be
applied to the right side of the equation in the numerator or the denominator (Duncan et
al. 2014). Two options are available as summarized in Table 7-6. The method used by
slope stability software can be determined using the simple approach suggested by
Duncan et al. (2014).
Table 7-6 Methods of Incorporating Geosynthetic Reinforcement Strength in

Factor of Safety Equation
Method of Including
Factor of Safety Equation
Reinforcement Strength
shear strength
Method A (Active) F =
shear stress required for equilibrum − reinforcement resistance
soil strength + reinforcement resistance

Method B (Passive) F =
shear stress required for equilibrium
If Method A is used to define F , then the strength of the geosynthetic used to calculate
F is Tal . To account for potential uncertainties in the geosynthetic Tal should be
divided by a factor of safety for geosynthetic strength ( FR ) of at least 1.3. If Method B is
used, then Tal will be reduced by F and the application of FR is not necessary.
The reinforcement force orientation assumed by limit equilibrium analysis can vary from
one that is parallel to the reinforcement to one that is tangent to the slip surface. Setting
the orientation parallel to the reinforcement in a slope stability computer program is
common practice and tends to result in a lower F compared to setting the orientation
tangent to the slip surface.
A detailed discussion of the procedure to determine the reinforcement requirements of

MSE slopes is provided by the FHWA (2009), particularly for the case in which Bishop’s
simplified method is used. The FHWA manual also considers the mechanics of internal
sliding failure and locally soft foundation soil at the slope toe. The steps provided in
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Table 7-7 are intended to help an engineer to construct a computer model for in MSE
slope design.
Table 7-7 Steps for Designing an MSE Slope

Step Procedure
Draw a scaled cross section of the slope that reflects the existing and proposed grades as well as
1 external water conditions, and permanent and temporary loads. Use high quality site plans to locate
these features as accurately as possible.
Use the available geotechnical information to determine and draw the boundaries of soil and rock strata
2
as well as groundwater.
3 Use the results of Steps 1 and 2 it to construct the cross section in a computer program model.
Assign physical, strength and hydraulic properties to the sections of the model as indicated by the
4
available geotechnical information.
5 Select either Spencer’s method or the Morgenstern and Price method to analyses failure surfaces.
Select one to three values of geosynthetic reinforcement strength ( Tai ) based on manufacturer
6
information, or assume typical values.
Determine whether the computer program used Method A or B to defined F. If the program provides an
7
option, select Method A and then reduce the geogrid strength by FR=1.3 or more.
8 Set the reinforcement force orientation parallel to the reinforcement.
Assign the F ∗ or Ci parameter based on recommendations by the manufacturer of the geosynthetic
9 reinforcement. If recommendations are not available or if they are not supported by test data, assume
that Ci = 0.67. This will be conservative for the soil parameters shown in Table 7-4.
If the computer program provides an option, select 100% reinforcement coverage as opposed to partial
10
coverage.
Assign layers of geosynthetic reinforcement to the cross section. Vertical spacing of 3 ft. is a good
11
initial starting point. In general, the vertical spacing of reinforcement should not exceed 3 ft.
Set the length ( L ) of each layer of reinforcement to about 0.7 H . The required L will increase if there
12
is a slope below (i.e. a toe slope) or behind (i.e. a crest slope) the MSE structure.
Perform preliminary analysis by setting the search limits to evaluate only those surfaces which exit
through the face of the slope. Evaluate both circular and noncircular failure surfaces. If the resulting F
13
is too low, then change the layers in the vicinity of the bottom of the failure surface to types with higher
strengths.
After designing for failure surfaces that exit through the slope face, change the search limits to evaluate
14
compound and global surfaces that pass below the toe of the structure.
If a compound failure surface is indicated that has an unacceptably low F , then increase the strength
15 of the lower reinforcement layers or decrease their vertical spacing. If a global failure surface is
indicated that has an unacceptably low F , then increase the length of the reinforcement layers.
Using the process in Table 7-7, the modifications required to obtain a satisfactory F
that is balanced with geosynthetic efficiency is an iterative process in which
geosynthetic strengths and lengths are adjusted. Once the optimum design is
determined, the parameters that may have significant uncertainty should be considered.
For example, if rock is thought to be present below the toe of the MSE slope, changing
the elevation of the top of rock by a few feet in the computer model can have a profound
effect that makes the difference between a stable slope and slope failure. Similarly,
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there is frequently significant uncertainty regarding the location of groundwater and soil
shear strength. The effect of these uncertainties should be investigated through
parametric analyses coupled with engineering judgment.
The designer is cautioned to avoid making the design over-complicated. Small savings
in material obtained from optimized lengths or spacings are often offset by potential for
error in construction or an increased difficulty in constructability.
7-9.6 Required Factor of Safety for MSE Slopes.
An MSE slope should be designed to a target F that is based on considerations of the

uncertainties regarding site conditions, material properties and the consequences of
slope failure. In general, the standard of practice is to provide a F in the range of 1.3 to
1.5. If the site conditions and material properties are understood well and if the
consequences of failure are relatively low, then a minimum F of 1.3 may be
appropriate. However, if site conditions are subject to unforeseeable changes and if soil
types, strengths and locations are poorly understood, then F of 1.5 or higher may be
necessary. Similarly, if the proper performance of the MSE slope is required for the
operation of important structures, then F of 1.5 or higher may be indicated. For MSE
slopes designed in accordance with FHWA guidance, a minimum factor of safety of 1.5
is required.
7-10 ROCK SLOPE STABILITY.
The stability of rock slopes may become a concern during the excavation for the
construction of roads, buildings and infrastructure components. Often a stability issue
cannot be identified until the excavation is underway and information that is needed to
assess the potential for various modes of failure becomes available. Analyses of rock
slope stability may need to be performed expeditiously to avoid project delays. The
potential for delays can be exacerbated by a need to design stabilization measures. In
other cases, rock slope instability does not occur until well after the initial excavation
and rock weathering has taken its toll.
This section provides an overview of some of the aspects of rock slopes and a more in-
depth discussion of others. The fundamental mechanics of rock slope failure are
covered by a discussion of sliding blocks, plane failure, wedge failure, and toppling
failure. Stabilization measures for rock slopes and mitigation of rock falls are also
addressed. More in-depth discussion can be found in FWHA (1998), Hoek and Bray
(1981), and Rowland et al. (2007).
7-10.1 Modes of Rock Slope Failure.
Fortunately, the stresses in most rock slopes are much less than the rock strength, and
for this reason most rock slopes are relatively stable. The potential for rock slope failure
becomes a concern under two general conditions. First, discontinuities in the rock mass
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propagate and the rock separates as blocks, wedges, columns, or other types of
sections. Second, rock that has already separated in the form of cobbles and boulders
can translate downslope under the influence of gravity as a rock fall. In both cases, the
separated rock may pose a hazard to both property and lives.
There are six typical configuration of rock slopes, some of which may pose risks of
instability. Four possible configurations of rock discontinuities are shown in Figure 7-13
while rock slopes with weak and weathered rock are depicted in Figure 7-14.
Figure 7-13 Rock Discontinuity Conditions (after FHWA 1998)
The term discontinuity refers to faults, joints, bedding planes, or any other surface upon
which rock may move. The pattern of discontinuities shown in Figure 7-13a is typical for
sedimentary rock, such as limestone, sandstone, and shale, that has been deposited in
bedded layers and later uplifted by geologic processes. The orientation of the
discontinuities in Figure 7-13a is roughly parallel to the rock slope face, but they do not
daylight at the slope face. That is, the discontinuities do not extend to and intersect the
exposed surface of rock at the slope face. In such a configuration the rock slope face is
expected to remain stable. In contrast, Figure 7-13b shows a rock slope with
discontinuities that daylight at the slope face. With such orientation of discontinuities, a
potential for rock slope failures exists.
Figure 7-13c shows a rock slope with generally favorable bedding in that the
discontinuities dip into the rock slope (dip is discussed later in this section) and there is
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little potential for rock to slide out of the slope face. However, blocks of rock at the
slope face may become instable when discontinuities daylight at the slope face. The
potential for the development of such conditions are increased if blasting was performed
during slope excavation.
A rock slope configuration that illustrates the conditions for the toppling of rock columns
is shown in Figure 7-13d. Toppling becomes a risk for rock with relatively thin bedding
with steeply dipping discontinuities. The stability of rock columns can degrade relatively
quickly if water seeps readily into the discontinuities from surface water runoff and
increases the rate of weathering. In regions with frequent freeze-thaw cycles the rate of
degradation may accelerate further because the frozen water can cause the rock
columns to separate.
Sandstone and shale are often found with near-horizontal bedding. Excavation of such
a bedding sequence generally results in a shale layers that weather faster than
sandstone. In such conditions layers of sandstone may be undermined and form ledges
that are prone to fracturing and failure. Such a condition is illustrated in Figure 7-14a.
Weak rock slopes with closely spaced impersistent joints may fail along a circular or
noncircular surface much like a soil slope as shown in Figure 7-14b.
Figure 7-14 Weathering and Weak Rock Conditions (after FHWA 1998)
An assessment of the potential for one of the above modes of failure to develop often
requires a geologic investigation, field mapping of discontinuities, stereographic
projection of geologic data, and an evaluation of rock strength. Each of these tasks
represent significant sections in comprehensive publications on the topic of rock slope
engineering.
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7-10.2 Mechanics of a Sliding Block.
The mechanics of a sliding block are central to two types of rock slope stability
analyses. Considering the rock slope depicted in Figure 7-15, the weight of the block
ABC is represented by force W , which acts through the block’s centroid. The
component of W that acts perpendicular to the sliding plane AC is FN . The component
of W that acts parallel to the sliding plane AC is FS . The relationship of these three
forces is defined by the angle, or dip, of the sliding plane ( Ψ P ) with respect to the
horizontal plane.
Figure 7-15 Rock Slope with Sliding Block
The factor of safety against sliding on plane AC is calculated as shown Figure 7-15.
This solution assumes no pore pressure is acting on the sliding plane. Positive pore
pressure will reduce the effective normal stress, shear strength, and the factor of safety.
7-10.3 Plane Failure.
The rock slope with a sliding block shown in Figure 7-15 is a simple version of a plane
failure. It is not a type of failure that is often encountered because the conditions for its
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development rarely occur. However, the mechanics of a plane failure also apply to the
more frequently encountered wedge failure. For this reason, it is instructive to further
consider plane failures.
7-10.3.1 Sloped Surface Orientation Terms.
A discussion of plane failure requires the definition of three terms that are used to
describe the orientation of sloped planes such as a slope face or a potential failure
surface. The dip ( Ψ ) of a sloped plane is the inclination of that surface as measured
from a horizontal plane, as shown in Figure 7-16. The dip direction or dip azimuth ( α )
is the direction of the horizontal trace of the line of dip measured clockwise from north.
Often the term strike is used to describe the orientation of a sloped plane. It is the
direction of a line that is formed by an intersection of the sloped plane with an imaginary
horizontal plane. The orientation of the strike of a sloped surface is perpendicular to the
dip direction of the sloped surface.
Figure 7-16 Definition of Sloped Surface Orientation Terms
7-10.3.2 General Conditions for Plane Failure.
For sliding to occur on a single plane the four conditions must be satisfied (Hoek and
Bray 1981):
1. The plane on which sliding occurs must strike nearly parallel (i.e., within about
20º) to the slope face.
2. The failure plane must daylight in the slope face (i.e., intersect the slope face).
Therefore, the dip of the failure plane ( Ψ P ) must be less than the dip of the slope
face, Ψ F . That is, Ψ P < Ψ F .
3. The dip of the failure plane must be greater than the angle of friction at the failure
plane. That is, φ ' < Ψ F .
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4. Release surfaces that provide negligible resistance to sliding must be present in
the rock mass to define the lateral boundaries of the slide.
The relative positions of the planes defined by Ψ F , Ψ P and φ ' are shown in Figure
7-17a. The release surfaces associated with a sliding plane are shown in Figure 7-17b.
Figure 7-17 Geometry for Plane Failure (after Hoek and Bray 1981)
7-10.4 Plane Failure Analyses.
Rock slopes analyzed for plane failure can also consider the presence of a tension
crack, which may contain water. The crack may be located either above or below the
slope crest as shown in Figure 7-18. The uplift force applied by water at the failure
plane is designated as U . The force applied by water in the tension crack is designated
as V . The factor of safety for such conditions can be calculated using the equation
presented in Figure 7-18.
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Figure 7-18 Rock Slopes with Tension Cracks (after Hoek and Bray 1981)
Further details regarding the stability analyses of sliding planes are described in the
FHWA (1998) and by Hoek and Bray (1981). While such analyses can be practically
performed using hand calculations, analytical efficiency can be significantly improved by
use of computer programs.
7-10.5 Wedge Failure.
The rock slope with a sliding wedge shown in Figure 7-19 is similar to a sliding plane,
but the presence of Plane A and Plane B make it possible to model a geometry that is
encountered in the field more frequently.
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Figure 7-19 Rock Slope with Sliding Wedge (after Hoek and Bray 1981)
The convention adopted in the FHWA (1998) and by Hoek and Bray (1981) is that the
release surface designated as Plane A is the flatter of the two release surfaces. The
steeper release surface is designated as Plane B. These two surfaces intersect along
the line of intersection.
As with a plane failure, there are certain geometrical requirements for wedge failure to
occur. Specifically, Ψ Fi > Ψ i > φ ' , where Ψ Fi is the inclination of the slope face as
measured at right angles to the line of intersection, Ψ i is the dip of the line of
intersection and φ ' is the average friction angle of Plane A and Plane B. Note that Ψ Fi
is not the same as Ψ F unless the dip direction of the line of intersection is the same as
the dip direction of the slope face (Hoek and Bray 1981).
The factor of safety of the wedge in Figure 7-19 may be determined by assuming that
sliding is resisted only by friction at the surface of Plane A and Plane B. With this
simplifying assumption, the resisting forces on these planes can be determined by
calculation of the normal forces RA and RB on each plane, as illustrated in Figure 7-20a.
The component forces of the weight of the wedge perpendicular and parallel to the line
of intersection are shown in Figure 7-20b.
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Figure 7-20 View of Wedge Geometry (after Hoek and Bray 1981)
Hoek and Bray (1981) relate the factor of safety against wedge failure ( FW ) to that for a
slope with a face that is inclined at Ψ Fi and a failure plane that is inclined at Ψ i by a
wedge factor K . This wedge factor can be graphically determined using a figure
provided in both the FHWA (1998) and Hoek and Bray (1981).
The analysis of a wedge failure for which cohesion or water are present is more
complicated than for a wedge failure in which only the friction angle at the failure planes
is considered. Both the FHWA (1998) and by Hoek and Bray (1981) discuss this
analytical process in detail. Such an analysis can also be performed efficiently using
the computer programs.
7-10.6 Toppling Failure.
For a column of rock to be subject to toppling failure, the center of gravity of the column
must be located on the side of the column at the slope face. In this way, the column is
loaded eccentrically. Such loading creates tensile stress on the side of the column
away from the slope face, as shown in Figure 7-21. If the tensile capacity of the column
is exceeded, failure can ensue.
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Figure 7-21 Rock Slope Subject to Toppling Failure
Analyses of rock toppling can be significantly more complicated than those for plane
failure or block failure. A method for hand calculations is described in FHWA (1998) but
such a procedure may have limited applicability to actual field conditions.
7-10.7 Circular Failure.
As described in section 7-10.1, if a rock slope consists of weak material with closely
spaced, impersistent joints, then the slope may fail along a circular or noncircular
surface much as a soil slope. Analyses of these types of failures can be performed
using the methods previously discussed for the stability of soil slopes. However, it
should be noted that even weak intact rock can exhibit significant cohesive strength. An
accurate assessment of the actual cohesive strength may be difficult to make. It is also
important to realize that zones of relatively strong rock may exist behind zones of
relatively weak rock. The presence of the strong rock zones may significantly affect the
location of the critical failure surface.
7-10.8 Rock Slope Stabilization and Protection.
Several measures can be taken to mitigate the hazards presented by unstable rock
slopes. These measures vary from the relatively simple to those which require
considerable analytical expertise, construction skill, and expense.
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7-10.8.1 Stabilization and Protection Options.
The range of stabilization measures that are typically available include rock
reinforcement, rock removal and protective barriers. A range of options is available
under each of these categories as shown in Figure 7-22.
Figure 7-22 Rock Slope Stabilization and Protection Measures (after FHWA 1998)
7-10.8.2 Reinforcement
Anchors have been used for many years to stabilize both soil and rock slopes. In
general, anchors can be classified as passive or active. A passive anchor that is
frequently used for top-down excavation stabilization is the soil nail. For rock slope
stabilization it may be referred to as a rock bolt or dowel. It typically comprises a steel
tendon in the form of an all-thread bar that is centered within a drill hole. A cement
grout is installed in the drill hole to bond the tendon to the adjacent soil or rock. Grout is
usually placed by tremie from the distal end to the anchor head. Typically, such holes
are drilled at a declination of about 15° or more below the horizontal plane to prevent
spilling of the grout at the excavated face.
The steel bars used for these applications typically exhibit a tensile capacity of at least
75 ksi. Higher capacity bars are readily available. Typical bar diameters correspond to
standard reinforcement steel sizes of #8 (i.e. 1.0-inch nominal diameter) to #11 (i.e.
1.41-inch nominal diameter). Resistance to bar corrosion is typically provided by an
epoxy coating or galvanization. The grout that surrounds the bar can also provide some
protection against corrosion, but the grout is subject to cracking and may not provide
complete coverage.
Different types of liquid resins are manufactured for use with rock anchors as an
alternative to cement grout. Resin cartridges include a hardener that can be selected to
provide a range of hardening times. The cartridges are inserted into the drill hole and
then followed by insertion of the steel bar tendon. The tendon is then spun at a
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prescribed rate to mix the resin and hardening agent and to distribute the mixture to
both bar and rock surfaces. Unfortunately, there is some skepticism within the ground
anchor contracting industry that resins can be consistently distributed to rock and
tendon surfaces and can reliably provide requisite bond capacities. Also, resins do not
provide the same level of corrosion protection as does cement grout.
The FHWA has published several manuals on the design of soil nail structures. The
current version is Geotechnical Engineering Circular No. 7 (FHWA 2015). This manual
should be consulted in the design of passive anchors for rock stabilization applications.
A passive anchor does not impart a stabilization force to the adjacent rock or soil until
the rock or soil tends to displace. At that point the passive anchor provides a resisting
force. In contrast, an active anchor is stressed as part of its installation process. Such
anchors may comprise steel bars like a soil nail. However, if a particularly long anchor
or one with high tensile capacity is required, then steel strand tendons may be
indicated. The installation of such an active anchor is similar to that of a passive
anchor. First, a hole for the anchor is drilled and then the strand is inserted and
centered within the hole. Next, the strand is grouted, but only for a certain length of the
strand starting from the distal end. A length between the top of the grouted section and
the anchor head is left unbonded. The reason for the unbonded section is that this
portion of the tendon must be left to strain under a design stressing load to provide an
active force at the anchor head. This force is transferred to a loading plate or block that
is secured against the rock slope face. In this way the active force can be used to
stabilize a large rock plane, wedge or unstable columns. This is perhaps the most
important distinction between an active anchor and a passive anchor. Unlike a passive
anchor, an active anchor does not depend on soil or rock movement to mobilize its
strength.
The bonded length of the tendon is that portion which is grouted. The length of the
bonded section is based on analyses that consider the load in the anchor and the bond
strength between the grout and the adjacent soil or rock.
Steel strand tendons are typically available in configurations that provide a working
tensile capacity of about 35 to 500 kips. Tendons with considerably higher capacities
can be fabricated.
The design of active anchor systems is discussed in the FHWA’s Geotechnical

Engineering Circular No. 4 (FHWA 1999). This manual should be consulted in the
design of active anchors for rock stabilization applications.
When anchors are designed for the stabilization of rock slopes, some consideration
should be given to the potential for conditions that may cause steel corrosion. Such
conditions often prevail in areas where coal and acidic runoff are present or where sodic
and pyritic soils are found. The FHWA provides the electrochemical parameters in
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Table 7-8 as limits for the use of steel reinforcement in MSE structures. These limits
should be considered when designing the corrosion protection for active and passive
anchors.
Table 7-8 Recommended Limits of Electrochemical Properties for Reinforced

Fill with Steel Reinforcement (after FHWA 2009b)
Property Criteria Test Method
Resistivity >3000 ohm-cm AASHTO T-288
pH >5 and <10 AASHTO T-289
Chlorides <100 ppm ASTM D4327
Sulfates <200 ppm ASTM D4327
7-10.8.3 Shotcrete.
Shotcrete is essentially concrete that has little to no gravel-size particles (i.e. larger than
the No. 4 sieve) that can be sprayed onto vertical and near-vertical surfaces. It is
usually applied in layers to build up the total coating to a specified thickness. Both steel
welded wire fabric (WWF) and bars may be used within the shotcrete to provide
reinforcement.
Shotcrete can be used to stabilize rock slopes that are subject to raveling and
dislodgement of gravel, cobble and boulders. It will tend to adhere to such unstable
faces but it should be secured with relatively short passive anchors (i.e. soil nails, rock
bolts or dowels). Otherwise, the shotcrete may delaminate from the rock slope in a
short period of time. The length of the anchors can be 10 feet or less if they are not
actually needed for stabilization of rock planes and wedges. If the rock slope surface is
mostly unstable then the anchors should be spaced at intervals no greater than about 6
feet.
A common cause for separation of shotcrete from the face of a rock slope is the
presence of water behind the shotcrete. Water frequently seeps from rock slopes, and
it should be drained using a drainage composite. Drainage is especially important in
climates where the water can freeze. Drainage composites can be successfully used
behind shotcrete. The composites typically are available in a width of 4 feet. They
should be installed with a coverage of 50%. A drainage grate should be installed at the
base of each drainage composite strip and daylighted by a weep hole through the
shotcrete.
7-10.8.4 Buttress.
Buttresses have been used for more than 1,000 years to stabilize slopes, walls and
buildings. In general, a buttress is a gravity structure that provides passive resistance
to displacement. For rock slope stabilization a buttress may comprise piled soil or rock,
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or it may be an engineered reinforced concrete structure or an MSE structure. The
engineering analyses of the mass requirements of a buttress are relatively
straightforward and similar to those of a gravity retaining wall.
7-10.8.5 Drains.
Drains are often used in conjunction with rock slope reinforcement measures to counter
the destabilizing effects of water that is retained behind a slope face. In general, relief
drains are drilled using equipment similar to that used for anchors. However, instead of
being angle below the horizontal plane relief drains are typically angle 2° or more above
the horizontal plane.
An unfortunate reality of relief drains is that they often become clogged by the
accumulation of organic material. Removal of the organic obstructions is generally not
practical.
7-10.8.6 Rock Removal.
The stability of rock slopes can often be improved by the removal of material. On a
large scale, such removal may take the form of mass excavation using drilling and
blasting measures followed by dozers, track hoes and haul trucks.
On a smaller scale trimming may be used to more selectively remove problematic

formations such as overhangs (see Figure 7-14), planes, wedges and columns.
Trimming may also employ drilling and blasting and may be preferable to mass
excavation in terms of cost and also in terms of potential disruption of adjacent
transportation or commercial operations.
If the rock slope includes loose cobbles and boulders that may present a rock fall
hazard, then scaling may be the most appropriate method of rock removal. In a scaling
operation, personnel traverse the slope face while secured by ropes and harnesses.
They use hand tools to dislodge rock and may even remove soil deposits and
vegetation
7-10.8.7 Rock Fall Protection Measures.
Measures to protect against rock falls may include the installation of surface restraints,
barriers or ditches. However, the design of any of these measures usually requires an
assessment of the risk posed by rock falls.
Rock fall analyses are generally beyond the capabilities of hand calculations, although
some graphical aids have been developed for this purpose (FHWA 1998). When the
FHWA (1998) was published, the computer-supported analysis of rock falls was at a
relatively early stage. The complexity of the problem was such that computer modeling
had to be paired with videography and surveying to render solutions with meaningful
accuracy.
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After a rock fall analysis has been performed and a reasonable estimate of the final
location of fallen rocks can be obtained, ditches can be designed to capture errant
cobbles and boulders. Guidelines for the dimensioning of capture ditches are provided
in FHWA (1998).
Steel face mesh can be installed over rock slope surfaces to restrain material that might
otherwise dislodge. The mesh may be relatively fine twisted wire or it may comprise
larger diameter elements depending on strength requirements. If the mesh is placed
directly on the slope face, it must be secured by anchors that are installed in a regular
pattern. In a separate application, the top of the mesh can be secured to a stable
section of rock slope and left suspended to drape in front of the slope at lower
elevations. Rocks that dislodge from the slope and would otherwise represent a rock
fall hazard are intercepted by the mesh curtain. In such applications, a ditch is usually
installed below the mesh to capture fallen material. As with reinforcement anchors,
steel mesh is subject to corrosion in aggressive electrochemical environments. Both
the mesh and its anchor components should be designed with corrosion protection.
Catch fences have become a common feature along highways that pass through
mountainous terrain. In general, they are a practical and cost-effective measure to
protect against rock falls. However, an assessment of rock fall trajectories is essential
to determine both the proper location, height, and structural capacities of such barriers.
7-10.8.8 Rock Sheds and Tunnels.
Rock sheds are an effective method of protecting vehicular traffic from rock falls, but
their use is rarely justified unless the cost of other mitigation measures is especially
high. They are typically needed on roadways that have been cut into hillsides. They
are designed with a roof that slopes downhill. On the uphill side of the roof, the roof
support beams are secured onto benches. On the downhill side of the roof, the roof
support beams can be supported by columns.
In locations where it is not practical to construct a rock shed it may be necessary to

construct a tunnel. FHWA (1998) describe such an example for a railway in Canada.

Topic Reference
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solution of stability problems, Proceedings of the ASCE Research
Conference on the Shear Strength of Cohesive Soils, Boulder, CO.
Stability Analysis - General Skempton, A. W. (1948). The Φ = 0 analysis of stability and its theoretical
basis, Proceedings of the 2nd International Conference on Soil Mechanics
and Foundation Engineering, Rotterdam, Vol. 1, pp. 72–78.
Terzaghi, K. Peck, R. B., and Mesri, G. (1996). Soil Mechanics in
Engineering Practice, 3rd ed., Wiley, Hoboken, NJ, 549 pages.
Bishop, A. W. (1955). The use of slip circles in the stability analysis of earth
Limit Equilibrium Methods
slopes, Geotechnique, 5(1), 7–17.
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Bromhead, E. N. (1992). The Stability of Slopes, 2nd ed., Blackie, New York.
Janbu, N. (1954a). Application of composite slip surface for stability analysis,
Proceedings of the European Conference on Stability of Earth Slopes,
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Janbu, N. (1954b). Stability Analysis of Slopes with Dimensionless
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Janbu, N. (1968). Slope stability computations, Soil Mechanics and
Foundation Engineering Report, The Technical University of Norway,
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Bjerrum, L. (1967). Progressive failure in slopes of overconsolidated plastic
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Division, 93(5), 1–49.
Chandler, R. J. (1977). Back analysis techniques for slope stabilization
works: a case record, Geotechnique, 27(4), 479–495.
Back Analysis Filz, G. M., Brandon, T. L., and Duncan, J. M. (1992). Back Analysis of
Olmsted Landslide Using Anistropic Strengths, Transportation Research
Record 1343, Transportation Research Board, National Research Council,
National Academy Press, Washington, DC, pp. 72–78.
Koerner, R. M. (1998). Designing with Geosynthetics, 4th ed., Prentice Hall,
Geosynthetic Reinforcement
Upper Saddle River, NJ.
Makdisi, F. I., and Seed, H. B. (1978). A simplified procedure for estimating
dam and embankment earthquake–induced deformations, ASCE, Journal of
the Geotechnical Engineering Division, 104(7), 849–867.
Seismic Slope Stability
Seed, H. B. (1979). Considerations in the earthquake–resistant design of
earth and rockfill dams, Nineteenth Rankine Lecture, Geotechnique, 29(3),
215–263.
Ladd, C. C. (1991). Stability evaluation during staged construction, ASCE,

Staged Construction of Slopes
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Residual and Fully-Softened Vol. 3, pp. 261–270.
Conditions
Skempton, A. W. (1985). Residual strength of clays in landslides, flooded
strata and the laboratory, Geotechnique, 35(1), 3–18.
7-12 NOTATION.
Symbol Description
bi Width of the slice
c Total stress cohesion
C Number of surfaces on which pullout resistance is mobilized
c' Effective stress cohesion
F Factor of safety
Fm Factor of safety for Bishop’s Simplified Method in effective stress example.
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Symbol Description
FN Component of W that acts perpendicular to the rock block sliding plane
FR Factor of safety for geosynthetic strength
FS Component of W that acts parallel to the rock block sliding plane
FW Factor of safety against wedge failure
F∗ Pullout resistance factor
Hi Average height of the slice
K Wedge factor
Le Length of reinforcement embedded behind the trial failure surface
MARV Minimum average roll value

Denominator in equation for calculating normal force at the base of a slice used for assessing
Mα validity of normal force.
Pr Geosynthetic reinforcement’s resistance to pullout
RFCR Reduction factor for creep
RFD Reduction factor for durability
RFID Reduction factor for installation damage
ru Pore pressure coefficient
s Shear strength
S Seepage force
SRF Strength reduction factor
ssu Undrained steady state shear strength
Tal Geosynthetic’s long-term tensile strength
TULT Ultimate tensile strength of the geosynthetic based on the MARV
U Uplift force applied by water at the failure plane
V Force applied by water in the tension crack
W Rock block weight
Wi Weight of each slice
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Symbol Description
α Dip direction or dip azimuth
α Scale correction factor to account for nonlinear stress reduction
αi Angle between the tangent to the failure surface and the horizontal
γ Unit weight
δ Effective soil-geosynthetic interface friction angle
θ Face angle
σ Total stress
σ' Effective stress
σ 'v Effective vertical stress
τ Shear stress required for equilibrium
τ' Effective stress friction angle
φ Tangent of the friction angle
φ' Effective friction angle
φ 'r Residual friction angle
Ψ Dip
Inclination of the slope face as measured at right angles to the line of intersection in a wedge
Ψ Fi failure analysis
Ψi Dip of the line of intersection in a wedge failure analysis
ΨP Angle of a sliding block plane with respect to horizontal
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CORRELATIONS FOR SOIL AND ROCK
8-1 INTRODUCTION.
Correlations are useful tools for obtaining values of engineering parameters based on
index properties or other easily measured soil parameters. Correlations are often used
when measured values are not available. The accuracy and applicability of correlations
depend on the data source, the statistical approach for determining the correlation, and
the causal relationship between the index property and the engineering parameter. For
these reasons, engineers should use correlations with caution. It is often prudent to
seek out the original source of the correlation to ensure that it is applicable to the
engineering problem at hand.
Often, correlations provide an uncertain, empirical prediction of the parameter, which

means that there are usually values above and below the proposed trend line.
Uncertainty results from both scatter in the measured data and the inability of the
chosen mathematical relationship to perfectly predict the observed trends. Because of
this uncertainty, correlations are typically most appropriate for preliminary design or as a
check that measured values are in general agreement with the behavior of the soils
used to develop the correlation. In addition, correlations can provide the basic form of
an equation that can be used with experimental data to create site-specific correlations
for an individual project or area.
For cases where the required property cannot be measured and correlations are used
for final designs, the uncertainty in the parameter should be evaluated by the engineer.
Different approaches can be used, including (1) the use of a range of values rather than
a single value of a parameter, (2) use of the lowest likely value, (3) application of
confidence limits, or (4) explicit consideration of uncertainty in the correlation using
formal reliability analysis.
Confidence limits are trend lines that are offset from the mean based on the standard
deviation ( S .D. ) of the residuals between the data and the mean. Confidence limit
boundaries plotted with mean trends help to illustrate the variability in a data set.
Confidence limits or the standard deviation can be used to select appropriately
conservative values from correlations. For example, if a correlated parameter is
assigned a value that is one standard deviation below the mean, the probability is only
16% that the actual value is lower than the assigned value. This probability reduces to
2% for an assigned value that is two standard deviations below the mean. These
probability margins assume that the error in the correlation follows a normal or log-
normal distribution and that the selected trend line fits the data well.
The selection of an appropriate confidence limit above or below the mean trend
depends on the effect of parameter on the particular analysis. In most cases, it is better
to use a confidence below the mean. This will decrease the probability that the
correlated parameter is greater than the actual value, which undesirable in many cases
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(e.g., shear strength, Young’s modulus, etc.). In a few cases, the opposite might be the
case (e.g., compression index).
8-2 EFFECTIVE STRESS (DRAINED) SHEAR STRENGTH.
8-2.1 Coarse-Grained Soils.
Most of the correlations presented for coarse-grained soils have been developed for
relatively clean sands unless otherwise noted. These correlations should not be used in
micaceous sands. The presence of mica tends to reduce some index properties (e.g.
the SPT N value) significantly but might not affect the drained friction angle when
compared to clean sands (Sabatini et al. 2002). These correlations should not be used
for gravelly soils unless specified.
8-2.1.1 Correlations with Soil Type.
Carter and Bentley (2016) summarized typical values for the effective stress friction
angles of coarse-grained soils as presented in Table 8-1 and Table 8-2. Table 8-1
presents values for the drained friction angle of different types of coarse-grained soils in
loose and dense conditions and Table 8-2 presents the values of effective stress friction
angle for coarse-grained soils compacted to the maximum dry density based on ASTM
D698.
Coarse-grained Soils (Carter and Bentley 2016)
φ ' (in degrees)
Soil Description
Loose Dense
Uniform sand, round grains 27 34
Well-graded sand, angular
33 45
grains
Sandy gravel 35 50
Silty sand 27-33 30-34
Inorganic silt 27-30 30-35
Compacted Coarse-grained Soils (Carter and Bentley 2016)
φ'
Soil Description USCS
(in degrees)
Well-graded sand-gravel mixtures GW >38
Poorly-graded sand-gravel mixtures GP >37
Silty gravels, poorly-graded gravel-sand-silt GM >34
Clayey gravels, poorly-graded gravel-sand-clay GC >31
Well-graded clean sand, gravelly sand SW 38
Poorly-graded clean sand, gravelly sand SP 37
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A correlation for the drained friction angle as a function of relative density, dry unit
weight and soil type is presented in Figure 8-1.
Figure 8-1 Approximate Relationship between the Effective Stress Friction

Angle and Dry Unit Weight for Various Relative Densities and Types of Soil
8-2.1.2 Correlations with Standard Penetration Test.
Many relationships have been presented in the literature to estimate drained shear
strength parameters of coarse-grained soils using results from the standard penetration
test. In older correlations where no energy correction for the SPT N value was used, it
was assumed that the reported N values were equal to N 60 .
Relationships between SPT N values and static cone tip resistance presented by
Duncan et al. (1989) summarizing the work presented by Meyerhof (1956) and Mitchell
(1981) are shown in Table 8-3 and Table 8-4, respectively.
Table 8-3 Relationship between SPT N Value, Relative Density and Effective
Stress Friction Angle (Meyerhof 1956)
Relative Static Cone Tip Effective Stress

State of
Density
N 60 Resistance, qc Friction Angle
Packing (blows/ft)
(%) (tsf) (degrees)
Very Loose < 20 <4 < 20 < 30

Loose 20 - 40 4 - 10 20 - 40 30 - 35
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Compact
40 - 60 10 - 30 40 - 120 35 - 40
(Medium)
Dense 60 - 80 30 - 50 120 - 200 40 - 45
Very Dense > 80 > 50 > 200 > 45
The effective stress friction angles presented in Table 8-3 are for clean sands and
should be decreased by 5° for clayey sands and increased by 5° for gravelly sands. To
use Table 8-3 on saturated very fine or silty sand, the measured SPT N should be
corrected using the equation below:
 N 60 for N 60 ≤ 15
N'= (8-1)
15 + 0.5 ( N 60 − 15 ) for N 60 > 15
where:
N ' = blow count corrected for dynamic pore pressure effects, and
N 60 = measured blow count corrected for 60% energy.
Table 8-4 Relationship between SPT N Value, Relative Density, and Angle of
Internal Resistance (after Mitchell 1981)
Relative Standard Penetration Static Cone Effective Stress Dry Unit

State of
Density1 Resistance, N 1,60 Resistance, qc Friction Angle Weight
Packing
(%) (blows/ft)2 (tsf) (degrees) (kN/m3)
V. Loose < 15 <4 < 50 < 30 < 14
Loose 15 - 35 4 - 10 50 - 100 30 - 32 14 - 16
M. Dense 35 - 65 10 - 30 100 - 150 32 - 35 16 - 18
Dense 65 - 85 30 - 50 150 - 200 35 - 38 18 - 20
V. Dense 85 - 100 > 50 > 200 > 38 > 20

1 Freshly deposited, normally consolidated sand
2 Corrected to an effective vertical overburden pressure of 1 atm.
Sowers (1979) related the effective stress friction angle to SPT N values for depths
less than 5 feet and greater than 20 feet as presented in Figure 8-2. Interpolation can
be used for depths between 5 and 20 feet. Shioi and Fukui (1982) also correlated the
effective stress friction angle to N 60 using (also shown in Figure 8-2):
=φ' 15 N 60 + 15 (For roads and bridges) (8-2)
=φ ' 0.3 N 60 + 27 (For buildings) (8-3)
where:
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N 60 = SPT N value corrected for 60% energy.
Figure 8-2 Relationship between Effective Stress Friction Angle of Coarse-

Grained Soils and SPT N 60 Value
Parry (1977) and Schmertmann (1975) considered the effects of overburden pressure
on the relationship between the effective stress friction angle and SPT N , resulting in
the correlations shown in Figure 8-3. Kulhawy and Mayne (1990) approximated the
trends in Figure 8-3(bottom) as:
0.34
 
 N 60 
φ ' = tan −1   (8-4)
12.2 + 20.3  σ 'v  
  Pa  
where:
φ ' = effective stress friction angle,
N60 = SPT N value corrected for 60% energy,
σ 'v = vertical effective stress, and
Pa = atmospheric pressure.
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Figure 8-3 Relationship between Peak Effective Stress Friction Angle,

Overburden Pressure, and SPT Blow Count for Sands (top) after Parry (1977) and
(bottom) after DeMello (1971) and Schmertmann (1975)
Peck et al. (1974) developed the correlation for the effective stress friction angle based
on SPT N values shown in Figure 8-4. According to Ameratunga et al. (2016), this
correlation is conservative. Wolff (1989) approximated this relationship as:
27.1+ 0.3 N1,60 − 0.00054 ( N1,60 )

2
φ'= (8-5)
where:
N1,60 = SPT N value corrected for 60% energy and 1 atm overburden pressure.
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Figure 8-4 Variation of Effective Stress Friction Angle with N 1,60

(after Peck at al. 1974, and Hatanaka and Uchida 1996)
Hatanaka and Uchida (1996) also correlated the effective stress friction angle of sands
to N1,60 as shown in Figure 8-4. This equation was developed using the results of
triaxial tests from high-quality intact frozen samples of natural sands. Using an SPT
hammer with an efficiency of 78%, the relationship was found to be:
=φ' 15.4 N1,60 + 20 (8-6)
8-2.1.3 Correlations with Cone Penetration Test.
Two relationships between CPT results and effective stress friction angle were already
presented in Table 8-3 and Table 8-4. Table 8-5 presents a similar relationship
developed by Bergdahl et al. (1993), according to Ameratunga et al. (2016).
Table 8-5 Relationship between Relative Density, Cone Tip Resistance, and
Effective Stress Friction Angle
(after Bergdahl et al. 1993, and Ameratunga et al. 2016)
Relative Density q (tsf) φ ' (degrees)

c
Very loose 0 – 26.1 29 – 32
Loose 26.1 – 52.2 32 – 35
Medium 52.2 – 104.4 35 – 37
Dense 104.4 – 208.8 37 – 40
Very dense > 208.8 40 – 42
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Mayne (2007) used the results of calibration chamber tests to estimate the effective stress
friction of coarse-grained soils for CPT results as
 qt 
 Pa 
φ '= 17.6 + 11⋅ log   (8-7)
 σ 'v 
 Pa 
 
where:
φ ' = drained friction angle,
qt =qc + u (1+ a) = corrected tip resistance,
u = pore pressure measured behind the cone tip, often named the u2 position,
a = cone net area ratio = ratio of the face area to shoulder area,
σ 'v = effective vertical stress, and
Pa = atmospheric pressure in same units as vertical stress and qc .
Figure 8-5 correlates the effective stress friction angle with CPT tip resistance as
summarized by Meyerhof (1976).
In cases where the tip resistance increases with depth, the method outlined in Figure
8-6 can be used to obtain the bearing capacity factor ( N q ). The correlation presented in
Figure 8-7 can be used to obtain the effective stress friction angle from N q .
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Figure 8-5 Relationship between Effective Stress Friction Angle and Cone Tip
Resistance (after Kerisel 1961, Kahl et al. 1968, Melzer 1968, Muhs and Weiss
1971, and Meyerhof 1976)
Figure 8-6 Estimation of φ ' from a Cone Resistance Profile

(after Duncan et al. 1989)
Figure 8-7 Relationship between Bearing Capacity Number N q and Peak

Effective Stress Friction Angle from Large Calibration Tests
(after Duncan et al. 1989)
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Robertson and Campanella (1983) correlated the effective stress friction angle to the
measured CPT tip resistance (electric cone) from tests performed in a calibration
chamber. The correlation used drained triaxial tests on uncemented, unaged,
moderately compressible quartz sands and considers the effect of overburden pressure
as shown in Figure 8-8. The relationship presented in Figure 8-8(a) was approximated
by Robertson and Cabal (2014) using the relationship:
1   qc  
=tan φ ' log   + 0.29  (8-8)
2.68   σ 'v  
where:
φ ' = drained friction angle,
qc = cone tip resistance, and
σ 'v = effective vertical stress in same units as qc .
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Figure 8-8 Variation of Peak Effective Stress Friction Angle with σ 'v and Cone
Resistance for Normally Consolidated, Uncemented, Quartz Sands
(after Robertson and Campanella 1983)
Schmertmann (1975) presented a correlation to determine the effective stress friction

angle based on the CPT tip resistance. To use this correlation, the relative density first
needs to be determined using Figure 8-9. Using the correlated value of relative density,
the effective stress friction angle can be determined from Figure 8-10.
For overconsolidated sands, the effect of overconsolidation on the measured tip

resistance needs to be considered before using correlations proposed for normally
consolidated sands. For this purpose, Schmertmann (1978) developed a correction
factor ( R ) such that:
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q
R =1 + 0.75 ( OCR β − 1) = c ,OC (8-9)
qc , NC
where:
OCR = overconsolidation ratio,
β = exponent,
qc ,OC = CPT tip resistance in overconsolidated sand, and
qc , NC = CPT tip resistance in normally consolidated sand.
An example of using this approach is shown in Figure 8-11. According to Lunne and
Christoffersen (1985), β can be assumed as 0.45 for all practical purposes.
Figure 8-9 Estimation of Relative Density for Normally Consolidated Sands from
Cone Penetration Resistance (after Schmertmann 1978)
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Figure 8-10 Relationship between Friction Angle and Relative Density based on
Triaxial Compression Tests on North Sea Sands
(after Schmertmann 1975, and Lunne and Kleven 1982)
Figure 8-11 Correction for Effects of Overconsolidation on Cone Penetration Tip

Resistance in Sand (after Lunne and Christoffersen 1985, and Duncan et al. 1989)
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8-2.1.4 Correlations with Dilatometer.
Marchetti (1997) proposed correlations to relate the effective stress friction angle for
clean sands to the horizontal stress index ( K D ) from the dilatometer test in which:
p0 − u0
KD = (8-10)
σ 'v
where:
p0 = corrected pressure required to initiate movement of the membrane against the soil,
u0 = hydrostatic pore pressure, and
σ 'v = effective vertical stress.
Ricceri et al. (2002) proposed that the upper bound estimate of effective stress friction
angle from dilatometer tests is:
KD
φ=' 31 + (8-11)
0.236 + 0.066 K D
and Marchetti (1997) proposed that the lower bound estimate is:
φ ' =28 + 14.6 ⋅ log K D − 2.1( log K D )

2
(8-12)
where:
φ ' = effective stress friction angle and
K D = horizontal stress index.
According to Marchetti, the lower bound solution can underestimate the in situ friction
angle by 2° to 4°. The upper and lower bound equations are plotted in Figure 8-12.
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Figure 8-12 Range of Effective Stress Friction Angle for Clean Sands based on
the Horizontal Stress Index from the Dilatometer Test
(after Marchetti 1997, and Ricceri et al. 2002)
8-2.2 Fine-Grained Soils.
The effects of overconsolidation on the shear strength preclude the development of

accurate correlations for the effective shear parameters of fine-grained soils. However,
multiple correlations have been developed for the fully softened ( φ 'FS ) and residual
( φ 'r ) friction angles of clays, where the fully softened friction angle is taken to be equal
to the normally consolidated peak value.
8-2.2.1 Correlations for Fully Softened Shear Strength.
Gibson (1953) presented a relationship for φ 'FS , which is plotted in Figure 8-13. The
standard deviation of the data about the trend is plotted as confidence limits. The mean
trend line can be approximated by the equation below according to Carter and Bentley
(2016):
φ=
'FS 0.0058 PI 1.73 − 0.32 PI + 36.2 (8-13)
where:
PI = plasticity index.
According to Carter and Bentley (2016) the polynomial fit is in agreement with the
original curve to within 1% for the fully softened condition.
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A similar correlation was presented by Ladd et al. (1977) based on triaxial tests on
intact normally consolidated clays as shown in Figure 8-14. The confidence limits in this
figure are plotted at one standard deviation above and below the mean trend.
Figure 8-13 Relationship between the Effective Stress Friction Angle of Fine-
Grained Soil and Plasticity Index (after Gibson 1953, Carter and Bentley 2016)
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Figure 8-14 Correlation between φ 'FS and PI based on Triaxial Tests on NC
Clays (after Kenney 1959, Bjerrum and Simons 1960, Ladd et al. 1977)
A similar relationship between φ 'FS and PI was proposed by Terzaghi et al. (1996) as
shown in Figure 8-15. This relationship was developed from the results of tests on
normally consolidated specimens with most of them being remolded.
Figure 8-15 Relationship between φ 'FS and PI (after Terzaghi et al. 1996)
Tiwari and Ajmera (2011) presented the correlations shown in Figure 8-16 and Figure
8-17 for the fully softened friction angle. These correlations were based on direct shear
tests performed on 36 artificially created soils. Tiwari and Ajmera (2011) created the
artificial soils for this study by mixing different proportions of quartz, kaolinite and
montmorillonite.
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Figure 8-16 Variation of the Fully Softened Friction Angle with Plasticity Index
(after Tiwari and Ajmera 2011)
Figure 8-17 Fully Softened Friction Angle based on Mineral Composition

(after Tiwari and Ajmera 2011)
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The preceding relationships only consider linear failure envelopes (i.e., constant friction
angle) while real soils often exhibit a nonlinear failure envelope. Nonlinear failure
envelopes have been described by multiple mathematical forms, including normal stress
dependent secant friction angle and two-parameter power function with parameters a
and b . Shear strength is calculated using a secant friction angle as:
s = σ ' ff tan φ 'sec (8-14)
where:
s = effective stress shear strength,
σ ' ff = effective normal stress on the failure plane, and
φ 'sec = stress dependent secant friction angle.
The two-parameter power function describes shear strength nonlinearly as:

b
σ ' 
s = aPa  ff  (8-15)
 Pa 
where:
s = effective stress shear strength,
a = empirical coefficient related to the steepness of the power function,
σ ' ff = effective normal stress,
Pa = atmospheric pressure in same units as stress, and
b = empirical coefficient related to the curvature of the power function.
Castellanos et al. (2021) presented the correlations for nonlinear fully softened shear
strength parameters shown in Figure 8-18 and Figure 8-19. These correlations were
developed based on over 400 direct shear tests on 97 soils (Castellanos et al. 2021).
The equations and standard deviations provided on the plots allow uncertainty in the
correlations to be explicitly considered. Castellanos et al. (2021) also provide statistical
measures of the covariance of the correlations for aFS and bFS .
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Plasticity Index (after Castellanos et al. 2021)
CF × PI (after Castellanos et al. 2021)
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8-2.2.2 Correlations for Residual Shear Strength.
Gibson (1953) also presented a relationship for φ 'r as plotted in Figure 8-13. The
residual friction angle trend line can be approximated by the equation below, according
to Carter and Bentley (2016):
φ=
'
r 0.084 PI 1.4 − 0.75 PI + 31.9 (8-16)
where:
According to Carter and Bentley (2016) this polynomial fit agrees with the original
curves to within 5% for the residual friction angle.
Skempton (1964, 1985) related the residual friction angle to the clay-sized fraction as
presented in Figure 8-20. The frictions angles were measured using ring shear tests
performed on normally consolidated and overconsolidated samples.
Using published results, Voight (1973) developed the correlation presented in Figure
8-21(top) to estimate the residual strength of clays based on the plasticity index.
Voight’s correlation was later supported by the residual strength measurements
performed by Bovis (1985) whose results can be seen in Figure 8-21(bottom).
Stark and Hussain (2013) correlated the residual friction angle of clays to the liquid limit
for various ranges of clay sized fraction as shown in Figure 8-22. This correlation needs
to be used with care because of the methods that were used to process the soil
samples for index property measurements were not consistent. For this correlation, the
index properties of clay samples were obtained from specimens sieved through a No.
40 sieve without any other processing. On the other hand, shale samples were ball-
milled and sieved through a No. 200 sieve. These differences in the procedures used to
process samples for measuring the index properties should be considered when using
the correlation.
Stark and Hussain (2013) correlation provides stress-dependent residual secant friction
angles for four different value of effective normal stress ranging from 50 to 700 kPa.
The trends can be calculated as
φr' =
C0 + C1 LL + C2 LL2 + C3 LL3 (8-17)
where:
LL = liquid limit and
C0 , C1 , C2 , and C3 = empirical coefficients listed in Table 8-6.
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Figure 8-20 Correlation between the Residual Friction Angle and Clay-sized
Fraction (after Skempton 1964, 1985)
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Figure 8-21 Residual Friction Angle vs. Plasticity Index – (top) Data Collected by
Voight (1973), and (bottom) Measurements by Bovis (1985)
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Figure 8-22 Drained Residual Secant Friction Angle as a Function of LL and CF

(after Stark and Hussain 2013)
Table 8-6 Coefficients for Stark and Hussain (2013) Residual Friction Angle
Correlation
Effective Coefficients for Equation 8-17
CF and LL Range normal stress
C0 C1 C2 C3
σ ' (kPa)
50 39.7 -0.29 6.63E-04 0
100 39.4 -0.298 6.81E-04 0
CF < 20%
400 40.2 -0.375 1.36E-03 0
700 40.3 -0.412 1.68E-03 0
50 31.4 -6.79E-03 -3.62E-03 1.86E-05
100 29.8 -3.63E-04 -3.58E-03 1.85E-05
25% < CF < 45%
400 28.4 -5.62E-02 -2.95E-03 1.72E-05
700 28.1 -0.2083 -8.18E-04 9.37E-06
50 33.5 -0.31 3.90E-04 4.40E-06
CF > 50% and 100 30.7 -0.2504 -4.21E-04 8.05E-06
LL < 120 400 29.4 -0.2621 -4.01E-04 8.72E-06
700 27.7 -0.3233 2.90E-04 7.11E-06
50 12.0 -0.0215 0 0
CF > 50% and 100 10.9 -0.0183 0 0
120 < LL < 300 400 8.3 -0.0114 0 0
700 5.8 -0.0049 0 0
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Laboratory testing at Virginia Tech has produced correlations for residual shear strength
based on the results of torsional ring shear tests on 102 clays with plasticity indices
between 6 and 112, liquid limits between 22 and 143, and clay fractions ranging from 13
to 90%. Figure 8-23 and Figure 8-24 present the relationship between power function
parameters (see Equation 8-15), plasticity index, and clay fraction.

Plasticity Index (after Castellanos et al. 2021)

Plasticity Index and Clay Fraction (after Castellanos et al. 2021)
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8-3 UNDRAINED SHEAR STRENGTH.
8-3.1 Correlations with Index Properties.
Skempton and Northey (1952) presented the relationship shown in Figure 8-25 that
related the undrained shear strength of normally consolidated clays to the liquidity
index. Terzaghi et al. (1996) demonstrated a strong relationship between the undrained
shear strength of remolded clays and the liquidity index as shown in Figure 8-26. The
results show similar behavior over a wider range of liquidity index to that observed by
Skempton and Northey (1952).
Figure 8-25 Relation between Liquidity Index and Undrained Shear Strength of
Remolded Clays (after Skempton and Northey 1952)
Skempton (1957) compiled data from various sources to estimate the undrained shear
strength of normally consolidated clay based on plasticity index as:
su
= 0.11 + 0.0037 PI (8-18)
σ 'v
where:
su = undrained shear strength,
PI = plasticity index (%).
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Figure 8-26 Relationship between Remolded Undrained Shear Strength and

Liquidity Index (after Terzaghi et al. 1996)
Skempton's (1957) correlation, which is plotted in Figure 8-27, was based primarily on
the results of field vane tests. It is unlikely that these shear strengths were corrected as
required by ASTM D2573, meaning the values might be too high. The correlation
presented by Skempton (1957) was later supported by Robertson and Campanella
(1984) using vane shear test results presented by Ladd and Foott (1974).
A similar correlation based on laboratory tests was presented by Ladd and DeGroot
(2004) and is shown in Figure 8-28. The correlation presented Ladd and DeGroot
(2004) shows the dependency of the undrained shear strength on the laboratory stress
path used to obtain the measurement.
Larson (1980) collected the undrained strength ratios and liquid limits of normally
consolidated Scandinavian clays from various sources. These data is plotted in Figure
8-29 along with the equation proposed by Hansbo (1957):
 su 
  = 0.0045 LL (8-19)
 σ 'v  NC
where:
su = undrained shear strength (normally consolidated),
LL = liquid limit (%).
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Field Vane (after Robertson and Campanella 1984)
Laboratory Testing (after Ladd and DeGroot 2004)
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The data collected by Larson (1980), which is from field vane tests, agrees well with the
Equation 8-19. Larson (1980) discusses various methods for correcting undrained
strength measured using the vane shear but it is not clear whether or not the data in
Figure 8-29 were corrected.
Figure 8-29 Variation of the Undrained Strength Ratio with Liquid Limit
(after Larson 1980)
8-3.2 Correlations with Stress History.
Mesri (1975) corrected the results of vane shear tests from the literature with the vane
correction factor proposed by Bjerrum (1972) and found that the undrained strength
ratio for normally consolidated soil is relatively constant. Similarly, Jamiolkowski et al.
(1985) found relatively constant values of normally consolidated undrained strength
ratio for clays with PI less than 60. Chandler (1988) and Ladd and DeGroot (2004)
also analyzed undrained strength data sets to determine typical values of undrained
strength ratio for various types of clay. These trends are summarized in Table 8-7.
Overconsolidation results in an increase in the undrained shear strength. Schmertmann

(1978) compared the undrained strength ratio for overconsolidated clays to that for
normally consolidated clays as a function of OCR as summarized in Table 8-8 and
Figure 8-30.
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Table 8-7 Typical Normally Consolidated Undrained Strength Ratios
Source NC Undrained Strength Ratio Comments
Mesri (1975) 0.22 ± 0.03 Collection of vane shear test results
Jamiolkowski et al. Clays with PI less than 60 based on several

0.23 ± 0.04
(1985) embankment failures
Range: 0.16 to 0.33
High value: 0.74
Based on the Mesri (1975) dataset of field vane shear
Chandler (1988) Mean (all): 0.28
tests on clay
Mean (discard extreme values):
0.22 ± 0.05
0.23 Saturated clay
Sabatini et al. (2002)
0.16 Soils with horizontal layering or features
Sensitive cemented marine clays, Canadian Champlain
Nominally 0.20
clays
Std. Dev. = 0.015
( PI < 30%, LI > 1.5)
 PI  Homogeneous CL and CH sedimentary clays of low to
0.22+ 0.05 
 100  moderate sensitivity, no shells or sand lenses layers
Ladd and DeGroot
Nominally 0.22 ( PI = 20 to 80%)
(2004)
Northeastern U.S. varved clays, direct simple shear
0.16
failure mode
Sedimentary deposits of silts and organic soils
Nominally 0.25
(Atterberg Limits plot below A-line) and clays with
Std. Dev. = 0.05
shells, excludes peat
Notes: PI = plasticity index, LI = liquid index
Shear strengths presented by Jamiolkowski et al. (1985) and Chandler (1988) were not corrected as
required by ASTM D2573 and may be too high.
Table 8-8 Approximate Relation of Undrained Strength Ratio and OCR

(after Schmertmann 1978)
OCR Undrained Strength Ratio
Less than 1 0 – 0.1
1 0.10 – 0.25
1 to 1.5 0.26 – 0.50
3 0.51 – 1.00
6 1.00 – 4.00
Greater than 6 > 4.00
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Figure 8-30 Normalized Undrained Strength Ratio vs. OCR

(after Schmertmann 1978)
For overconsolidated clays, Jamiolkowski et al. (1985) and Ladd and DeGroot (2004)
showed that the effects of stress history on the undrained shear strength ratio can be
accounted for by:
 su   su 
  =  OCR
m
(8-20)
 σ 'v OC  σ 'v  NC
where:
su = undrained shear strength ( NC = normally consolidated, OC = overconsolidated),
σ 'v = vertical effective stress,
OCR = overconsolidation ratio, and
m = semi-empirical fitting parameter.
The value of m is theoretically related to the recompression and compression indices

as shown by Roscoe et al. (1958) and Mitachi and Kitago (1976) for ideal or remolded
soils. 14 For real soils, undrained laboratory shear strength tests on soil specimens at
different values of OCR can be used to determine m . Ladd et al. (1977) observed that
14For this to be true, ideal or remolded soils must be assumed to follow the tenets of critical state soil
mechanics. This assumption breaks down for soils exhibiting post-peak strain softening. This idealization
also assumes that Cc and Cr are log-linear for all ranges of stresses and for any rebound pressure.
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m is approximately 0.8 based on direct simple shear tests. Typical values for m are
summarized in Table 8-9.
Table 8-9 Typical Values of m
Source m Soil Description

Roscoe et al. (1958), Cr
Mitachi and Kitago m ≈ 1− Saturated clay (theoretical)
Cc
(1976)
Range: 0.8 to 1.35
Jamiolkowski et al. High value: 1.51
(1985), Chandler Mean (all): 1.03 Field vane shear tests on clay
(1988) Mean (discarding
extreme values): 0.97
Sabatini et al. (2002) 0.8 Saturated clay
Sensitive cemented marine clays,
1.00 Canadian Champlain clays
( PI < 30%, LI > 1.5 )
 Cr  Homogeneous CL and CH sedimentary

0.88 ⋅  1 −  ± 0.06 clays of low to moderate sensitivity, no
 Cc  shells or sand lenses layers
Ladd and DeGroot Nominally 0.8 ( PI = 20 to 80%)
(2004)
Northeastern U.S. varved clays, direct
0.75
simple shear failure mode
 Cr  Sedimentary deposits of silts and organic

0.88 ⋅  1 −  ± 0.06 soils with Atterberg Limits below the A-line
 Cc 
and clays with shells, excludes peat
Nominally 0.8
Note: Cc = compression index and Cr = recompression index
For very soft clays with overconsolidation ratios less than 2, Sabatini et al. (2002) found
that the undrained shear strength could be estimated as (assumes m equals 1):
su ≈ 0.21σ ' p (8-21)
where:
su = undrained shear strength, and
σ ' p = preconsolidation pressure.
The consolidation stress state in triaxial compression tests also influences the
undrained shear strength. Based on data from 48 normally consolidated clays, Mayne
(1985) found that the ratio was about 0.87 for undrained shear strengths from K 0
consolidated and isotropically consolidated undrained triaxial compression tests.
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Kulhawy and Mayne (1990) examined the data and found that the normally consolidated
undrained strength ratio for K 0 -consolidated tests can be related to the isotopically
consolidated tests by:
 su   su 
  = 0.15 + 0.49   (8-22)
 σ 'v  ACU  σ 'v  ICU
where:
( su σ 'v ) ACU = undrained strength ratio in CK0U triaxial compression, and
( su σ 'v ) ICU = undrained strength ratio in ICU triaxial compression.
8-3.3 Correlations with Cone Penetration Test.
Undrained shear strength is typically estimated from the cone tip resistance measured
in the CPT using methods based on bearing capacity. The three methods are the N c ,
N k , and N kt methods as defined by Lunne et al. (1997). The empirical bearing capacity
factors ( N c , N k , and N kt ) should be calibrated on a site- or region-specific basis by
relating known values of undrained shear strength measured using the triaxial device
(ASTM D2166, ASTM D2850), laboratory miniature vane shear (ASTM D4648), field
vane shear (ASTM D2573), or direct simple shear (ASTM D6528) to the predicted
values based on Equations 8-23 to 8-25.
The N c method is the simplest and directly relates the CPT tip resistance to the
undrained shear strength as:
qc
su = (8-23)
Nc
where:
N c = empirical bearing capacity factor.
In most cases, the value of N c is in the range of 17 to 23 for normally consolidated and
slightly overconsolidated clays. The N c method may be less accurate than other
methods at depths greater than 15 m because the overburden pressure is not
considered.
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The N k method considers the overburden pressure acting at the point of the
measurement. Based on this method, the undrained shear strength can be determined
from CPT results as:
qc − σ v
su = (8-24)
Nk
where:
qc = cone tip resistance,
σ v = total vertical stress, and
N k = empirical bearing capacity factor.
Data presented by Lunne and Kleven (1982), shows that N k ranges from about 10 to
about 19 with an average of 15. Carter and Bentley (2016) suggest values of 17 or 18
for normally consolidated clays and 20 for overconsolidated clays.
The N kt method is a modification of the N k method that considers the pore pressure
acting at the tip of the cone. The undrained shear strength is calculated as:
qt − σ v
su = (8-25)
N kt
where:
qc = cone tip resistance.
σ v = total vertical stress.
N kt = empirical bearing capacity factor.
qt =qc + u (1+ a) = corrected tip resistance.
u = pore pressure measured behind the cone tip, often called the u2 position.
a = cone net area ratio = ratio of the face area to shoulder area.
Modern cones have net area ratios above 0.8 and little difference is typically observed
in the N k and N kt methods. Values of N kt are often in the range of 14 to 16.
8-3.4 Correlations with Standard Penetration Test.
Various attempts have been made to correlate the undrained shear strength of clays to
SPT N values. Observed ranges of undrained shear strength based on soil
consistency and N are summarized in Table 8-10.
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Table 8-10 Approximate Undrained Shear Strength for Cohesive Soils Based on
SPT N
Undrained Shear Strength (psf)
Soil SPT N
Consistency Value Parcher and Means Tschebotarioff Terzaghi et al.
(1968) (1973) (1996)
Very soft <2 300 - < 250
Soft 2–4 300 – 600 250 – 500 250 – 500
Medium 4–8 600 – 1200 500 – 1000 500 – 1000
Stiff 8 – 15 1200 – 2400 1000 – 2000 1000 – 2000
Very stiff 15 – 30 2400 2000 – 4000 2000 – 4000
Hard > 30 > 4500 > 4000 > 4000
Stroud and Butler (1975) developed a correlation for the undrained shear strength of
overconsolidated clays as a function of the SPT N value. As shown in Figure 8-31, the
relationship exhibits significant scatter, which reduces the reliability of the correlation.
Carter and Bentley (2016) approximated the trendline in Figure 8-31 as:
su 8910
= + 4.36 (8-26)
N PI 3
where:
su = undrained shear strength (in kPa),
N = SPT-N value, and
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Figure 8-31 Correlation between Undrained Shear Strength, SPT N Value, and
Plasticity Index for Overconsolidated Clays (after Stroud and Butler 1975)
Sowers (1979) correlated undrained shear strength to SPT N for different USCS soil
classifications as presented in Figure 8-32. Relationships proposed by Hara et al.
(1974) and Terzaghi and Peck (1967) are also presented in Figure 8-32. Hara et al.’s
correlation is based on undrained shear strengths from triaxial compression tests.
Figure 8-32 Relationship between Undrained Shear Strength and SPT N

(after Terzaghi and Peck 1967, Hara et al. 1974, and Sowers 1979)
8-3.5 Correlations with Dilatometer.
The undrained shear strength of overconsolidated clays has been correlated to the
horizontal stress index ( K D ) and the dilatometer modulus ( ED ) as summarized in Table
8-11.
Table 8-11 Undrained Shear Strength Correlations to Dilatometer

Undrained Strength Ratio or Undrained Shear Strength (kPa) Source
 su   su  1.25 1.25
= σ '   σ '  ( 0.5 K D ) ≈ 0.22 ( 0.5 K D ) Marchetti (1980)
 v OC  v  NC
 su  1.14
 σ '  = 0.35 ( 0.47 K D ) Kamei and Iwasaki (1995)
 v OC
su = 0.018 E D Iwasaki and Kamei (1994)
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Note:
su = undrained shear strength in kPa, σ 'v = vertical effective stress in kPa,
K D = horizontal stress index from dilatometer, and E D = dilatometer modulus in kPa.
8-4 CONSOLIDATION PARAMETERS.
8-4.1 Compression and Recompression Indices – Fine-Grained.
The compression index ( Cc ) is the slope of the virgin consolidation line of the e vs.
log( σ 'v ) plot. The recompression index ( Cr ) is the slope of the recompression line of
the e vs. log( σ 'v ) plot. These parameters are used to calculate the compression of the
clay when subjected to an increase in stress in the normally consolidated and
overconsolidated ranges (See Section 5-5.2.1). An alternative of the compression and
recompression indices are the modified compression index ( Cε c ) and modified
recompression index, Cε r (a.k.a., compression and recompression ratio). The modified
compression and recompression indices are equal to the compression and
recompression indices divided by ( 1 + e0 ), respectively. These are the slopes of the
compression and recompression curves when vertical strain is used instead of void
ratio.
8-4.1.1 Typical Values.
Typical values of the compression index for different clays and silts are summarized in
Table 8-12.
Table 8-12 Typical Values for C c for Undisturbed Clays
Soil Cc Reference
Boston Blue Clay, undisturbed (CL) 0.35
Chicago clay undisturbed (CH) 0.42
Cincinnati Clay (CL) 0.17
Louisiana Clay, undisturbed 0.33 Lambe and
New Orleans Clay, undisturbed (CH) 0.29 Whitman (1969)
Siburua clay (CH) 0.21
Kaolinite 0.21 – 0.26
Na-Montmorillonite (CH) 2.6
Normally consolidated medium sensitive clays 0.2 – 0.5
Organic silt and clayey silts (ML-MH) 1.5 – 4.0
Organic clays (OH) > 4.0
Peat (Pt) 10 – 15
Chicago silty clay (CL) 0.15 – 0.30
Holtz and Kovacs
Boston Blue Clay (CL) 0.3 – 0.5
(1981)
Vicksburg Buckshot Clay (CH) 0.5 – 0.6
Swedish medium sensitive clays (CL-CH) 1–3
Canadian Leda clays (CL-CH) 1–4
Mexico City Clay 7 – 10
San Francisco Bay Mud (CL) 0.4 – 1.2
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Bangkok Clays (CH) 0.4
Uniform sand, loose (SP) 0.05 – 0.06
Uniform sand, dense (SP) 0.02 – 0.03 USACE (1990)
Uniform silts (ML) 0.2
8-4.1.2 Correlations with Index Properties.
Many relationships have been developed to estimate the compression and

recompression indices based on parameters, such as water content, liquid limit, and
void ratio. Some of these correlations are summarized in Table 8-13 and Table 8-14
and plotted in Figure 8-33 thru Figure 8-38. As can be seen from these plots, the
presented correlations estimate values of compression and recompression indices that
vary significantly. For the compression index, the correlations using the natural water
content tend to be in closer agreement compared to those based on other index
properties. Prior to use, the soil type(s) used to develop the correlations and the
sensitivity of the project to errors in the prediction of settlement should be considered to
determine if the intended application matches.
Leroueil et al. (1983) showed that the sensitivity of the clay also affects the value of the
compression index, especially for marine deposits. The results presented in Figure
8-39 show a significant effect of the sensitivity on the compression index.
Lambe and Whitman (1969) presented typical ranges of the modified compression
index of clays as a function of the natural water content and these are shown in Figure
8-40.
Table 8-13 Compression Index Correlations

Correlation Comments References
=Cc 0.007( LL − 10) Remolded clays. Skempton (1944)
=Cc 0.0046( LL − 9) Clays from Sao Paulo, Brazil Cozzolino (1961)

1.673
LL
Cc = Hong Kong soft marine clay Lumb and Holt (1968)
2040
=Cc 0.0083( LL − 9) Remolded clays Schofield and Wroth (1968)
Cohesive soils of the Rhone
=Cc 0.003( LL − 10) Alpes and Valley of the Gielly et al. (1969)
Seine River
=Cc 0.006( LL − 9) Clays for Greece and USA Azzouz et al. (1976)
=Cc 0.008( LL − 5) Dredging material Salem and Krizek (1976)
=Cc 0.00797( LL − 8.16) Indiana soils Lo and Lovell (1982)
=Cc 0.01( LL − 13) All clays USACE (1990)

Undisturbed clay of
=Cc 0.009( LL − 10) sensitivity less than 4. Terzaghi et al. (1996)
Reliability 30%
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=Cc 1.15(e0 − 0.91) All clays (Lower limit) Nishida (1956)
=Cc 0.30(e0 − 0.27) Inorganic silty clays Hough (1957)
Cc =0.256 + 0.43(e0 − 0.84) Brazilian motley clays

Cozzolino (1961)
1.21 + 1.055(e0 − 1.87)
Cc = Brazilian soft silty clays
Table 8 13 (cont.) Compression Index Correlations
=Cc 0.75(e0 − 0.50) Soils of very low plasticity Sowers (1970)
=Cc 0.40(e0 − 0.25) Clays for Greece and USA Azzouz et al. (1976)
Weathered and soft
Cc 0.22 + 0.29e0
= Adikari (1977)
Bangkok clays
=Cc 0.575e0 − 0.241 French clays Vidalie (1977)
=Cc 0.5363(e0 − 0.411) Indiana soils
=Cc 0.5673(e0 − 0.4422) Wabash Lowland

Goldberg et al. (1979)
=Cc 0.4941(e0 − 0.3507) Crawford Upland
Outwash and alluvial
=Cc 0.5621(e0 − 0.4215)
deposits
=Cc 0.496e0 − 0.195 Indiana soils Lo and Lovell (1982)
Cc = 0.3745e0 Saturated clays

Soils from nine states in the Rendon-Herrero (1983)
=Cc 0.434(e0 − 0.336)
USA
3/2
 w 
Cc = 0.85  n  Finnish muds and clays Helenelund (1951)
 100 
=Cc 0.01404 wn − 0.189 All clays Nishida (1956)
Cc = 0.01wn Chicago and Canada clays Koppula (1981)
=Cc 0.01( wn − 5) Clays for Greece and USA Azzouz et al. (1976)
Weathered and soft
=Cc 0.008wn + 0.2 Adikari (1977)
Bangkok clays
=Cc 0.0147 wn − 0.213 French clays Vidalie (1977)
=Cc 0.0133wn − 0.1621 Crawford Upland Goldberg et al. (1979)
=Cc 0.0126 wn − 0.162 Indiana soils Lo and Lovell (1982)

Soils from nine states in the
=Cc 0.01wn − 0.07549 Rendon-Herrero (1983)
USA
Cc = 0.0115wn Organic soils, peats
USACE (1990)
Cc = 0.012 wn All Clays
Cc = 0.135 PI
Remolded clays Wroth and Wood (1978)
=Cc 0.005 PI ⋅ Gs
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Cc =0.37(e0 + 0.003LL − 0.34) Clays for Greece and USA Azzouz et al. (1976)
Weathered and soft
Cc =0.009e0 + 0.008 LL + 0.20 Adikari (1977)
Bangkok clays
Cc= 0.0101(e0 LL − 0.5765LL + 12.665) Crawford Upland Goldberg et al. (1979)
Cc =0.40(e0 + 0.001wn − 0.25)

Clays for Greece and USA Azzouz et al. (1976)
Cc = 0.009 wn + 0.002 LL − 0.1
Table 8 13 (cont.) Compression Index Correlations
Cc = 0.0129( wn + 0.1015 LL − 16.1875) Indiana soils Goldberg et al. (1979)
Cc = 0.0114( wn + 0.2491LL − 18.8134) Crawford Upland Goldberg et al. (1979)

Cohesive soils in Alberta,
Cc =0.0082 wn + 0.0043CF − 0.1403 Koppula (1981)
Canada
Cc =0.37(e0 + 0.003LL + 0.0004wn − 0.34) Clays for Greece and USA Azzouz et al. (1976)
Cc = 0.0153( wn + 0.1022 LL −
0.3104 PL − 11.623)
Indiana soils
Cc = 0.5684(e0 + 0.033LL − 0.0082 PL
+ 0.0329σ ' p − 0.4322)
Cc = 0.6076(e0 + 0.003LL − 0.0095 PL Outwash and alluvial
+ 0.43σ ' p − 0.4186) deposits
Cc = 0.0025CF + 0.1165e0 + 0.0036 wn Cohesive soils in Alberta,

Koppula (1981)
+ 0.0014 PI + 0.0009 PL − 0.997 Canada
2.4
 1 + e0  Al‐Khafaji and
Cc = 0.5   Saturated clay
Andersland (1992)
 Gs 
Saturated sediment fine-
Cc = 0.0121wnGs
grained soil
 (1 + e0 )2 
=Cc 0.185  − 0.144 
 Gs 
  (1 + e )2   Soils from nine states in
Rendon-Herrero (1983)
0
=Cc 0.489 ln   + 0.296 
  Gs   USA
2.382
 1 + e0 
1.2
Cc = 0.141G  s 
 Gs 
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Table 8-14 Recompression Index Correlations

Marine clays of Southeast
Cr = 0.0045 LL Asia
Cox (1968)
=Cr 0.002( LL + 9) Clays for Greece and USA Azzouz et al. (1976)
Balasubramaniam and
=Cr 0.00463LL − 0.013 Bangkok clays
Brenner (1981)
=Cr 0.00238 LL + 0.0294 Indiana soils Lo and Lovell (1982)
=Cr 0.208e0 + 0.0083 Chicago clays Peck and Reed (1954)

Inorganic and organic clayey Elnaggar and Krizek
=Cr 0.156e0 + 0.0107 and silty soil (1970)
=Cr 0.14(e0 + 0.007) Clays for Greece and USA Azzouz et al. (1976)
=Cr 0.2037(e0 − 0.2465) Indiana soils

=Cr 0.221(e0 − 0.3074) Wabash Lowland
=Cr 0.152e0 + 0.0125 Indiana soils Lo and Lovell (1982)

Marine clays of Southeast
Cr = 0.0043wn Asia
Cox (1968)
=Cr 0.003( wn + 7) Clays for Greece and USA Azzouz et al. (1976)
0.0039 wn + 0.013 for wn < 100%

Cr =
French clays Vidalie (1977)
=Cr 0.403log ( wn ) − 0.478
=Cr 0.0065( wn − 11.6361) Wabash Lowland Goldberg et al. (1979)
Balasubramaniam and
=Cr 0.00566 wn − 0.037 Bangkok clays
Brenner (1981)
=Cr 0.003wn + 0.0249 Indiana soils Lo and Lovell (1982)
Cr = PI 370 Remolded clays Wroth and Wood (1978)
Cr = 0.126(e0 + 0.003LL − 0.06)

Cr =0.142(e0 − 0.0009 wn + 0.006)
=Cr 0.0034(e0 wn + 8.3647) Wabash Lowland
=Cr 0.0033(e0 wn + 12.5168) ) Crawford Upland
Cr =0.003wn + 0.0006 LL + 0.004

Cr = 0.135(e0 + 0.01LL − 0.002wn − 0.06)
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Figure 8-33 Range of Compression Index based on Liquid Limit Predicted by

Correlations
Figure 8-34 Range of Compression Index based on Initial Void Ratio Predicted by
Correlations
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Figure 8-35 Range of Compression Index based on Natural Water Content

Predicted by Correlations
Figure 8-36 Correlations for Recompression Index based on Liquid Limit
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Figure 8-37 Range of Recompression Index based on Initial Void Ratio Predicted
by Correlations
Figure 8-38 Range of Recompression Index based on Natural Water Content

Predicted by Correlations
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Figure 8-39 Sensitivity ( St ), In situ Void Ratio, and Compression Index

Relationship (after Leroueil et al. 1983)
Figure 8-40 Correlation between Modified Compression Index and Water Content
(after Lambe and Whitman 1969)
Burland (1990) compiled data from normally consolidated clays, both undisturbed and
reconstituted at a liquidity index ranging from 1 to 1.5. Recognizing that vertical
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stresses between 100 and 1000 kPa control most consolidation calculations, he defined
the normally consolidated void ratios at these stresses as e∗100 and e∗1000 , respectively.
From these void ratios, the intrinsic compression index was defined as:
C
=c
* *
e100 *
− e1000 (8-27)
where:
C ∗c = intrinsic compression index,
e∗100 = intrinsic void ratio at 100 kPa, and
e∗1000 = intrinsic void ratio at 1000 kPa.
By normalizing the current void ratio with respect to e∗100 , Burland defined the void index
as:
*
e − e100
Iv = (8-28)
Cc*
where:
I v = void index,
e = void ratio,
e∗100 = intrinsic void ratio at 100 kPa, and
C ∗c = intrinsic compression index.
With the data normalized in this way, he defined the sedimentation compression line
(SCL) and intrinsic compression line (ICL), which describe the typical variation of the in
situ void index ( I v 0 ) with effective stress for a wide range of clays. The SCL represents
the typical relationship for the compression of naturally sedimented clays. The ICL
represents the typical relationship for the compression of remolded clays. Burland’s
SCL and ICL are plotted in Figure 8-41.
While the values of e∗100 and C ∗c are best determined from laboratory tests, Burland
(1990) found that these parameters could be estimated from the void ratio at the liquid
limit ( eL ) by:
*
e100 =0.109 + 0.679eL − 0.089eL2 + 0.016eL3 (8-29)
and
=Cc* 0.256eL − 0.04 (8-30)
where:
e∗100 = intrinsic void ratio at 100 kPa,
eL = void ratio at water content equal to the liquid limit, and
C ∗c = intrinsic compression index.
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Figure 8-41 Sedimentation and Intrinsic Compression Lines (after Burland 1990)
Using Equations 8-28 through 8-30 and the relationships shown in Figure 8-41, the
virgin consolidation line can be approximated solely based on liquid limit for both freshly
deposited soils (using ICL) or structured, aged deposits (using SCL). This approach is
particularly useful for validation of laboratory consolidation tests. Examples of the ICL
and SCL for liquid limits ranging from 25 to 125 are plotted in Figure 8-42.
Figure 8-42 Example NC Compression Curves based on a) Intrinsic

Compression Line and b) Secondary Compression Line
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8-4.2 Compression and Recompression Indices – Coarse-Grained.
The compressibility of coarse-grained soils is normally significantly smaller than that

from fine-grained soils. The compressibility of coarse-grained soils is not typically
defined in terms of the compression and recompression indices. Instead the modulus-
based methods presented in Chapter 5 are employed. The constrained modulus for
coarse-grained soils is usually stress dependent (Kulhawy and Mayne 1990) and may
be estimated using correlations in the following section. If required, typical values for the
modified compression index of coarse-grained soils are summarized in Table 8-15 and
Table 8-16.
Table 8-15 Modified Compression Indices for Saturated, Normally Consolidated

Sands (after Burmister 1962, Coduto et al. 2011)
Cε c Cc (1+ e0 )
=
Soil Type
Dr = 0% Dr = 20% Dr = 40% Dr = 60% Dr = 80% Dr = 100%
Medium to coarse
0.010 0.008 0.006 0.005 0.003 0.002
sand (SW & SP)
Fine to coarse sand

0.011 0.009 0.007 0.005 0.003 0.002
(SW)
Fine to medium sand

0.013 0.010 0.008 0.006 0.004 0.003
(SW & SP)
Fine sand (SP) 0.015 0.013 0.010 0.008 0.005 0.003
Fine sand with little

0.017 0.014 0.012 0.009 0.006 0.003
fine to coarse silt (SM)
Table 8-16 Compressibility Data for Six Sands (Been et al. 1987)
Cc
Sand e0 Cr
σ 'v Pa = 1 to 3 σ 'v Pa = 20 to 30
0.854 0.021 0.085 0.006
Monterrey 0
0.782 0.018 0.090 0.007
0.917 0.025 0.130 0.007
Ticino
0.827 0.026 0.085 0.006
0.870 0.024 0.095 0.005
Hokksund
0.790 0.018 0.056 0.005
0.760 0.025 0.030 0.007
Ottawa
0.560 0.005 0.100 0.003
Reid- 0.900 0.013 0.090 0.005
Bedford 0.650 0.005 0.019 0.003
Hilton Mines 0.950 0.038 0.210 0.009
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0.732 0.022 0.100 0.006
8-4.3 Constrained Modulus.
The secant drained constrained modulus ( M ds ) was found to be a function of the vertical
effective stress ( σ 'v ) and a modulus number ( m ) by Janbu (1963). The constrained
modulus for normally consolidated clays, silts, and sands is related in either a linear or
nonlinear fashion to vertical effective stress by:
M ds = mσ 'v (8-31)
or
0.5
σ ' 
M ds = mPa  v  (8-32)
 Pa 
where:
M ds = constrained modulus,
m = modulus number,
Pa = atmospheric pressure (same units as M ds and σ 'v ).
Janbu (1963) related the modulus number to the void ratio (or porosity) and the natural
water content as shown in Figure 8-43 and Figure 8-44, respectively. In addition, Janbu
(1985) presented the relationship for the modulus number of NC silts and sands as a
function of the porosity, as presented in Figure 8-45.
8-4.3.1 Correlations with Standard Penetration Test.
Based on results from nine British clays, Stroud (1974) correlated the constrained
modulus of clays to the SPT N value as:
M ds = f ⋅ N ⋅ Pa (8-33)
where:
f = empirical coefficient related to plasticity index from Figure 8-46,
Pa = atmospheric pressure, and
N = SPT blow count.
According to Kulhawy and Mayne (1990), Stroud’s correlation is not very reliable and
should be used with caution.
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Figure 8-43 Relationship between Modulus Number and Void Ratio for NC Soils
(after Janbu 1963)
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Figure 8-44 Modulus Number for NC Clays (after Janbu 1985)
Figure 8-45 Modulus Number for NC Silts and Sands (after Janbu 1985)
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Figure 8-46 Variation of Empirical Coefficient f used for Calculating
Constrained Modulus with PI (after Stroud 1974)
8-4.3.2 Correlations with Cone Penetration Test.
Numerous correlations have been presented to estimate the value of constrained

modulus using the results from the cone penetration tests. Most of these correlations
use the form of the equation shown below:
M ds= α ⋅ qc (8-34)
where:
α = empirical coefficient.
Mitchell and Gardner (1975) compiled values of α for different soils and showed that α
can range from 0.4 to 8. In most cases, α is between 1 and 3. These values were
obtained using a variety of cones with different geometries and testing procedures.
Kulhawy and Mayne (1990) presented the correlation shown in Figure 8-47 to obtain the
constrained modulus of clays based on CPTu data in the form of tip resistance
corrected for pore pressure and overburden stress.
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Figure 8-47 Correlation between Normalized Constrained Modulus and
Normalized qt from CPTu for Clays (after Kulhawy and Mayne 1990)
8-4.4 Coefficient of Secondary Compression.
The coefficient of secondary compression defines the settlement as a function of time

after primary consolidation is completed. As with the compression ratios, the coefficient
of secondary compression can be defined as a function of strain ( Cεα ) or as a function
of the void ratio ( Cα ). See Section 5-5.4 for more details.
Mesri (1973) summarized the data shown in Figure 8-48 to estimate the coefficient of
secondary compression of NC clays using the natural water content. Based on that
data, the modified coefficient of secondary compression can be estimated as:
Cεα = 0.0001wn (8-35)
where:
Cεα = secondary compression ratio and
wn = natural water content.
According to Kulhawy and Mayne (1990), Cεα ranges from 0.0005 and 0.001 for most
overconsolidated clays. The ratio of the coefficient of secondary compression to the
compression index ( Cα Cc = Cεα Cε c ) is more or less constant for a given soil (Mesri and
Godlewski 1977). Values of Cα Cc are summarized in Table 8-17. Another correlation
for the coefficient of correlation for silts and clays is presented in Figure 8-49.
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Figure 8-48 Correlation between Modified Secondary Compression Index and
Natural Water Content for Normally Consolidated Clays
(after Mesri 1973, Holtz and Kovacs 1981)
Table 8-17 Typical Values of Cα C c for Natural Soils

(after Mesri and Godlewski 1977)
Grouping Soil Type Cα Cc

Whangamarino clay 0.03 – 0.04
Leda clay 0.025 – 0.06
Soft blue clay 0.026
Portland sensitive clay 0.025 – 0.055
Inorganic clays San Francisco Bay Mud 0.04 – 0.06
and silts New Liskeard varved clay 0.03 – 0.06
Silty clay C 0.032
Nearshore clays and silts 0.055 – 0.075
Mexico City clay 0.03 – 0.035
Hudson River silt 0.03 – 0.06
Norfolk organic silt 0.05
Calcareous organic silt 0.035 – 0.06
Organic clays
Post-glacial organic clay 0.05 – 0.07
and silts
Organic clays and silts 0.04 – 0.06
New Haven organic clay silt 0.04 – 0.075
Amorphous and fibrous peat 0.035 – 0.083
Canadian muskeg 0.09 – 0.10
Peats Peat 0.075 – 0.085
Peat 0.05 – 0.08
Fibrous peat 0.06 – 0.085
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Figure 8-49 Secondary Compression Index for Silts and Clays
8-4.5 Coefficient of Consolidation.
The coefficient of consolidation ( cv ) is a difficult parameter to estimate for design use

because in situ stratigraphy can include sand seams and lenses, varved layers, etc.
Small laboratory test specimens may not contain these fabric elements. In addition, the
coefficient of consolidation can differ in the vertical and horizontal directions.
Laboratory tests normally only measure the coefficient of consolidation in one direction.
A first-order approximation for the coefficient of consolidation can be obtained using
Figure 8-50.
Figure 8-50 Approximate Relationship between Coefficient of Consolidation and

Liquid Limit
8-5 ELASTIC PARAMETERS.
8-5.1 Definitions.
Correlations to four elastic parameters will be presented in this section: (1) Young’s
modulus ( E ) is the ratio of the change in normal stress in a given direction to the strain
in the same direction within the elastic range; (2) bulk modulus ( K ) is the change in
mean stress divided by the corresponding volumetric strain; (3) shear modulus ( G ) is
the ratio of the change shear stress divided by the shear strain caused by that stress;
and (4) Poisson’s ratio (ν ) is the ratio of the lateral strain to the axial strain caused by a
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change of stress. The relationships between these parameters are summarized in
Table 8-18.
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Table 8-18 Relationships between Common Elastic Parameters
Parameter E= K = G= ν =
EG E − 2G
E &G --- ---
3( 3G − E ) 2G
E E
E &ν --- ---
3(1− 2ν ) 2(1+ν )
3 KE 3K − E
E&K --- ---
9K −E 6K
2G (1+ν )
G &ν 2G (1+ν ) --- ---
3(1− 2ν )
9 KG 3 K − 2G
G&K --- ---
3 K +G 2( 3 K +G )
3 K (1− 2ν )
K &ν 3 K (1− 2ν ) --- ---
2(1+ν )
In fine-grained soils, the value of the Young’s modulus from field testing is normally
derived under undrained conditions ( Eu ). Assuming a Poisson’s ratio for undrained
conditions of 0.5, the Young’s modulus under drained conditions can be found:
2
E
= (1 +ν ) Eu (8-36)
3
where:
E = Young’s modulus for drained conditions,
ν = Poisson’s ratio for drained conditions, and
Eu = Young’s modulus for undrained conditions.
The shear modulus represents the response of the soil skeleton and it is independent of
drainage conditions (i.e., G = Gu ) (Kulhawy and Mayne 1990). Assuming a Poisson’s
ratio for undrained conditions of 0.5, the relationships in Table 8-18 indicate that Eu is
three times greater than the shear modulus. According to Kulhawy and Mayne (1990),
the undrained Young’s modulus of soils is stress path dependent.
8-5.2 Undrained Young’s Modulus of Fine-Grained Soils.
8-5.2.1 Typical Values.
Typical values of undrained Young’s modulus are summarized in Table 8-19.
8-5.2.2 Correlations with Undrained Shear Strength.
Duncan and Buchignani (1987) proposed a correlation to obtain the ratio of the
undrained shear modulus to the undrained shear strength for fine grained soils as a
function of the plasticity index and overconsolidation ratio as shown in Figure 8-51. This
correlation is based on results obtained from direct simple shear tests.
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Table 8-19 Typical Range of Undrained Young’s Modulus for Clays
Eu Pa
Clay
Consistency Kulhawy and Mayne
USACE (1990)
(1990)
Very soft 5 – 50
Soft 50 – 200 15 – 40
Medium 200 – 500 40 - 80
Stiff or silty 500 – 1000 80 – 200
Sandy 250 – 2000
Clay shale 1000 – 2000
Figure 8-51 Correlation of Undrained Modulus Normalized by Undrained Shear

Strength to Overconsolidation Ratio (after Duncan and Buchignani 1987)
8-5.2.3 Correlations with Standard Penetration and Pressuremeter Tests.
The pressuremeter test is used to directly measure a soil modulus in the horizontal
direction. According to Kulhawy and Mayne (1990), the horizontal modulus measured
in clay using the pressuremeter test is approximately equal to the undrained Young’s
modulus.
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Ohya et al. (1982) presented the relationship for the undrained Young’s modulus from
the pressuremeter ( EPMT ) and the SPT N value shown in Figure 8-52. The relationship
shown in this figure exhibits a large amount of scatter ( r 2 = 0.39 ), and it should be used
with caution.
Figure 8-52 Correlation between PMT Modulus for Clays and SPT N
(after Ohya et al. 1982)
8-5.2.4 Correlations to Load Tests.
Poulos and Davis (1980) presented the relationship for the undrained Young’s modulus
calculated from load tests on drilled shafts and driven piles as a function of the
undrained shear strength of the clay, as shown in Figure 8-53.
Callanan and Kulhawy (1985) presented the relationship for the undrained Young’s
modulus of clays back calculated from load tests on drilled shafts and spread
foundations presented in Figure 8-54. In the right side of Figure 8-54, σ vm is the mean
total vertical stress over the foundation depth.
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Figure 8-53 Undrained Modulus for Deep Foundations in Compression

(after Poulos and Davis 1980)
Figure 8-54 Undrained Modulus for (left) Drilled Shafts in Compression and Uplift
and (right) Spread Foundations in Uplift (after Callanan and Kulhawy 1985)
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8-5.3 Drained Young’s Modulus of Coarse-Grained Soils.
Coarse-grained soils only experience undrained conditions for a very short period of
time and under special circumstances. For this reason, the Young’s modulus is only
required for drained conditions in most cases. Typical values of drained Young’s
modulus ( E ) based on relative density are summarized in Table 8-20 for coarse-
grained soils.
Table 8-20 Typical Ranges of Drained Young’s Modulus for Coarse-Grained

Soils
Normalized Drained Young’s Modulus ( E Pa )

Relative
Density or Poulos (1975)
Soil Type USACE (1990)
Typical Driven Piles
Loose 100 – 200 275 – 550 100 – 250
Medium 200 – 500 550 – 700 250 – 1000
Dense 500 – 1000 700 – 1100 1000 – 2000
Silty sand --- --- 250 – 2000
8-5.3.1 Correlations with SPT N Values.
Many different correlations have been developed to estimate the drained Young’s
modulus of coarse-grained soils using SPT N values. Some of these correlations are
presented in Table 8-21 and plotted in Figure 8-55.
Table 8-21 Correlations for Drained Young’s Modulus of Coarse-Grained Soils

using SPT N Values
Equation Applicable to Reference

2
E Pa ≈ 7.1(1 − ν ) N Sandy soils After Ferrent (1963)
E Pa ≈ 4.7 ( N +15 ) Sands

After Webb (1969)
E Pa ≈ 3.1 ( N +5 ) Clayey sands
E Pa ≈ C ( N + 6 ) Silts with sand and gravels with sand with N < 15

Begemann (1974)
E Pa ≈ 39.5 + C ( N −6 ) Silts with sand and gravels with sand with N ≥ 15
E Pa ≈ 5 N 60 Sands with fines

Kulhawy and Mayne
E Pa ≈ 10 N 60 NC clean sands
(1990)
E Pa ≈ 15 N 60 OC clean sands
E = drained Young’s modulus,

Pa = atmospheric pressure (in same units as Ed ),
N = SPT blow count value (use N60 for modern samplers)
ν = Poisson’s ratio, and
C = empirical coefficient equal to 3 for silts with sand and 12 for gravel with sand
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Figure 8-55 Correlations for Drained Young’s Modulus of Granular Soils
8-6 CALIFORNIA BEARING RATIO (CBR).
The California bearing ratio ( CBR ) is a penetration test developed by the California
Department of Transportation to assess the load-bearing capacity of soils used for
building roads. This test is described in ASTM D1883. The CBR is predominately a
laboratory test, but there are methods available to measure CBR in the field. CBR can
be determined for test specimens that have been soaked, or specimens at their
compaction water content. Because these correlations apply to specific soil types, the
original references should be consulted before using the correlations.
8-6.1 Correlations with Index and Compaction Properties.
Several researchers have presented correlations to estimate the CBR using index
properties. Correlations to grain size, Atterberg limits, and other index properties are
summarized in Table 8-22. Correlations between CBR and soil compaction properties
are summarized in Table 8-23.
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Table 8-22 CBR Correlations with Grain Size, Atterberg Limits, and Unit Weight
Equation Reference / Comment
75
CBR = NCHRP (2001), FC × PI > 0
1 + 0.728 ( FC × PI )
0.358
CBR = 28.09 ( D60 ) NCHRP (2001), FC × PI =
0
CBR = 95 NCHRP (2001), D60 ≥ 30 mm
CBR = 5 NCHRP (2001), D60 ≤ 0.01 mm
CBR 0.24GC + 3.1

=
Yildirim and Gunaydin (2011)
= 18.5 − 0.18 FC
CBR
4.75 − 0.044 LL + 0.15 PL

CBRunsoaked =
Patel and Desai (2010)
5.18 − 0.028 LL − 0.047 PL
CBR =
23 + 1.42γ d − 0.213 PI − 0.916 w − 0.368 LL

CBR = George et al. (2009)
9.7915
CBR = 0.0004Gs
−8 6.6141
= 1.36 × 10
CBR (γ m ) Yashas et al. (2016)
−10 7.4106
= 9.27 × 10
CBR (γ d )
= 1.93β − 31
CBRsoaked Al-Hashemi and Bukhary (2016)
FC = fines content = percent passing #200 sieve,

SC = sand content = percent retained between #4 and #200 sieve,
GC = gravel content = percent retained between 75 mm and #4 sieve,
LL = liquid limit, PL = plastic limit, PI = plasticity index,
γ d = dry unit weight (in kN/m3), γ m = moist unit weight (in kN/m3),
w = water content (in percent), Gs = specific gravity of solids, and
β = angle of repose
Table 8-23 CBR Correlations to Index and Compaction Properties

(after Singh et al. 2011)
Equation
 γd  w 
= 33 
CBR  − 5.5   − 1.15 PL − 2.21
 γ d ,max   wopt 
 γd   w 
= 24 
CBRunsoaked  − 67   − 2 PL + 104.7
 γ d ,max   wopt 
PL = plastic limit,
γ d = dry unit weight, γ d ,max = maximum dry unit weight from Modified Proctor
test,
w = water content, and wopt = optimum water content from Modified Proctor test.
Water contents should be either both in decimal or both in percentage form.
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8-6.2 Correlations with Dynamic Cone Penetration.
Many correlations have been developed to estimate CBR from the results of dynamic
cone penetration tests. Most of these correlations can take the form
CBR= A ⋅ DCP x (8-37)
where:
CBR = California Bearing Ratio,
A = empirical coefficient,
DCP = dynamic cone penetration index (mm/blow), and
x = empirical exponent.
The values of A and x found by various researchers are summarized in Table 8-24.
The range of CBR values expected from these correlations is presented in Figure 8-56.
Table 8-24 CBR Correlations with DCP

Reference / Comment A x
Gabr et al. (2000) 25 -0.55
Feleke and Araya (2016), fine-grained soils 47 -0.90
George et al. (2009) 47 -0.79
White et al. (2018), CL with CBR < 10 59 -2.00
Feleke and Araya (2016), coarse-grained soils 90 -1.17
Feleke and Araya (2016), fine-grained soils 104 -0.91
Feleke and Araya (2016), coarse-grained soils 157 -0.85
George and Kumar (2018) 246 -1.35
White et al. (2018), All soils except CL with CBR < 10 292 -1.12
Smith and Pratt (1983) 363 -1.15
Harrison (1986), clay-like soils with DCP < 10 mm/blow 501 -1.12
Harrison (1986), clay-like soils with DCP > 10 mm/blow 363 -1.16
Kleyn and Van Heerden (1983) 425 -1.27
Nazzal (2003) developed a similar correlation to those in Table 8-24 for soils with DCP
values between 6.3 and 67 mm/blow:
2559
=CBR +1 (8-38)
DCP1.84 − 7.35
where:
CBR = California Bearing Ratio, and
DCP = dynamic cone penetration index (mm/blow).
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Figure 8-56 Range of CBR based on DCP Predicted by Correlations
Livneh (1989) found that the CBR could be estimated from the results of the Standard
Penetration Test as shown below. The data used to obtain this correlation is presented
in Figure 8-57.
−0.26
 300 
log CBR =
−5.13 + 6.55  log  (8-39)
 N 
where:
CBR = California Bearing Ratio, and
N = SPT blow count value (use N 60 for modern hammers).
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Figure 8-57 Correlation for CBR in Terms of SPT N Value (Livneh 1989)
8-7 HYDRAULIC CONDUCTIVITY.
The hydraulic conductivity (a.k.a., the coefficient of permeability) governs the flow rate
and head loss as water flows through a soil mass. The hydraulic conductivity has one
of the widest ranges of any engineering parameter as can be seen in Table 8-25. The
wide range of possible values and the influence of variable ground conditions on the
hydraulic conductivity make this parameter difficult to evaluate with a high degree of
accuracy.
8-7.1 Typical Values.
Typical values of hydraulic conductivity based on soil type are presented in Table 8-25.
Other typical values based on soil type can be found in Section 6-3.3.
8-7.2 Correlations for Coarse-Grained Soils.
One of the first correlations to estimate the hydraulic conductivity based on grain size
was presented by Hazen (1911). This correlation was developed for saturated clean
sands with a fines content less than 5% and D10 values ranging from 0.1 mm to 3 mm:
k = C ( D10 ) 2 (8-40)
Where:
k = hydraulic conductivity (cm/s),
C = empirical coefficient, usually taken to be 1 cm/s/mm2, and
D10 = grain size corresponding to 10% passing on the grain-size distribution (mm).
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Table 8-25 Typical Ranges of Hydraulic Conductivity based on Soil Type
(after Terzaghi et al. 1996)
Soil Hydraulic Conductivity (m/sec) Relative Permeability
Gravel > 10-3 High
Sandy gravel
Clean sand 10-3 to 10-5 Medium
Fine sand
Sand
Dirty sand 10-5 to 10-7 Low
Silty sand
Silt
10-7 to 10-9 Very low
Silty clay
Clay < 10-9 Practically impermeable
Based on data from the middle and lower Mississippi River Valley, the USACE (1993)
correlated the in situ horizontal permeability of fine to medium, relatively uniform sands
(USCS classifications of SP or SW) as shown in Figure 8-58. This correlation was
recommended for use only within the geographic area for which it was developed.
Figure 8-58 Horizontal Hydraulic Conductivity based on D10 (after USACE 1993)
As shown in Figure 8-59, Kenney et al. (1984) correlated the hydraulic conductivity of
coarse-grained soils to the grain size corresponding to 5% passing on the cumulative
grain-size distribution curve ( D5 ). A similar correlation can be developed between the
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hydraulic conductivity for clean coarse-grained soils and the D10 grain size as shown in
Figure 8-60.
Figure 8-59 Hydraulic Conductivity based on D5 (after Kenney et al. 1984)
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Figure 8-60 Hydraulic Conductivity of Sands and Sand-Gravel Mixtures as a

Function of D5 , D10 , C u , and e
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Carrier (2003) presented a modified version of the Kozeny-Carman equation ( Kozeny
1927; Carman 1938, 1956) to estimate the hydraulic conductivity using the full grain-
size distribution of a soil. The modifications introduced by Carrier (2003) to the original
equation simplify its use. The hydraulic conductivity can be estimated by:
2
 
 100%   1   e3 
k 1.99 ×102 
=   2   (8-41)
 fi   S  1+ e 
∑
 Dli + Dsi0.596
0.404

where:
k = hydraulic conductivity (cm/s),
fi = fraction of particles (by mass) between two adjacent sieve sizes,
Dli = the particle size of the coarser sieve (mm),
Dsi = the particles size of the finer sieve (mm),
S = surface area factor ranging from 6 for spheres to 8.5 for angular particles, and
e = void ratio.
Additional discussion and correlations of k to grain size can be found in Section 6-3.3.
8-7.3 Correlations for Fine-Grained Soils.
Unlike coarse-grained soils, the hydraulic conductivity of fine-grained soils is difficult to

estimate from index properties. Hydraulic conductivity correlations for fine-grained soils
should be used with caution.
Carrier and Beckman (1984) related the hydraulic conductivity and the Atterberg limits
and void ratio of fine-grained soils using:
4.29
 e − 0.027 ( PL − 0.242 PI )   1 
k = 0.0174     (8-42)
 PI   1+ e 
where:
k = hydraulic conductivity in m/s,
e = void ratio,
PL = plastic limit, and
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Benson et al. (1994) measured the hydraulic conductivity on intact test specimens
obtained from compacted clay liners from 67 landfills in North America. The results of
those tests were found to relate to other measured properties of the clays as:
 
894  w 
ln k =
−18.35 + − 0.08 PI − 2.87  + 0.32 GC + 0.02CF (8-43)
W γw − 1 
 γ Gs 
 d
where:
k = hydraulic conductivity in m/s,
CF = clay-sized fraction (percent by mass smaller than 0.002 mm),
GC = gravel content = percent retained between 75 mm and #4 sieve,
W = weight of field compactor (kN),
PI = plasticity index,
w = molding water content,
γ w = unit weight of the water,
γ d = dry unit weight, and
Gs = specific gravity of the solids.
Benson and Trast (1995) performed hydraulic conductivity tests on 13 compacted clays
used for compacted clay liners around the United States. The test specimens were
prepared at different water contents, compacted, and then tested to measure the
hydraulic conductivity. The results were correlated by Benson and Trast to k using:
 
 w 
ln k =−15 − 0.087  − 0.054 PI + 0.022CF + 0.91E (8-44)
γw − 1 
 γ Gs 
 d
where:
k = hydraulic conductivity (m/s),
w = molding water content,
γ w = unit weight of the water,
γ d = dry unit weight,
Gs = specific gravity of the solids.
PI = plasticity index,
CF = clay-sized fraction, and
E = compactive effort index (equal to -1, 0, and 1 for modified, standard, and reduced
Proctor compactive effort, respectively).
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8-8 SHEAR WAVE VELOCITY.
Many correlations have been developed to estimate the shear wave velocity ( Vs ) from
SPT N values. Judgment is required regarding the use of uncorrected vs. corrected
blow count. Older correlations where likely developed using 60% hammer efficiency as
represented by N 60 . The general form for most of the Vs correlations is:
Vs = BN x z y (8-45)
where:
Vs = shear wave velocity (m/s),
B , x , and y = empirical coefficients,
N = SPT blow count, and
z = depth to the soil layer (m).
Correlations using the form of Equation 8-45 are summarized in Table 8-26. Other
correlations are presented in Table 8-27. The range of values expected from these
correlations can be seen in Figure 8-61.
Figure 8-61 Range of Shear Wave Velocities based on SPT N Value Predicted by
Correlations
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Table 8-26 Shear Wave Velocity Correlated to SPT N Value and Depth
B x y Comments Reference
131 0.21 0 Sand, use N 60
108 0.24 0 Use N 60
105 0.26 0 Use N 60
Hasancebi and Ulusay (2007)
98 0.27 0 Clay
90 0.31 0
91 0.32 0 Sands
101 0.27 0 Sand
96 0.30 0 Maheswari et al. (2010)
89 0.36 0 Clay
101 0.29 0 Sykora and Stokoe (1983)
68 0.29 0 Kiku et al. (2001)
80 0.29 0 Clay
81 0.33 0 Sand Imai (1977)
91 0.34 0
84 0.31 0 Ohba and Toriumi (1970)
114 0.31 0 Clay
57 0.49 0 Sand Lee (1990)
106 0.32 0 Silt
97 0.31 0 Imai and Tonouchi (1982)
76 0.33 0 Imai and Yoshimura (1975)
92 0.34 0 Fujiwara (1972)
90 0.34 0 Imai et al. (1975)
85 0.35 0 Ohta and Goto (1978)
108 0.36 0 Athanasopoulos (1995)
59 0.39 0 Dikmen (2009)
81 0.39 0 Ohsaki and Twasaki (1973)
61 0.50 0 Seed and Idriss (1981)
52 0.52 0 Iyisan (1996)
59 0.11 0.43 Akin et al. (2011)
69 0.17 0.20 Clay Jamiolkowski et al. (1988)
82 0.25 0.14 Gravel Yoshida et al. (1988)
Table 8-27 Shear Wave Velocity Correlated to SPT N Value

Equation Reference
0.202
Vs 116 ( N + 0.3185 )
= Jinan (1987)
( )
0.36
Vs ,1 = 61.89 N1,60 CS Ulmer et al. (2020)
0.25
Vs ,1 = normalized shear wave velocity = Vs ( Pa σ 'v ) , and
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Equation Reference
N1,60 CS = N1,60 corrected for fine content
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8-8.2 Correlations with Cone Penetration Test.
Several correlations relating the shear wave velocity to the results of the cone
penetration test have been developed and are summarized in Table 8-28. It should be
noted that some cones are equipped with a geophone or accelerometer and can
measure shear wave velocity directly.
Table 8-28 Correlations for Shear Wave Velocity with CPT results
Equation Comments Reference
Vs 134 + 0.0052 qc
= Sands Sykora and Stokoe (1983)
Vs = 0.1qc Clays Jaime and Romo (1988)

0.33 0.27
Vs = 17.5qc σ 'v Sands Baldi et al. (1989)
0.23
Vs1 = 102 qc1 Fear and Robertson (1995)
0.3
1.67  100 fs 
=Vs (10.1log qc − 11.4 )  q 
 c 
0.359 −0.473
Vs = 14.1qc e
Clays Hegazy and Mayne (1995)
0.549 0.025
Vs = 3.18qc fs
0.192 0.179
Vs = 13.2 qc σ 'v
Sands
0.319 −0.0466
Vs = 12 qc fs
0.435 −0.532
Vs = 9.44 qc e
Clays Mayne and Rix (1995)
0.627
Vs = 1.75 qc
0.089 0.1219 0.215
Vs = 32.3qc fs z
0.269 0.109 0.127
Vs = 11.9 qc fs z Clays Piratheepan (2002)
0.103 0.029 0.155
Vs = 25.3qc fs z Sands
0.25
0.103  σ 'v  1.788 I c

Vs ,1 = 0.0831qc  P  e Hegazy and Mayne (2006)
 a 
=Vs 119 log f s + 18.5 Mayne (2006)
0.5
 0.55 I c +1.68 q − σ 'v 
Vs = 10 
t
 Robertson (2009)

  Pa 
0.205
Vs ,1 = 149 qc1 Karray et al. (2011)
0.489
Vs ,1 = 16.88 ( qc1NCS ) Ulmer et al. (2020)
Vs = shear wave velocity (m/s), Vs ,1 = normalized shear wave velocity = Vs ( Pa σ 'v ) 0.25 ,
n
qc = cone tip resistance (kPa), qc1 = normalized cone tip resistance = ( qc Pa )( Pa σ 'v )
qc1NCS = normalized cone tip resistance as detailed by Boulanger and Idriss (2014)
f s = cone sleeve resistance (kPa), e = void ratio, z = depth of the soil layer,
σ 'v = effective vertical stress (kPa), I c = soil index for estimating grain characteristics, and
n = 0.5 for I c < 2.6, else 0.75.
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Topic Reference
Kulhawy, F. H. and Mayne, P. W. (1990). Manual on estimating soil properties
Soil Properties for foundation design (No. EPRI-EL-6800). Electric Power Research Inst., Palo
Alto, CA; Cornell Univ., Ithaca, NY, Geotechnical Engineering Group.
Carter, M. and Bentley, S. P. (2016). Soil properties and their correlations.
John Wiley & Sons.
Geotechnical Correlations Duncan, J. M., Horz, R. C., and Yang, T. L. (1989). Shear Strength
Correlations for Geotechnical Engineering, CGPR#4, Center for Geotechnical
Practice and Research, Blacksburg, VA, 93 pp.
8-10 NOTATION.
Symbol Description
A Empirical coefficient for Equation 8-37
CPT cone net area ratio, and empirical coefficient related to the steepness of the power function
a (Equation 8-15)
B Foundation width or diameter, and empirical coefficient in Equation 8-45
b Empirical coefficient related to the curvature of the power function
CBR California Bearing Ratio
CBRsoaked Soaked California Bearing Ratio
CBRunsoaked Unsoaked California Bearing Ratio
Cε c Modified compression index or compression ratio

∗
C c
Intrinsic compression index
CF Clay-sized fraction
Cε r Modified recompression index or recompression ratio
Cu Coefficient of uniformity
cv Coefficient of consolidation
Cα Secondary compression index
Cαε Modified secondary compression index or secondary compression ratio
D Foundation embedment
DCP Dynamic cone penetration index

D ji The particle size of the coarser sieve in Equation 8-41
Dr Relative density
Dsi The particle size of the finer sieve in Equation 8-41

Particle-size diameter corresponding to 5% passing on the cumulative particle-size distribution
D5 curve
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Symbol Description
D10 curve
D60 curve
E Young’s modulus, and compactive effort index in Equation 8-44
e Void ratio
ED Dilatometer modulus
E Young’s modulus, typically drained
eL Void ratio at a water content equal to the liquid limit
EPMT Young’s modulus from pressuremeter
Eu Undrained Young’s modulus

∗
e100 Intrinsic void ratio at 100 kPa
∗
f Empirical coefficient for Equation 8-33
FC Fine contents
fi Fraction of particles (by mass) between two adjacent sieve sizes in Equation 8-41
fs CPT sleeve resistance
G Shear modulus
GC Gravel content
Gu Undrained shear modulus
Gs Specific gravity of solids
Ic Soil index
Iv Void index
I v , ICL Void index for the intrinsic compression line
I v , SCL Void index for the sedimentation compression line
K Bulk modulus
k Hydraulic conductivity
KD Dilatometer horizontal stress index
KO Coefficient of earth pressure at rest
LL Liquid limit
m Semi-empirical fitting parameter for Equation 8-20, and modulus number
M ds Constrained modulus
N SPT N value. May be assumed to be equal to N 60

n Porosity, number of datapoints, and coefficient for qc1
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Symbol Description
NC Normally consolidated
Nc Empirical bearing capacity factor
Nk Empirical bearing capacity factor
N kt Empirical bearing capacity factor
Np Slope of the σ 'v vs qc plot
Nq Bearing capacity number
N 60 SPT N value corrected for 60% hammer energy efficiency
N1,60 N 60 value corrected to an effective vertical overburden of 1 atm
N1,60CS N1,60 corrected for fine content
N' SPT N value corrected for dynamic pore pressure effects

OC Overconsolidated
PI Plasticity index
PL Plastic limit
p0 Pressure required to initiate movement of the dilatometer
qc Static CPT tip resistance
qc1 Cone penetrometer tip resistance normalized to 1 atm overburden pressure.
qc , NC Static CPT tip resistance in normally consolidated sand
qc ,OC Static CPT tip resistance in overconsolidated sand
qd Dynamic cone resistance
qp Static CPT net resistance
qt Cone penetrometer tip resistance corrected for pore pressure effects
R Correction factor for overconsolidated static cone tip resistance
r2 Coefficient of determination
s Effective stress shear strength
S Surface area factor for Equation 8-41
S .D. Standard deviation
St Sensitivity
su , fv Undrained shear strength from field vane tests
u or u2 Pore pressure measured behind the cone tip
u0 Hydrostatic initial pore pressure
Vs Shear wave velocity
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Symbol Description
Vs ,1 Normalized shear wave velocity
W Weight of the field compactor in Equation 8-43
w Water content
wopt Optimum water content referenced to a given standard
x Empirical exponent for Equation 8-37 and empirical coefficient for Equation 8-45
y Empirical coefficient for Equation 8-45
z Depth below the soil layer
α Empirical coefficient for Equation 8-34

Inclination of the failure wedge for foundation loading, exponent in Equation 8-9, and angle of
β repose
∆q p Change in static CPT cone net resistance
∆σ 'v Change in vertical effective stress
φ 'FS Fully softened friction angle
φ 'r Residual friction angle
φ 'sec Effective stress secant friction angle (stress dependent)
γd Dry unit weight
γ d ,max Maximum dry unit weight referenced to a given standard
γm Moist unit weight
γw Unit weight of the water
σv Vertical total stress
σ vm Mean vertical total stress
σ v0 Initial vertical total stress
σ ' ff Effective normal stress on the failure plane at failure
σ 'p Preconsolidation pressure or maximum past pressure
σ 'v Vertical effective stress
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APPENDIX A. REFERENCES
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498
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APPENDIX B. LIST OF COMPUTER PROGRAMS
Table B-1 List of Computer Programs

Name Company Application Website
ADAMA Assessing stresses and settlements under
FOSSA Engineering embankment and footings acting on http://www.geoprograms.com/
Inc. horizontal ground surfaces
ADAMA
Interactive program for the design of
GeoCoPS Engineering http://www.geoprograms.com/
geosynthetic tubes
Inc.
ADAMA
Design and analysis of mechanically
MSEW Engineering http://www.geoprograms.com/
stabilized earth walls
Inc.
ADAMA
Interactive, design-oriented, program for
Reslope Engineering http://www.geoprograms.com/
geosynthetic-reinforced slopes
Inc.
ADAMA
Assessing the rotational and translational
ReSSA+ Engineering http://www.geoprograms.com/
stability of reinforced slopes and walls
Inc.
ADINA R&D Stress analysis of solids (2D and 3D) and
ADINA http://www.adina.com/
Inc. structures in statics and dynamics
Analyzing stability of slopes for road,
AEC Logic
AEC Slope railways, river training works, canal http://www.aeclogic.com/
Pvt. Ltd
embankment, dams etc.
Reporting and managing subsurface data,
gINT Bentley
including borehole logs, well logs, CPT https://www.bentley.com/
Professional System Inc.
data, and geophysical logs.
Bridge
Static axial capacity program used for
FB-Deep Software https://bsi.ce.ufl.edu/
drilled shafts and driven piles
Institute
Bridge Nonlinear finite element analysis program
FB-MultiPier Software capable of analyzing multiple bridge pier https://bsi.ce.ufl.edu/
Institute structures interconnected by bridge spans.
Used to select a suitable pile type for known
soil strata by investigating the effects of soil
parameters and different pile types. Allows
the bearing capacity of individual piles and
Bearing Pile
CADS groups of piles of various lengths and types https://cads.co.uk/
designer
to be checked, including bored piles,
continuous flight auger (CFA) piles, driven
cast in place, driven tubular steel, driven
steel H piles and driven precast piles.
Piled Wall
Suite
Analysis and design of embedded walls in
concrete or steel. Includes analysis and
CADS https://cads.co.uk/
design for sheet piles, king piles, contiguous
and secant bored piles and diaphragm walls
Designing and checking of bases. Can be

RC Pad Base used stand-alone or as part of the CADS
CADS https://cads.co.uk/
Designer integrated analysis, design, and detailing
solution.
Pile Cap Designer software that
automatically produces a selection of
RC Pile cap
CADS suitable designs to BS 8110 and EC2 for https://cads.co.uk/
designer
pile caps with 2-9 piles supporting circular
or rectangular columns
499
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Slope stability software package for
calculating the factor of safety of earth
Reslope CADS https://cads.co.uk/
slopes. Uses Bishop’s simplified method
and circular slip surfaces.
The AWall CAD Tool allows a user to
Callide
accurately represent the Plan and Elevation
AWALL Technologies http://www.ctiware.com/
views of a retaining wall on their grading
Inc
plan
Callide
Design and drawing of mechanically
VESPA2MSE Technologies http://www.ctiware.com/
stabilized earth retaining walls.
Inc
Mobile application designed to bridge the
Canary gap between data collection and
Mfield http://canarysystems.com/
System Inc. observations in the field, and the hosted
project database.
Performs for calculation of three-
CDM
dimensional boulders falling on a slope
IS GeoMassi Dolmen and https://www.cdmdolmen.it/
using the "Lumped Mass hybrid" method
omnia IS srl
associated with a statistical analysis.
CDM
Stability analysis of slopes in loose terrain
IS GeoPendii Dolmen and https://www.cdmdolmen.it/
based on limit equilibrium methods.
omnia IS srl
CDM Classification of the quality of rock masses
IS GeoRocce Dolmen and using the most widespread theories in the https://www.cdmdolmen.it/
omnia IS srl geo-mechanical field.
Numerical interpretation and graphic
CDM
representation of the results of SPT, DP
IS Geostrati Dolmen and https://www.cdmdolmen.it/
(Dynamic Probing), and CPT tests
omnia IS srl
performed on project sites.
Finite element analysis, according to the
CDM
NTC 2018 and Eurocode, of inland walls
IS Muri Dolmen and https://www.cdmdolmen.it/
with constant or variable section, with
omnia IS srl
buttresses, teeth, poles, and tie rods.
Designing flexible containment structures
CDM for which the soil-structure interaction is
IS Paratie Dolmen and analyzed in the nonlinear field with https://www.cdmdolmen.it/
omnia IS srl hysteresis taking into account the
deformability of the face.
CDM Geotechnical modules useful for the rough
IS ProGeo Dolmen and design of structures in contact with the https://www.cdmdolmen.it/
omnia IS srl ground.
CDM
IS PL Dolmen and Complete analysis of piles. https://www.cdmdolmen.it/
omnia IS srl
CDM
IS Plinti Dolmen and Analysis and design of surface foundations. https://www.cdmdolmen.it/
omnia IS srl
Civil and
Structural
MasterKey: Designing retaining walls with full control
Computer https://www.masterseries.com/
Retaining wall over the design process.
Services
Limited
Windows-based analysis program that
handles virtually all types of piles, including
steel pipes, H-piles, pre-cast concrete piles,
AllPile Civiltech, Inc. auger-cast piles, drilled shafts, timber piles, https://civiltech.com/
jetted piles, tapered piles, piers with bell,
micropiles (minipiles), uplift anchors, uplift
plate, and shallow foundations.
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Liquefaction analysis and settlement
Liquefy Pro Civiltech, Inc. https://civiltech.com/
analysis due to liquefaction.
Design and analysis tool containing four
Shoring Suit Civiltech, Inc. modules for shoring, earth pressure, https://civiltech.com/
surcharge, and heave.
Generating boring log and test pit graphical
Super Log Civiltech, Inc. reports for field drilling and geotechnical https://civiltech.com/
investigation.
Clover
Galena Slope stability analysis. http://www.galenasoftware.com/
Associates
Datgel Pty
CPT Tool 3.2 Analysis of CPT data. https://www.datgel.com/
Ltd
Geotechnical in situ and lab result storage
and reporting, including logs for boreholes,
Datgel Pty
DGD Tool 4 test pits, DCPs and vibrocores, and a large https://www.datgel.com/
Ltd
range of summary graphs, histograms,
fence, table, and map reports.
Datgel Lab and Datgel Pty
Analysis of laboratory and in situ tests. https://www.datgel.com/d
In Situ Tool 3 Ltd
Analysis of bearing capacity in accordance
DC Bearing DC Software https://www.dc-software.de/
with Eurocode 7.
Analysis of cantilever walls in accordance
DC Cantilever DC Software https://www.dc-software.de
with Eurocode 7.
Analysis and design of single, block and
DC Footing DC Software sleeve footings, rectangular, strip and https://www.dc-software.de
circular footings according to Eurocode 7.
Design and analysis of gabions and
DC Gabion DC Software supporting structures of layered blocks and https://www.dc-software.de
concrete stack stones.
Analysis of reinforced earth with
DC Geotex DC Software geosynthetics in accordance with Eurocode https://www.dc-software.de
7.
DC Inflit DC Software Analysis of infiltration. https://www.dc-software.de/
3D display of foundation pits with exact wall
DC Integra 3D DC Software geometry and automatic generation of slope https://www.dc-software.de/
intersections.
Stability analysis of diaphragm wall
DC Lamellae DC Software https://www.dc-software.de/
lamellae.
Analysis of soil nailing in accordance with
DC Nail DC Software https://www.dc-software.de/
Eurocode 7.
DC Pile DC Software Analysis and design of piles. https://www.dc-software.de/
Settlement analysis according to Eurocode
DC Settlement DC Software https://www.dc-software.de
7.
Slope stability analysis according to
DC Slope DC Software https://www.dc-software.de
Eurocode 7.
DC Analysis and design of underpinning and
DC Software https://www.dc-software.de/
Underpinning retaining walls.
Bore well logs, layer specifications, well,
DC Bore DC Software https://www.dc-software.de/
and gauge sinking.
DC Cone DC Software Conducting and interpreting CPT data. https://www.dc-software.de/
Data reduction of Atterberg limits according
DC Cons DC Software to DIN 18 122 / SN 670 345 / OENORM B https://www.dc-software.de/
4411 /CEN ISO/TS 17892-12.
Determination of lime content according to
DC Lime DC Software https://www.dc-software.de/
DIN 18 129.
Conduction and interpretation of plate load
DC Load DC Software https://www.dc-software.de/
testing.
DC Pump DC Software Pump test graphics and evaluation. https://www.dc-software.de/
501
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Shear strength test according to DIN 18 137
DC Shear DC Software https://www.dc-software.de/
and interpretation of results.
DC Sieve DC Software Sieve and sedimentation analysis. https://www.dc-software.de/
Deep
DeepFND Excavation Analysis and design of deep foundation. http://www.deepexcavation.com/
LLC
Geotechnical and structural design for many
wall types that include soldier pile walls,
Deep
sheet pile walls, and diaphragm walls with
DeepEX Excavation http://www.deepexcavation.com/
multiple sections of reinforcement. Can also
LLC
perform slope stability analysis with soil
nailing.
Deep
Inspecting pile installation records in three
Deviate VR Excavation http://www.deepexcavation.com/
dimensions or using virtual reality.
LLC
Deep
Helixpile Excavation Design and analysis of helical piles. http://www.deepexcavation.com/
LLC
Deep
Full design-visualization program for deep
HoloDeepex Excavation http://www.deepexcavation.com/
excavations.
LLC
Deep Soil nail analysis software. Follows the
Snail Plus Excavation FHWA methodology for the design of soil http://www.deepexcavation.com/
LLC nail walls.
Deep Evaluating the stability of slurry supported
Trench Excavation trenches and panels for 2D and 3D http://www.deepexcavation.com/
LLC analyses.
Deep
TriAxial PRO Excavation Processing triaxial test data. http://www.deepexcavation.com/
LLC
Package of eight design software: namely
Deltares D-Foundations, D-Geo Pipeline,
Geotechnical Deltares D-Geo Stability, D-Pile Group, https://www.deltares.nl/
Softwares D-Settlement, MWell, MSeep and
D-Sheet Piling
Simulation of storm surges, hurricanes,
tsunamis, detailed flows and water levels,
Delft3D
waves, sediment transport and morphology,
Flexible Mesh Deltares https://www.deltares.nl/
and water quality and ecology. Capable of
Suite
handling the interactions between these
processes.
D- Design of foundations following Eurocode 7
Deltares https://www.deltares.nl/
Foundations and Dutch and Belgian annexes.
Design of a pipeline installation in a trench
D-Geo and trenchless installation, using the micro
Deltares https://www.deltares.nl/
Pipelines tunneling technique or the Horizontal
Directional Drilling (HDD) technique.
D Geostability Deltares Slope stability analysis. https://www.deltares.nl/
Three-dimensional behavior of single piles
D Pile Groups Deltares and pile groups, interacting via the pile cap https://www.deltares.nl/
and the soil, as a function of loading.
Settlement analysis, offering accurate and
robust models, capturing consolidation,
D Settlement Deltares https://www.deltares.nl/
creep, submerging, drains, staged loading,
and unloading and reloading
Design retaining walls and horizontally
D- sheet piling Deltares https://www.deltares.nl/
loaded piles.
502
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Simulation of two-dimensional stationary
groundwater flow in a cross section of
M Seep Deltares layered soil structures or in one phreatic https://www.deltares.nl/
aquifer, composed of different material
areas.
Groundwater modeling to analyze time-
M Well Deltares dependent hydrogeological problems, such https://www.deltares.nl/
as dewatering, in multilayer soil profiles.
Finite element software package for
structural, geotechnical, tunneling,
Diana Diana FEA https://dianafea.com/
earthquake disciplines, and oil & gas
engineering.
Designing foundations for industrial
equipment such as horizontal exchangers,
Dimensional horizontal vessels, vertical vessels,
Foundation3D
Solutions, fractionation towers, air filters, pipe racks https://www.dimsoln.com/
2018
Inc and other plant supports or simply any
structure that needs a simple spread or
combined footing.
Dimensional
Design of soil and pile supported, multi-load
Mat3D 2018 Solutions, https://www.dimsoln.com/
point mat foundations.
Inc
Dimensional
DSAnchor Solutions, Designing anchors for concrete foundations. https://www.dimsoln.com/
Inc
Dimensional
Design and analysis of drilled shafts or
Shaft3D Solutions, https://www.dimsoln.com/
caisson type foundations.
Inc
Dr. Shallow foundation analysis, including
SoFA Konstantinos settlement calculations and static and http://sofasoftware.weebly.com/
Nikolaou seismic bearing capacity.
Axial capacity, as a function of depth, of a
APILE Ensoft Inc. driven pile in clay, sand, or mixed-soil https://www.ensoftinc.com/
profiles.
Equivalent dynamic stiffness and damping
DynaMat Ensoft Inc. of machine foundations using a three- https://www.ensoftinc.com/
dimensional hybrid method.
Dynamic response of both shallow and
deep foundations under harmonic, transient,
DynaN Ensoft Inc. https://www.ensoftinc.com/
and random loadings using the improved
Novak’s method.
Dynamic stiffness of single piles or pile
DynaPile Ensoft Inc. https://www.ensoftinc.com/
groups.
Analysis of mats or structural slabs
GeoMat Ensoft Inc. https://www.ensoftinc.com/
supported on soils.
Analysis of pile groups subjected to both
GROUP Ensoft Inc. https://www.ensoftinc.com/
axial and lateral loadings.
Analysis of a pile under lateral loading using
LPILE Ensoft Inc. https://www.ensoftinc.com/
the p-y method.
Distribution of load and axial deformation of
PileGPw Ensoft Inc. https://www.ensoftinc.com/
the piles within a pile group.
Flexible retaining wall systems considering
PYWALL Ensoft Inc. the soil-structure interaction using the https://www.ensoftinc.com/
beam-column model.
Settlement calculation for shallow and deep
SETOFF Ensoft Inc. https://www.ensoftinc.com/
foundations.
Axial capacity and the short-term, load-
SHAFT Ensoft Inc. settlement curves of drilled shafts or bored https://www.ensoftinc.com/
piles in various types of soils.
503
UFC 3-220-10
1 February 2022
2-D slope stability analysis using limit
STABLPRO Ensoft Inc. https://www.ensoftinc.com/
equilibrium method.
The t-z method to estimate the
TZPILE Ensoft Inc. displacement as a function of load for driven https://www.ensoftinc.com/
piles and drilled shafts.
Walls Retain Fides DV Analysis and design of retaining walls. http://www.fides-dvp.eu/
Fides Stability computations in geotechnics using
Fides DV http://www.fides-dvp.eu/
geostability kinematic element analysis methods (KEA).
FIDES Interactive generation and calculations of
Fides DV http://www.fides-dvp.eu/
Groundslab elastic semi-infinite space model.
Interactive graphical preprocessing for
FIDES-
Fides DV tunneling and geotechnical models for http://www.fides-dvp.eu/
WinTube-3D
SOFiSTiK solvers.
Geotechnical analysis based on analytical
Geo5 Fine https://www.finesoftware.eu/
and finite element methods.
Fitts Simulation and prediction of groundwater
AnAqSim Geosolutions conditions and groundwater/surface-water http://www.fittsgeosolutions.com/
LLC interactions. Alternative to MODFLOW.
Fitzroy Bundle software for structural design with
SCALE https://fitzroy.com/
System Ltd. foundation design components.
It is a bundle software for structural design
Fitzroy
LUCID and useful for design of different type of https://fitzroy.com/
System Ltd.
foundations and retaining walls.
Application suite for subsurface mapping
and data management to evaluate
GAEA contaminants, soil and rock properties,
Strata Explorer Technologies minerals, oil and gas deposits, and oil http://gaea.ca/
Ltd. sands. It is ideal for the environmental,
geotechnical, mining, oil sands, and
petroleum industries.
GAEA
Creation of boring and well logs and
WinLog RT Technologies http://gaea.ca/
managing boring and well data.
Ltd.
GAEA
Creation of grain-size analysis charts in
Winsieve Technologies http://gaea.ca/
several standard or custom formats.
Ltd.
Design of reinforced slopes in a wide variety
Geocentrix
Reactiv of soil types, using reinforced soil or soil http://www.geocentrix.co.uk/
Ltd.
nails.
Design of embedded retaining walls,
Geocentrix incorporating several UK and international
ReWard http://www.geocentrix.co.uk/
Ltd. design standards including BS 8002 and
Eurocode 7.
Geocentrix
Repute Onshore pile design and analysis. http://www.geocentrix.co.uk/
Ltd.
Geogiga
Geogiga Seismic data processing and interpretation
Technology http://www.geogiga.com/
Seismic Pro software.
corp.
CPeT IT Geologismiki Interpretation of Cone Penetration data. http://geologismiki.gr/
Cone Penetration Based soil liquefaction
software that for CPT data interpretation,
Cliq Geologismiki factor of safety, liquefaction potential index https://geologismiki.gr/
and post-earthquake displacements (both
vertical and lateral).
Liquefaction analysis that accepts SPT
LiqSvs Geologismiki https://geologismiki.gr/
and Vs field data.
Seismic signal processing and seismic
SPAS Geologismiki https://geologismiki.gr/
analysis.
504
UFC 3-220-10
1 February 2022
Assessment of liquefaction potential based
LiqIT Geologismiki https://geologismiki.gr/
on commonly used field data.
Vibro-replacement and design of stone
StoneC Geologismiki https://geologismiki.gr/
columns.
Settlement calculation taking into
SteinP 3DT Geologismiki consideration the influence of nearby https://geologismiki.gr/
footing elements.
Preliminary settlement analysis below a
SteinN Pro Geologismiki https://geologismiki.gr/
rectangular footing.
BLogPro Geologismiki Creation of simple soil borehole logs. https://geologismiki.gr/
Estimation of various soil properties from
SPTCorr Geologismiki https://geologismiki.gr/
the Standard Penetration Test blow count.
Prediction of settlement from in situ
GEODelp GEOS http://www.geos-ic.com/
measurements.
Calculation of settlement under
GEO Fond GEOS embankments and dimensioning of shallow http://www.geos-ic.com/
and deep foundation.
Design of retaining walls and analysis of
GEOMUR GEOS http://www.geos-ic.com/
internal and external stability.
Design of nailed wall cladding and
GEOSpar GEOS calculation of steel section and support http://www.geos-ic.com/
plates.
Slope stability, calculation of general
GEO Stab GEOS stability of supports, and dimensioning http://www.geos-ic.com/
reinforced floor and nailed walls.
Calculation of elastoplastic equilibria and
RIDO GEOS http://www.geos-ic.com/
dimensioning of retaining screens.
2-D and 3-D finite element numerical
Z-soil GEOS simulation and geotechnical calculation of http://www.geos-ic.com/
simple and complex structures.
Finite element simulation of air transfer in
AIR/W Geoslope https://www.geoslope.com/
mine waste and other porous media.
Finite element simulation of solute and gas
CTRAN/W Geoslope https://www.geoslope.com/
transfer in porous media.
Finite element simulation of earthquake
Quake/W Geoslope https://www.geoslope.com/
liquefaction and dynamic loading.
Finite element simulation of groundwater
SEEP/W Geoslope https://www.geoslope.com/
flow in porous media.
Finite element simulation of stress and
SIGMA/W Geoslope deformation in earth and structural https://www.geoslope.com/
materials.
SLOPE/W Geoslope https://www.geoslope.com/
equilibrium method.
Finite element simulation of heat transfer
TEMP/W Geoslope https://www.geoslope.com/
and phase change in porous media.
Integrated suite for simulation of slope
Geo Studio Geoslope stability, ground deformation, and heat and https://www.geoslope.com/
mass transfer in soil and rock.
Slope stability analysis, including features
ILA GeoSoft https://www.geoandsoft.com/
for retaining system designing
CE.CA.P GeoSoft Analysis and design of foundations https://www.geoandsoft.com/
Solution of dimensioning problems and
DIADIM GeoSoft https://www.geoandsoft.com/
verification through finite difference model.
Interpretation of static and dynamic
INSITU GeoSoft https://www.geoandsoft.com/
geotechnical in situ tests.
Analysis and design of retaining, gravity and
VERCAM GeoSoft https://www.geoandsoft.com/
in concrete walls.
505
UFC 3-220-10
1 February 2022
Determination of safety factors pertaining to
the liquefaction of incoherent saturated
LIQUITER GeoSoft https://www.geoandsoft.com/
terrains subjected to earthquake
phenomena.
Computerized structural geology data
collection and analysis, which recognizes
the discontinuity sets of a rock mass
CLUSTAR GeoSoft https://www.geoandsoft.com/
through hierarchical and non-hierarchical
clustering procedures derived from the
multivariate analysis.
3-D model for rock fall analysis and the
ROTOMAP GeoSoft https://www.geoandsoft.com/
design of rock fall protective systems.
Stability analysis of removable blocks on
ROCK3D GeoSoft https://www.geoandsoft.com/
planar rock slopes.
Computation of bearing capacity on rocky or
Load cap Geostru loose soils and analysis of soil reinforced https://www.geostru.eu/
with geogrid.
Design and analysis of gabion walls, simple
GDW Geostru concrete weirs, and gabion weirs in static https://www.geostru.eu/
and seismic conditions.
Mechanical analysis of soil using the finite
GFAS Geostru https://www.geostru.eu/
element method.
Calculation of the bearing capacity of the
Pile and
Geostru foundation terrain of a pile or micropile https://www.geostru.eu/
Micropile
(Screw-piles).
Design and analysis of reinforced concrete
retaining walls resting either on their own
MDC Geostru https://www.geostru.eu/
foundation or on piles, optionally supported
by tiebacks.
Design and analysis of sheet pile walls,
SPW Geostru https://www.geostru.eu/
drilled piles, and diaphragm walls.
Evaluation of localized instability rocky
elements affected by seismic movements
Rock Plane Geostru https://www.geostru.eu/
and/or by presence of water pressures
within intersurface fractures.
Down Hole Geostru Processing borehole seismic tests. https://www.geostru.eu/
Dynamic
Geostru Interpretation of Dynamic Penetration test. https://www.geostru.eu/
Probing
Geosysta Integrated data management system for
Adamas http://geosysta.com/
Ltd. geotechnical data.
Geosysta
Drillysis Borehole logging application. http://geosysta.com/
Ltd.
Stability analysis of cantilevered and
WALLAP Geosolve http://www.geosolve.co.uk/
propped cantilever retaining walls.
Slope Geosolve Slope stability analysis. http://www.geosolve.co.uk/
Analysis of retaining wall problems including
GWALL Geosolve http://www.geosolve.co.uk/
gravity walls and cantilever wall with bases.
Geotec Analysis of single piles, pile groups, and
ELPLA https://www.elpla.com/
Software piled raft foundation.
Soil liquefaction analysis, including
liquefaction potential, seismic settlement
Geotechnical
(dry and saturated) and lateral spreading
GeoLiqu Software and http://geoadvanced.com/
based on standard penetration test (SPT)
Services
data, cone penetration test (CPT) data and
shear wave velocity (Vs) data profiles.
Calculation of compression deformation
Geotechnical
utilizing Standard Penetration test (SPT),
GeoComp Software and http://geoadvanced.com/
cone penetration test (CPT), and shear
Services
wave velocity (Vs) data.
506
UFC 3-220-10
1 February 2022
Geotechnical
GeoBP Software and Bearing capacity analysis of soil. http://geoadvanced.com/
Services
Calculation of static and seismic lateral
Geotechnical earth pressures, utilizing trial wedge
GeoEP Software and method, for surface configurations such as http://geoadvanced.com/
Services level, ascending and/or descending or
stepped surfaces.
Simulation of steady-state groundwater flow
GGU 3D
GGU Soft in three-dimensional groundwater systems https://www.ggu-software.com/
SSFLOW
using finite element methods.
Analysis of transient groundwater flow using
GGU 3D the finite element method based on a 3-D
GGU Soft https://www.ggu-software.com/
Transient groundwater system analyzed using
GGU 3D SSFLOW.
Bored and driven pile calculations and
GGU-Axpile GGU Soft https://www.ggu-software.com/
graphical representation of results.
Analysis of 1-D consolidation processes in
GGU single-layered systems (analytical), multi-
GGU Soft https://www.ggu-software.com/
Consolidate layered systems (numerical), and single- or
multi-layered systems with vertical drains.
Analysis of plane and axis-symmetrical
GGU Elastic GGU Soft deformation using the finite element https://www.ggu-software.com/
method.
Analysis of retaining walls based on the
Recommendations of the German Working
GGU Retain GGU Soft https://www.ggu-software.com/
Group for Excavations and for Waterfront
Structures (EAB + EAU).
Settlement analysis of triangular and
GGU Settle GGU Soft rectangular foundations, including mutual https://www.ggu-software.com/
influence of neighboring foundations.
Analysis of elastically-supported slabs
based on the modulus of subgrade reaction
GGU Slab GGU Soft https://www.ggu-software.com/
and constrained modulus methods using
the finite element method.
Slope stability analysis and analysis of soil
nailing and reinforced earth walls. Nailing
GGU Stability GGU Soft can consist of anchors, soil nails, https://www.ggu-software.com/
geosynthetics (reinforced earth), or injection
piles.
Analysis of diaphragm wall stability in
GGU Trench GGU Soft https://www.ggu-software.com/
accordance with DIN 4126
GGU-Underpin GGU Soft Analysis and design of underpinning. https://www.ggu-software.com/
Gookin
BorinGS Creation and management of boring logs. http://www.gookinsoftware.com/
Software
Determination of the rate and magnitude of
GWP Geo consolidation of soil slurries, such as mine
FSCONSOL http://www.fsconsol.com/
software Inc. tailings, deltaic deposits, and other soft
soils.
Inducta Pty.
FTG Design of pad and strip footings. https://www.inducta.com.au/
Ltd.
Innovative Analysis of single pile behavior under axial
Geotechnics loading applied at the pile head for both
PileAXL https://www.pilegroups.com/
Pty Ltd, onshore and offshore engineering
Australia problems.
Innovative
Geotechnics Deep foundation analysis and design for
PileSuite https://www.pilegroups.com/
Pty Ltd, both onshore and offshore projects.
Australia
507
UFC 3-220-10
1 February 2022
Innovative Finite element simulation of deformations
Geotechnics and loads of pile groups subject to general
PileGroup https://www.pilegroups.com/
Pty Ltd, 3-D loading, such as axial and lateral forces
Australia and moments applied on the pile caps.
Finite-element simulation of laterally loaded
Innovative
piles (single piles mainly under lateral
Geotechnics
PileLAT loading) based on automatically generated https://www.pilegroups.com/
Pty Ltd,
nonlinear p-y curves for various soil and
Australia
rock types.
Innovative Prediction of settlement for piles socketed
Geotechnics into rock under compressive axial loading
PileROC https://www.pilegroups.com/
Pty Ltd, and estimates of ultimate and factored axial
Australia capacities for a range of socket lengths.
Interstudio Analysis of stratified slopes in the presence
Geo Tec B http://en.interstudio.net/
S.r.1 of water and loads.
3-D simulation for advanced geotechnical
Itasca
analysis of soil, rock, ground water,
3DEC Consulting https://www.itascacg.com/
structural support, and masonry using the
Group
distinct element method.
Itasca 2-D finite difference simulation for advanced
FLAC Consulting geotechnical analysis of soil, rock, https://www.itascacg.com/
Group groundwater, and ground support.
Itasca 3-D finite difference simulation for advanced
FLAC3D Consulting geotechnical analysis of soil, rock, https://www.itascacg.com/
Group groundwater, and ground support.
Itasca
Distinct Element Method (DEM) for
PFC Consulting https://www.itascacg.com/
advanced, fast multi-physics simulation.
Group
2-D simulation of the quasi-static or
Itasca dynamic response to loading of media
UDEC Consulting containing multiple, intersecting joint https://www.itascacg.com/
Group structures using the distinct element
method.
Itasca
FLAC/Slope Consulting Slope stability analysis. https://www.itascacg.com/
Group
Simulation of stability and deformation using
CESAR- LPCP itech-soft http://www.cesar-lcpc.com/
the finite element method in 2-D and 3-D.
Design of concrete or masonry leaning
Lean Wall Javasoft https://javasoft-softwares.com/
walls.
Design of concrete or masonry retaining
Retain Wall Javasoft https://javasoft-softwares.com/
walls.
Key wall Design and analysis of gravity walls and soil
Retaining reinforced wall sections for all Keystone
Key Wall Pro http://keystonewalls.com/
wall system structural units and most common soil
Inc reinforcement materials.
Geotechnical knowledge management
HoleBaseSI Keynetix Ltd system for inclusion of geotechnical data https://www.keynetix.com/
within the BIM process.
External stability analysis of reinforced
concrete cantilever walls (sliding,
Twall Design LG Soft https://www.dec.uc.pt/
overturning and bearing capacity) under
both static and seismic conditions.
Geotechnical stability analysis using the
Limit state LimitSTATE
limit state approach to determine the critical http://www.limitstate.com/
GEO Ltd
failure mechanism.
508
UFC 3-220-10
1 February 2022
Finite element simulation of deep
foundations, excavations, complex tunnel
Midas GTS MIDAS IT systems, seepage, consolidation, http://midasgtsnx.com/
embankments, dynamic conditions, and
slope stability analysis.
4D
Mira Quantitative forecasting of geotechnical
Geotechnical
Geoscience hazard for design or real time monitoring http://www.mirageoscience.com/
Hazard
Ltd. applications.
assesment
Limit equilibrium slope stability analysis of
Mitre
existing natural slopes, unreinforced man-
GSLOPE Software http://www.mitresoftware.com/
made slopes, or slopes with soil
Cooperation
reinforcement.
Mitre
GTILT Software Management of slope inclinometer data. http://www.mitresoftware.com/
Cooperation
Design and verification of foundation of
EDIPLIN Newsoft SAS https://www.newsoft-eng.it/
reinforced concrete poles.
NISEE -
Analysis of earthquake generation and
University of
FEQDrain dissipation of pore water pressure in https://nisee.berkeley.edu/
California,
layered sand deposits with vertical drains.
Berkeley
Novo Tech http://www.novotechsoftware.co
NovoCPT CPT interpretation.
Software Inc. m/
Novoformula Geotechnical correlations.
Software Inc. m/
NovoLiq Soil liquefaction analysis.
Software Inc. m/
Vislog 3-D soil profile visualization.
Software Inc. m/
Frew Oasys Ltd Embedded retaining wall analysis. https://www.oasys-software.com/
Greta Oasys Ltd Stability analysis for gravity retaining walls. https://www.oasys-software.com/
Soil settlement calculation and
PDisp Oasys Ltd https://www.oasys-software.com/
displacement analysis.
Calculation of load capacity and settlement
Piles Oasys Ltd https://www.oasys-software.com/
for single piles.
2-D finite element simulation in plane
Safe Oasys Ltd https://www.oasys-software.com/
stress, plane strain, or axial symmetry.
Siren Oasys Ltd Seismic site response analysis. https://www.oasys-software.com/
Slope Oasys Ltd 2-D slope stability analysis. https://www.oasys-software.com/
Prediction of ground movement, settlement,
Xdisp Oasys Ltd and assessment of building and utility https://www.oasys-software.com/
damage.
Alp Oasys Ltd Analysis of laterally loaded piles. https://www.oasys-software.com/
Creation of detailed velocity models
Optim from surface refraction array data using a
Seisopt 2D http://www.optimsoftware.com/
Software proprietary simulated annealing optimization
algorithm.
Implementation of the Refraction
Optim
Seisopt Remi Microtremor (ReMi) method to measure the http://www.optimsoftware.com/
Software
in situ shear wave velocity profile.
Optum 2-D finite element simulation for
Optum G2 Comput. geotechnical stability and deformation https://optumce.com/
Engineering analysis in plane strain or axisymmetry.
Optum 3-D finite element simulation for
Optum G3 Comput. geotechnical stability and deformation https://optumce.com/
Engineering analysis in plane strain or axisymmetry.
509
UFC 3-220-10
1 February 2022
SPW911 Pile Buck Sheet pile design. http://www.pilebuck.com/
Pile 1-D wave equation analysis to simulate
GRLWEAP Dynamics motions and forces in a pile when driven by https://www.pile.com/
Inc. either an impact or vibratory hammer.
Pile
Pile Driving Dynamic load testing and pile driving
Dynamics https://www.pile.com/
Analyzer monitoring.
Inc.
Simulation of static load test in compression
Pile
and tension, prediction of load displacement
CAPWAP Dynamics https://www.pile.com/
behavior, and determination of stresses at
Inc.
each depth along the pile.
Quality control or assessment of drilled
shafts/bored piles, auger cast in place
Thermal Pile
(ACIP)/continuous flight auger (CFA) or
Integrity Dynamics https://www.pile.com/
drilled displacement piles, slurry walls,
Profiler Inc.
barrettes, soil nails, and jet grouted
columns.
Pile 3-D tomography imaging tool for analyzing
PDI TOMO 3D Dynamics wave speeds to yield a wave speed of https://www.pile.com/
Inc. entire shaft volume.
Pile
PDA-DLT Dynamics Dynamic load testing for drilled shafts. https://www.pile.com/
Inc.
2-D finite element simulation with add-ons
Plaxis 2D Plaxis for ground water flow, dynamic loading, and https://www.plaxis.com/
thermal analysis of soils and rocks.
3-D finite element simulation with add-ons
Plaxis 3D Plaxis for ground water flow, and dynamic loading https://www.plaxis.com/
of soils and rocks.
Geo Program Presta Shop Complete geotechnical analysis. http://www.programgeo.it/
Various books Prof. Arnold Analysis of sheet pile walls in layered soils,
and software Verruijt (Deflt slope stability, piles, groundwater flow, and
http://geo.verruijt.net/
for soil University of finite element simulation of steady and non-
mechanics Technology) steady groundwater flow.
Prokon Structural design that is useful for slope
PROKON Software stability analysis, rock stability and capacity https://www.prokon.com/
Consultants analyses and pile capacity analysis.
Q System
Geotech Analysis and design of foundations and
Engineering http://qsystemsengineering.net/
Masters piles.
LLC
Retainpro
Software div. Design and analysis of earth retaining
Retain Pro 10 https://retainpro.com/
ENERCAL, structures.
Inc.
Evaluation of the stability of surface or
CPillar Rocscience underground crown pillars, and laminated https://www.rocscience.com/
roof beds.
Dips Rocscience Stereographic projection. https://www.rocscience.com/
Stress analysis and data visualization tool
Examine Rocscience https://www.rocscience.com/
for underground excavations in rock.
Analysis of rock and soil strength data, and
RocData Rocscience determination of strength envelopes and https://www.rocscience.com/
other physical parameters.
2-D statistical analysis to assist with
RocFall Rocscience https://www.rocscience.com/
assessment of slopes at risk for rock falls.
Planar rock slope stability analysis and
RocPlane Rocscience https://www.rocscience.com/
design.
Estimation of support requirements of
RocSupport Rocscience https://www.rocscience.com/
tunnels in weak rock.
510
UFC 3-220-10
1 February 2022
Toppling analysis and support design for
RocTopple Rocscience https://www.rocscience.com/
rock.
RS2 Rocscience 2-D finite element simulation. https://www.rocscience.com/
RS3 Rocscience 3-D finite element simulation. https://www.rocscience.com/
RSPile Rocscience Pile analysis. https://www.rocscience.com/
Settle Rocscience 3-D soil settlement analysis. https://www.rocscience.com/
2D slope stability analysis using limit
Slide2 Rocscience equilibrium method and finite element https://www.rocscience.com/
seepage analysis.
Slide3 Rocscience 3-D slope stability analysis https://www.rocscience.com/
Evaluation of the geometry and stability of
SWedge Rocscience https://www.rocscience.com/
surface wedges in rock slopes.
3-D stability analysis and visualization
program for underground excavations in
UnWedge Rocscience https://www.rocscience.com/
rock containing intersecting structural
discontinuities.
ELK Sharper Geo Analysis of in situ tests and slope stability. http://www.sharpergeo.com/
3-D conceptual modeler and visualization
Soil Vision
SVDesigner tool for the geotechnical https://soilvision.com/
system Ltd
and hydrogeological fields.
Soil Vision Dynamic analysis by the finite element
SVseismic https://soilvision.com/
system Ltd direct time step-by-step integration method.
Soil Vision
SVslope 3-D slope stability analysis. https://soilvision.com/
system Ltd
Estimation and mathematical representation
Soil Vision
SVsoils of soil constitutive models for subsequent https://soilvision.com/
system Ltd
numerical modeling.
Soil Vision 1-D, 2-D, and 3-D finite element simulation
SVflux https://soilvision.com/
system Ltd of groundwater.
Determination of the stress state and
Soil Vision deformation of soils under various loading
SVsolids https://soilvision.com/
system Ltd conditions and solving stress-deformation
problems.
TAGA
TSLOPE 3D Engineering 3-D slope stability analysis. https://tagasoft.com/
Software Ltd.
Technology
CASTeR Development Generation of soil test reports. http://www.tdcindia.com/
Center
Technology Analysis of slope stability, bearing capacity,
GTeCS Development pile capacity, settlement, and under-reamed http://www.tdcindia.com/
Center pile capacity.
Tensar
TensarSoils International Analysis and design of retaining walls. https://www.tensarcorp.com/
Corporation
Tensar Calculation of bearing capacity and
Dimensions International projected settlement beneath shallow https://www.tensarcorp.com/
Corporation foundations.
Design of shallow, deep, and raft
Foxta V3 Terrasol https://www.terrasol.fr/
foundations.
Design of retaining walls using the
subgrade reaction method, including
K-REA V4 Terrasol https://www.terrasol.fr/
diaphragm walls, sheetpile walls, and
soldier pile walls.
Semi-automatic processing of geotechnical
Straticad Terrasol data within drawings and their display in 2-D https://www.terrasol.fr/
and 3-D.
511
UFC 3-220-10
1 February 2022
Slope stability analysis, including stability of
geotechnical structures, reinforcement,
Talren V5 Terrasol https://www.terrasol.fr/
natural slopes, cut or fill slopes, earth dams,
and dikes.
Unipile 5.0 UniSoft GS Analysis of piles and pile groups. https://www.unisoftgs.com/
Stress and settlement calculations involving
Unisettle 4.0 UniSoft GS complex load combinations and site https://www.unisoftgs.com/
conditions.
University of http://www.ce.utexas.edu/prof/wri
UTEXAS Texas at ght/UTEXASED4/UTEXASED4%
equilibrium method.
Austin 20Home.htm
Selection of suites of earthquake ground
University of motions from a library of ground motions
https://github.com/arkottke/sigma
SigmaSpectra Texas at such that the median of the suite matches a
spectra
Austin target response spectrum at all defined
periods.
University of 1-D linear-elastic and equivalent-linear site
Strata Texas at response analyses using time series or https://github.com/arkottke/strata
Austin random vibration theory ground motions.
University of
Sliding-block analyses to evaluate seismic
SLAMMER Texas at https://pubs.usgs.gov/tm/12b1/
slope performance.
Austin
Simulation and prediction of groundwater
conditions and groundwater/surface-water https://water.usgs.gov/ogw/modfl
MODFLOW USGS
interactions. MODFLOW is the USGS's ow/
modular hydrologic model.
Simulation of multi-phase groundwater flow
HYDRO- https://volcanoes.usgs.gov/softw
USGS and associated thermal energy transport in
THERM are/hydrotherm/
three dimensions.
Wutec
Finite element simulation for quasi-3D
Geotechnical
nonlinear dynamic analyses of single piles
VERSAT-P3D International, http://www.wutecgeo.com/
and pile groups in the frequency and time
B.C.,
domains.
Canada
Software package (VERSAT-2D Processor,
VERSAT-S2D and VERSAT-D2D) for 2-D
Wutec
finite element simulation of stresses,
Geotechnical
deformations, and soil-structure interactions
VERSAT-2D International, http://www.wutecgeo.com/
for static loading and dynamic analyses of
B.C.,
earth structures subjected to dynamic loads
Canada
from earthquakes, machine vibration,
waves, or ice action.
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APPENDIX C. SYMBOLS USED IN GEOTECHNICAL ENGINEERING
One potentially confusing aspect of geotechnical engineering is the lack of

standardization used for common engineering parameters. Different symbols were
adopted at the various U.S. and overseas universities, as well as organizations such as
NGI, USBR, and USACE, who were involved in the early years of soil mechanics. As
an example, when presenting the results of the compression curve of a conventional
consolidation test, the symbols used for the x-axis (vertical effective stress) include: p ,
p ' , σ 'v , σ v , and others. The purpose of this Appendix is not to offer suggestions for
standardization, but to provide a listing of the different symbols that have been used
historically in the geotechnical literature to be used as a cross-reference when
consulting old figures, papers, and texts.
Table C-1 Symbols Used in Geotechnical Engineering
Symbol Description
a Isotropic transformation factor for flow nets
a CPT net area ratio used for pore pressure corrections
a Acceleration
a Strength parameter used with a power function for nonlinear failure envelope
a Attraction
A Cross sectional area of the flow region perpendicular to the flow direction
A Skempton pore pressure parameter
A Skempton pore pressure parameter

ACU Anisotropically consolidated-undrained triaxial test
Af Skempton pore pressure parameter at failure
av Coefficient of compressibility
b Strength parameter used with a power function for nonlinear failure envelope
B Width of a foundation, loaded area, or tunnel
B Skempton’s pore pressure parameter
B Skempton’s pore pressure parameter

Bc Diameter of a flexible pipe
Bd Width of trench in pipe loading calculations
Bt Bulk modulus of soil

c Total stress cohesion intercept (sometimes undrained shear strength)
c' Effective stress cohesion intercept.
C Number of surfaces on which pullout resistance is mobilized
Cαε Modified secondary compression index or secondary compression ratio
CAU Anisotropically consolidated undrained triaxial test
CBR California Bearing Ratio
CBRsoaked Soaked California Bearing Ratio
CBRunsoaked Unsoaked California Bearing Ratio
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Symbol Description
Cc Coefficient of curvature for grain-size distribution curve.
Cc* Intrinsic compression index
Ccε Modified compression index
cCU Total stress cohesion intercept from CU triaxial test
Cd Load coefficient in pipe loading calculations

CF Clay-sized fraction
ch Coefficient of consolidation in horizontal direction
CIU Isotropically consolidated-undrained triaxial test
Cr ε Modified recompression index
CRR Cyclic resistance ratio

Cs Swelling index, often used as a synonym for recompression index
CSR Cyclic stress ratio
Ct Creep factor for coarse-grained settlement methods
Cu Coefficient of uniformity for grain-size distribution curve
C 'u Linear coefficient of uniformity (geotextile design)
cv Coefficient of consolidation in vertical direction
Cε c Modified compression index
Cε r Modified recompression index
Cεα Modified coefficient of secondary compression

d Distance between the loaded points
d Y-intercept of the failure envelope (Kf-line) in MIT stress path space (p vs q)
d' Y-intercept of the failure envelope (Kf-line) in MIT stress path space (p' vs q)
dc Effective drainage diameter
dw Equivalent diameter of well or PVD

D Diameter of the lab or field vane
D Diameter
D Outer diameter of pipe
D Foundation embedment
D Damping ratio
Particle-size diameter corresponding to 5% passing on the cumulative particle-size
D5
distribution curve
D10
distribution curve
D15
distribution curve
D30
distribution curve
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Symbol Description
D60
distribution curve
DCP Dynamic cone penetrometer or DCP penetrometer index
De equivalent core diameter
Dr Relative density
DSS Direct simple shear test
Particle size for which X% of the soil is finer for linearized particle distribution (geotextile
D 'x
design)
Dx B Particle size for which X% of the soil is finer for a base soil
Dx F Particle size for which X% of the soil is finer for a filter material
e Void ratio
*
*
ef Final void ratio or void ratio at failure
eL Void ratio at a water content equal to the liquid limit
emax Maximum index void ratio
emin Minimum index void ratio
E Elastic modulus our Young's Modulus

E Compactive effort index
E' Equivalent modulus
Ea Active earth pressure force
ED Dilatometer modulus
Ei Initial tangent modulus

EI Expansion index
Em Modulus of elasticity of mat
EP Pressuremeter modulus
Es Modulus of elasticity of soil

ESP Effective stress path (MIT p' vs q stress path space)
Eu Undrained Young’s modulus
F Factor of safety
Percentage passing a No. 200 (75 μm) sieve (only considering the particles passing a 3-
F inch sieve)
F size correction factor
F* Pullout resistance factor
FC Fine contents
fi Fraction of particles between two adjacent sieve sizes (Kozeny-Carman equation)
Fn Radial drainage factor related to drain spacing
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Symbol Description
Fr Radial drainage factor related to well resistance
Fr Factor of safety for geosynthetic strength

FR Cone penetration test friction ratio
fs Cone penetrometer friction sleeve resistance
Fs Radial drainage factor related to soil disturbance (smear)

FS Factor of safety
FS g Factor of safety for geotextile permeability
Fw Factor of safety against wedge failure

G Shear modulus
GC Gravel Content
GI group index
Gs Specific gravity of solids
GSI Geological Strength Index
Gu Undrained shear modulus
H Height
H Depth of soil cover or vertical distance between ground surface and tunnel roof
H Initial thickness in settlement
H dr Drainage path length
Hi Thickness of each soil layer (may be listed without subscript)
Hi Average height of the slice in slope stability analysis
Hi Initial height of the test specimen
hl Head loss across flow region
hp Pressure head
ht Total hydraulic head
Ht Tunnel height
Ht Total thickness of transformed soil system
hv Velocity head
hz Elevation head
I Influence factor for change in stress calculations
I1 First stress invariant
Ic Soil index
ICU Isotropically consolidated-undrained triaxial test
ID Dilatometer material index
Is Uncorrected point load strength index
I s( 50 ) Size corrected point load strength index
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Symbol Description
Iv Void index
I v , ICL Void index for the intrinsic compression line
I v , SCL Void index for the sedimentation compression line
Iz Schmertmann strain influence factor
I zp Schmertmann peak influence factor

k Hydraulic conductivity or permeability
K Wedge factor
K Bulk modulus
K0 Coefficient of lateral earth pressure at rest
K0 -line Line through p’ and q (MIT) for at-rest conditions
Ka Active earth pressure coefficient
Kb Bulk modulus parameter for Duncan-Chang model
Kc Anisotropic consolidation stress ratio = σ '1,con σ '3,con

KD Dilatometer horizontal stress index
K f -line Failure envelope in MIT stress path space (p'f vs. qf)
kg Hydraulic conductivity of geotextile across plane of fabric
kh Hydraulic conductivity in horizontal direction

Km Mat stiffness factor
Kp Passive earth pressure coefficient
ks Coefficient of subgrade reaction
ks Hydraulic conductivity of the disturbed zone
K ur Unload-reload modulus parameter Duncan-Chang model
kv Hydraulic conductivity in horizontal direction

l Distance between two points along a structure
L Length
L Longest dimension of a foundation or loaded area
L Length of flow path
Le Length of reinforcement embedded behind the trial failure surface
LI or I L Liquidity index
LIR Load Increment Ratio for consolidation test
LL Liquid limit
Lm maximum distance water must flow through a vertical drain
m Modulus number
MARV Minimum average roll value used for various properties of geosynthetics
M ds Constrained modulus
MSE Mechanically stabilized earth
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Symbol Description
mv Coefficient of volume compressibility
n Porosity
n Vertical drain spacing ratio
N Standard Penetration Test blow count (blows/ft). Often assumed to be N60
N' Average Standard Penetration Test value
N1 Standard Penetration Test blow count normalized to overburden pressure of 1 tsf
Standard Penetration Test blow count corrected for 60% of hammer energy and normalized
N1,60
to an overburden pressure of 1 tsf.
SPT blow count corrected for 60% of hammer energy and fines content, and normalized to
N1,60 ,cs
an overburden pressure of 1 tsf.
N60 SPT blow count corrected for 60% of hammer energy
Nc Bearing capacity factor

NC Normally consolidated
N crit Undrained stability factor
Nd Number of equipotential (head) drops in the flow net
Nf Number of flow channels in the flow net
Nk Bearing capacity factor used for reduction of CPT data in fine-grained soil
N kt Bearing capacity factor used for reduction of CPT data in fine-grained soil
Nq Bearing capacity factor
N 'silty Standard Penetration Test blow count for saturated silty sands
O95 Geotextile apparent opening size

OC Overconsolidated
p MIT stress path parameter = (σ1 + σ3)/2 or (σv + σh)/2
p' MIT stress path parameter = (σ'1 + σ'3)/2 or (σ'v + σ'h)/2
p0 Pressure required to initiate movement of the dilatometer
p0 Pressuremeter liftoff pressure
pc Tunnel air pressure
Pc Maximum past pressure or preconsolidation pressure
pf MIT stress path parameter p at failure

Inflection point in pressuremeter curve assumed to delineate the change from pseudo
pf
elastic to plastic response and the point where creep may be expected
p 'f MIT stress path parameter p' at failure
PI Plasticity index
Pressuremeter limit pressure where the curve becomes asymptotic on a pressure versus
pL
volume curve
PL Plastic limit
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Symbol Description
Pp Maximum past pressure or preconsolidation pressure
P 'p Maximum past pressure or preconsolidation pressure

Pressuremeter yield point during the reloading portion of an unload–reload cycle where
pr
recompression ends and the soil reinitiates plastic shearing
Pr Geosynthetic reinforcement’s resistance to pullout
PSR Principal stress ratio
pu Pressuremeter minimum pressure during unloading during the unload–reload cycle
q Volumetric flow rate
q MIT stress path parameter = (σ1 - σ3)/2 or (σv - σh)/2
Q Rock tunneling quality index
Q Quantity of flow
Q Unconsolidated-undrained (UU) triaxial test
q0 Applied pressure or load
q0 Applied stress at the base of the foundation or structure
q0 − net Net vertical stress applied by the structure
qc Cone penetrometer tip resistance or cone bearing (not corrected for pore pressure effects)
Cone penetrometer tip resistance or cone bearing normalized to an overburden pressure of
qc1
1 tsf
qd Dynamic cone resistance
qf Applied stress following removal of surcharge
qf MIT stress path parameter q at failure
qs Surcharge load
qt Cone penetrometer tip resistance corrected for pore pressure effects
qu Unconfined compressive strength
qw Discharge capacity of the drain

r Horizontal distance from centerline of a foundation
R Radius of influence in well design
R Correction factor for overconsolidated static CTP cone tip resistance
R Isotropically consolidated undrained triaxial test with pore water pressure measurements.
Isotropically consolidated undrained triaxial test without pore water pressure
R measurements.
RC Relative compaction
Rf Reduction factor for Duncan-Chang model
RFCR Reduction factor for creep
RFD Reduction factor for durability
RFID Reduction factor for installation damage

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Symbol Description
RQD Rock Quality Designation
ru Pore pressure coefficient
s Ratio of the disturbed zone diameter to the diameter of the drain
s Settlement
s Shear strength
S Degree of saturation
S Surface area factor for grain shape (Kozeny-Carman equation)
S Seepage force
sc Primary consolidation settlement
SRF Strength reduction factor
ss Secondary compression settlement
ssu Undrained steady state shear strength
St Sensitivity
St , fv Sensitivity measured using field vane shear apparatus
su Undrained shear strength for a φ = 0 envelope = (σ1f - σ3f)/2
su , fv Undrained shear strength determined using field vane apparatus
sur , fv Remolded undrained shear strength determined using field vane apparatus
t Thickness
t Stand up time for tunneling in raveling soils
t Time after start of consolidation
t Time
T Elapsed time between excavation and completion of permanent structure
T Temperature
T Time factor in consolidation theory
t50 Time for 50% consolidation to be achieved
t50 Time for 90% consolidation to be achieved
Tal Geosynthetic’s long-term tensile strength
tg Geotextile thickness
tm Thickness of mat
Tmax Maximum net torque for vane shear test
tp Time required to finish primary consolidation
Tr Time factor for radial consolidation
Tres Residual torque reading for vane shear test

TSP Total stress path (MIT)
(T-us)SP Total stress path – static pore water pressure (MIT)
TULT Ultimate tensile strength of the geosynthetic based on the MARV
Tv Time factor for vertical drainage

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Symbol Description
U Uplift force applied by water at the failure plane
U Degree of consolidation
u0 Initial pore water pressure
Cone penetrometer pore water pressure for sensing element located directly behind cone
u2
tip
ua Pore air pressure
Uc Combined degree of consolidation

UC Unconfined compression test
U fs Degree of consolidation following surcharge application
Ur Degree of radial consolidation

USR Undrained strength ratio = su/σ'v
USRNC Undrained strength ratio for normally consolidated conditions
UU Unconsolidated-undrained triaxial test
ux Excess pore water pressure
Uz Degree of compression
Uz Average degree of consolidation

v Specific volume = 1 + e
V Total volume
V Total volume (phase relationships)
V0 Initial calculated volume within the uninflated membrane for pressuremeter
Va Volume of air (phase relationships)
vd Discharge velocity
vs Seepage velocity
Vs Volume of solids (phase relationship)
Vs Shear wave velocity
Vs1 Normalized shear wave velocity
Vv Volume of voids (phase relationship)
Vw Volume of water (phase relationships)

w Water content (gravimetric)
W Width of the system perpendicular to the page
W Weight
W Total weight (phase relationships)
w0 Initial water content
Wc Flexible pipe load
Wd Rigid pipe load
wf Final water content
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Symbol Description
Wi Weight of each slice (limit equilibrium slope stability analysis)
wL Liquid limit
wopt Optimum water content
wP Plastic limit
Wp Prism load on pipe
Ws Weight of solids (phase relationships)
WT Total weight of sample (phase relationships)
Ww Weight of water (phase relationships)

y Height of the flow region
z Depth along vertical drain
z Elevation of a point of interest above the elevation datum
z Depth below the soil layer
zcrit Critical depth for unsupported shafts in clay soils
zi Layer thickness for settlement calculations
zp Depth below an applied load
α Angle between the major principal plane and the plane of interest
α Settlement correction factor
α Dip direction or dip azimuth
α Scale correction factor to account for nonlinear stress reduction
α Slope of the failure line (Kf) in MIT p - q space
α’ Slope of the failure line (Kf) in MIT p’ - q space
βα Empirical or semi-empirical coefficient relating k to Dα
δ Effective soil-geosynthetic interface friction angle
∆e Change in void ratio
∆H Change in layer thickness
∆H Change in height
∆hL ∆hL Total head loss for one equipotential drop on a flow net
δL Angular distortion
∆L Deflection ratio
δ max Differential settlement
∆q p Change in cone tip resistance
∆σ Change in applied stress
∆σ 'v Change in vertical effective stress
∆σ d Change in deviatoric stress or change in principal stress difference = ∆σ1 - ∆σ3
∆σ v Change in total vertical stress
∆u Change in pore water pressure
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Symbol Description
εa Vertical or axial strain
ε crit Critical strain for structural distress
ε Strain rate
εh Horizontal strain
εr Radial strain
ε vol Volumetric strain
εv Volumetric strain
εv Vertical strain
φ Total stress friction angle
φUU or φU Total stress friction angle from UU triaxial test (S < 100%)
φ Effective stress friction angle
φCU Total stress friction angle from CU triaxial test

φ 'FS Fully softened friction angle
φ FS Fully softened friction angle
φ 'RES Residual friction angle
φ RES Residual friction angle
φ 'R Residual friction angle
φR Residual friction angle
φ 'SEC Effective stress secant friction angle (stress dependent)
φ SEC Effective stress secant friction angle (stress dependent)
γ Unit weight
γ Shear strain
γ' Effective unit weight
γb Buoyant unit weight
γd Dry unit weight
γ d − max Maximum dry unit weight
γm Moist unit weight
γ SAT Saturated unit weight
γT Total, wet, or moist unit weight
γ Shear strain rate
λ Ratio of the circumferential stress to the vertical stress in circular openings
µ Coefficient of friction
µ' Coefficient of friction for trench backfill
µ0 Influence factor associated with embedment of load
µ1 Influence factor associated with geometry and Poisson’s ratio
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Symbol Description
µR Vane correction factor
νm Poisson’s ratio of mat
θ Volumetric moisture content
σ Total normal stress
σ' Effective normal stress
σ Effective normal stress

σ1 Total major principal stress
σ '1 Effective major principal stress
σ1 Effective major principal stress

σ '1,con Effective major consolidation stress
σ 1,con Effective major consolidation stress
σ2 Total intermediate principal stress

σ '2 Effective intermediate principal stress
σ2 Effective intermediate principal stress

σ3 Total minor principal stress
σ '3 Effective minor principal stress
σ3 Effective minor principal stress

σ '3 ,con Effective minor consolidation stress
σ 3 ,con Effective minor consolidation stress
σ 'c Consolidation stress
σc Consolidation stress
σ cell Cell pressure for triaxial test
σd Principal stress difference or deviatoric stress
σ1 − σ 3 Principal stress difference or deviatoric stress
σ ' fc Effective normal stress on the failure plane during consolidation
σ fc Effective normal stress on the failure plane during consolidation
σ ff Total normal stress on failure plane at failure
σ ' ff Effective normal stress on failure plane at failure
σh Total horizontal stress

σ 'h Effective horizontal stress
σh Effective horizontal stress

σ 'm Mean effective stress
σm Mean effective stress

σ 'N Effective normal stress on failure surface
σN Effective normal stress on failure surface
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Symbol Description
σ 'p Maximum past pressure or preconsolidation stress
σp Maximum past pressure or preconsolidation stress
σ 'crit Critical confining stress

σ ' ps Effective stress after perfect sampling
σ 'p Effective stress after perfect sampling
σt Interior tunnel pressure from compressed air or breasting

σv Total vertical stress
σ 'v Vertical effective stress
σv Vertical effective stress

σ v0 Initial vertical total stress
σ 'z Vertical effective stress
σ z0 initial geostatic vertical total stress
σ 'z 0 Initial or in situ vertical effective stress
σ 'zp Initial vertical effective stress at depth of Schmertmann peak influence factor
τ Shear stress
τ cyc Applied peak cyclic shear stress
τf Shear stress at failure
τ ff Shear stress on the failure surface at failure
τ eq Shear stress required for equilibrium

ω Tilt angle due to differential settlement
ψ Matric suction
Ψ Dip
ψg Geotextile permittivity, provided by manufacturers or from testing (ASTM D4491)
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Table C-2 Acronyms and Abbreviations
Term Definition
2-D or 2D Two dimensional
3-D or 3D Three dimensional
AASHTO American Association of State Highway and Transportation Officials
ACU Anisotropically consolidated undrained
AMTS Automated total station
AR Augmented reality
BLM U.S. Bureau of Land Management
BPT Becker penetration test
CB Cement-bentonite
CD Consolidated drained
CU Consolidated undrained
CK0U K0 consolidated undrained
CPMT Cone pressuremeter
CPT Cone penetration test
CPTu Piezocone test
CRS Constant rate of strain (consolidation test)
CYCDSS Cyclic direct simple shear
DCP Dynamic cone penetration
DMT Flat plate dilatometer test
DOT Department of Transportation
DPI Dynamic cone penetration index
DPT “Dutch” cone penetrometer test
DSS Direct simple shear
DST Direct shear test
DTM Digital terrain model
EDG Electrical density gauge
EIS Electrical impedance spectroscopy
EROS Earth Resources Observation System
FDM Finite difference method
FEA or FEM Finite element analysis or method
FEMA Federal Emergency Management Agency
FERC Federal Energy Regulatory Commission
FHWA Federal Highway Administration
GIS Geographic information system
GPS Global positioning system
HCMM Heat Capacity Mapping Mission
ICL Intrinsic compression line for remolded clays (Burland 1990)
ICOLD International Committee on Large Dams
ICU Isotropically consolidated undrained
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Term Definition
ID Inside diameter
ISO International Organization for Standardization
ISRM International Society for Rock Mechanics
IST Impact soil tester
LEL Lower explosive limit
LFD Lightweight falling deflectometer
LIDAR Light detection and ranging
LIR Load increment ratio
LPT Large penetration test
LVDT Linear pvariable displacement transducer
M-DI Moisture-density indicator
MEMS Micro-electro-mechanical
MSE Mechanically stabilized earth
MSW Municipal solid waste

NAIP National Agriculture Imagery Program
NCIC National Information Center
NCHRP National Cooperative Highway Research Program
NCRS National Resources Conservation Service

NDG Nuclear density gauge
NFS Not frost-susceptible
NP Nonplastic
NRC Nuclear Regulatory Commission
OD Outside diameter
OSHA Occupational Safety and Health Administration
PFS Possibly frost-susceptible
PLT Plate load test
PMT Pressuremeter test
RTD Resistance temperature device
SAR Synthetic aperture radar
SB Soil-bentonite
SBPMT Self-boring pressuremeter
SBT Soil behavior type
SCB Soil-cement-bentonite
SCL Sedimentation compression line for natural clays (Burland 1990)
SCPTu Seismic piezocone test
SDG Soil density gauge
SLAR Side-looking airborne radar
SPT Standard Penetration Test
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Term Definition
SS Surface stiffness
TVA Tennessee Valley Authority
UAV Unmanned aerial vehicle
UEL Upper explosive limit
USACE United States Army Corps of Engineers
USBR United Stated Bureau of Reclamation
USCS Unified Soil Classification System
USDA United State Department of Agriculture
USEPA United States Environmental Protection Agency
USFS United States Forest Service
USGS United States Geological Survey
UU Unconsolidated undrained
VST Vane shear test
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APPENDIX D. GLOSSARY
Active zone – The near-surface zone affected by seasonal variation in water content.
Also, zone within a soil mass subjected to active earth pressure conditions.
Activity of clay – The ratio of plasticity index to percent by weight of the total sample
that is smaller than 0.002 mm in grain size. This property can be correlated with the
type of clay mineral.
Adobe – Sandy clays and silts of medium plasticity usually found in the semiarid
regions of the southwestern United States. The name is also applied to some high
plasticity clays with high clay content and high swell and shrink potential usually found
in the western United States.
Aeolian soil – Material transported and deposited by wind.
Aquiclude – A relatively impervious rock or soil layer underlying or overlying an aquifer.
Aquifer – Relatively permeable rock or soil stratum that can store and easily transmit
water. Also used for the sand layer often found beneath levees in the lower Mississippi
Valley.
Alluvial soils – Materials transported and deposited by running water.
Anisotropic soil – A soil mass having different properties in different directions, often
referring to strength or permeability characteristics.
As-compacted – Condition of the soil after compaction is completed.
Azimuth – Is the angle of a feature measured from North at 0° in a spherical coordinate

system.
Baby poop – Very soft clay located just above limestone in karst. Frequently orange
and formed by dissolution.
Back-packing – Any material (commonly granular) that is used to fill the empty space
between the lagging of a wall system and a rock surface.
Backswamp – The prolonged accumulation of floodwater sediments in flood basins

bordering a river; materials are generally clays but tend to become siltier near the
riverbank
Balanced load – See Compensated foundation
Bank-run sand and gravel – Raw material excavated from a borrow pit, but not sorted
or separated into specific grades.
Bedding – Planes of dissimilar materials caused by deposition normally encountered in

sedimentary rocks.
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Bentonite – High plasticity clay consisting of mostly montmorillonite, resulting from the
weathering of volcanic ash mainly in the presence of water. It is normally hard when dry
but swells considerably when wet. This clay is commonly used with water as drilling
mud and as liner in landfills.
Black cotton soil – Black expansive soil commonly encountered in India. The name
originates from the fact that this soil is common in areas where the main crop is cotton.
Blocky – Adjective for soils that can be broken down into small angular lumps which
are difficult to break down further.
Blow sand – Wind-driven or drifted sands.
Blue marl – Name given to a bluish-green clay from the Miocene that can be found
along the fall line from Richmond, VA, into Maryland. This soil is considered to be
acidic, usually with a pH less than 4.0, which can affect water quality and prevent plant
or aquatic life.
Bog – Wetland covered with peat with a high water table that accumulates dead plants,
usually mosses, and mainly sphagnum. It is generally nutrient poor and acidic.
Boney ground – Ground containing significant amounts of large gravel, cobbles, and
boulders.
Borehole jack – An in situ test device used to estimate the deformability of rocks.
Equipment description and operating procedures are presented in ASTM D4971.
Boulder – Rock particles that have a greatest dimension of at least 12 inches.
Boulder clay – Geological term used to designate clays formed from glacial drift that
has not been subjected to the sorting action of water and therefore contains particles
from boulders to clay sizes. Boulder clays are also called tills.
Boundary condition – Physical parameters assigned to the edges or boundaries of the

domain in numerical analysis. Examples are constant total head boundaries in FE
seepage analysis or restrained displacement boundaries in FE stress analysis.
Breaker run – Crushed rock with large particles refers to large broken stone obtained
as part of quarrying or mining activities.
Buckshot – Term applied to clays of the southern and southwestern United States that
cracks into small, hard, relatively uniform sized lumps on drying. The lumps are similar
to the size of buckshot and the soil is very sticky when wet.
Bull’s liver – An inorganic silt or silty sand usually encountered in the New York City
area. The name Bull’s liver comes from its red color and jelly-like behavior when it is
subjected to vibration.
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Bull’s tallow or Bull tallow clay – Tan or gray high plasticity clay typically found in
relatively thin layers directly above partially weathered rock or rock in the Charlotte, NC,
area. This clay normally has high shrink and swelling potential.
Caliche – Sedimentary rock from arid and semiarid climate in which soil particles, such
as gravel, sand, clay, and silt, are cemented and coated by carbonate (often calcium or
magnesium carbonate). The level of cementation varies significantly within a deposit.
The soil has light coloration often exhibits light colored concretions of various sizes
depending on the level of development of the soil profile. The consistency of caliche
varies from soft rock to firm soil.
Capillary stresses – Pore water pressures less than atmospheric values produced by
surface tension of pore water acting on the meniscus formed in void spaces between
soil particles.
Channel fill – Deposits laid down in abandoned meander loops isolated when rivers
shorten their courses; composed primarily of clay. However, silty and sandy soils are
found at the upstream and downstream ends
Chip – Name given to crushed angular rock fragments smaller than a few centimeters.
Clay – Soil particles passing a No. 200 (75-μm) sieve that exhibit plasticity (putty-like
properties) within a range of water contents, and considerable strength when air dried.
For classification of clayey soils, refer to Section 1-3.3.
Clay size fraction – The portion of the soil which is finer than 0.002 mm. This is not a
viable measure of the plasticity of the material or its characteristics as a clay.
Coarse-grained soils – Soils that contain 50% or more particles retained on a No. 200
(75 μm) sieve.
Cobbles – Rock particles that pass through a 12-inch square opening sieve but are
retained on a 3-inch square opening sieve.
Coffee grounds – Soil formed from freshwater marshes that has been dry for decades
and has decomposed to the point that is black and inert with little to no plasticity. It is
black and granular even when wet.
Colluvial soils – Material transported and deposited by gravity, often found in the
vicinity of slopes.
Colluvium – Loose soil deposited at the bottom of a slope.
Compacted – Soil specimen formed by compaction in a mold at a given water content

and relative compaction usually referred to a given compaction standard.
Compensated foundation – Method used to support heavy structures over

compressible strata. In this approach, the weight of the structure is balanced,
completely or partially, by soil that is permanently excavated from the building footprint.
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Compression index – Parameter which quantifies the compressibility of normally
consolidated soil in one-dimensional compression. Normally, it is the log-linear slope of
the compression curve defined by void ratio (y-axis) and the logarithm of vertical
effective stress (x-axis).
Cone Penetration Test (CPT) – An in situ test that utilizes a standard cone-shaped
instrument that is pushed at a standard constant rate from the ground surface to obtain
a continuous record of the penetration resistance of the cone tip and the frictional
resistance of the soil acting on the friction sleeve of the probe. Testing is currently
conducted in accordance with ASTM D5778.
Consolidation tests – Tests in which the volume change of the soil is determined for a
change in applied stress, normally for one-dimensional compression.
Coquina – Soft, porous sedimentary rock, mainly limestone, composed largely of

shells, coral, and fossils cemented together, with particles averaging 0.079 in (2 mm) or
greater in size.
Compression curve – Curve relating the void ratio or strain to the effective stress
applied (usually in log scale).
Critical depth – The depth over which soil compression caused by changes in stress
contributes to significant surface settlement. The critical depth in fine-grained soils
corresponds to the depth at which the change in stress is less than 10% of the existing
vertical effective stress. In coarse-grained soils, the critical depth occurs when the
change in stress is less than 20% of the existing vertical effective stress. Critical depth
can also be used to refer to the depth from the ground surface for which no support is
required for vertical shafts in clay.
Cyclic Stress Ratio (CSR) – Amplitude of the cyclic shear stress imposed by an
earthquake normalized by the initial effective vertical stress. In a cyclic triaxial test, this
is equal to one-half of the applied cyclic deviator stress divided by the isotropic
consolidation stress.
Deflection ratio – The maximum expected deviation from uniform settlement divided by
the overall length of the structure, which is an approximate measure of the curvature
caused by settlement.
Deltaic – Deposits formed at the mouths of rivers, which result in extension of the
shoreline.
Desert varnish – Also called patina, rock varnish, or rock rust, is thin, dark red to black
mineral coating found on pebbles and rocks surfaces in arid regions.
Desiccation – The process of shrinkage or consolidation of the fine-grained soil

produced by increase of effective stresses in the grain skeleton accompanying the
development of capillary stresses in the pore water. Desiccation is often a result of soil
drying.
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Dewatering – Process where water is pumped from a foundation excavation or pumped
from a pervious soil stratum with the purpose of lowering the water table.
Diatomaceous earth – Soft, siliceous sedimentary rock that usually crumbles into
powder. When crumbled, the particles are silty and contain large amounts of diatoms,
the siliceous skeletons of minute marine or freshwater organisms.
Differential settlement – Difference in vertical displacement between horizontally

spaced points. Often, difference in settlement between structural elements, such as
footings or columns.
Dip – Angle that the surface of the rock forms with a horizontal plane.
Dissipate – Increase or decrease of pore water pressure in order to achieve an

equilibrium condition. Can also refer to the decrease in the magnitude of a value with
depth, such as the dissipation of stress increase with depth.
Dispersive clays – Clays containing a high percentage of dissolved sodium in the pore
water, such that when exposed to water, are very susceptible to erosion.
Distortion – The slope of the expected settlement profile or the ratio of the settlement
between two points to the distance separating the points.
Disturbed specimen – Soil specimen obtained without care taken to preserve the
volume or structure of the soil. Disturbed specimens are used for index tests, and are
not used for strength or compressibility tests.
Double Drainage – Condition when the excess pore water pressure can drain from the
top and bottom boundaries of the laboratory test specimen or from a layer of clay in situ.
Dune sands – Mounds, ridges, and hills of uniform fine sand characteristically
exhibiting rounded grains
Dynamic cone penetration (DCP) – An in situ test performed by driving a standard-

sized cone into the ground using a drop hammer. This test is detailed in ASTM D6951.
Effective diameter – The grain size that has the primary influence on the average pore
size of the soil, which is typically selected as the grain size corresponding to 5 to 20%
passing on the cumulative particle-size distribution curve.
Effective Stress – The net stress across points of contact of soil particles, generally
considered as equivalent to the total stress minus the pore water pressure.
Ejecta – Loose deposits of volcanic ash, lapilli, bombs, etc.
Elevation head – measure of potential energy of water defined by the vertical distance
of the water surface from a datum.
Equalization – Action of letting something reach equilibrium.

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Equipotential line – Lines or curves define points of constant total head.
Equivalent Fluid Pressure – Horizontal pressures of soil, or soil and water in

combination, which increase linearly with depth and are equivalent to those that would
be produced by a heavy fluid of a selected unit weight.
End of Primary Consolidation (EOP) – When all the excess pore pressure in the soil
created by the increase in stress is dissipated and the soil enters into secondary
compression.
Estuarine – Mixed deposits of marine and alluvial origin laid down in widened channels
at mouths of rivers and influenced by tide of body of water into which they are deposited
Excess Pore Pressures – Increment of pore water pressures greater than hydrostatic
values, produced by application of normal stresses or shear stresses.
Exit Gradient – The hydraulic gradient (difference in head at two points divided by the
distance between them) at the point where water exits soil. Exit gradients are often used
as an indicator of erosion at the downstream toe of dams and levees.
Expansion Index – Percent swell multiplied by 10 for the ASTM D4829 test.
Extraction wells – Pumped wells that withdraw groundwater or contaminated

groundwater from an aquifer.
Fibric peat – Peat in which the original plant fibers are slightly decomposed and contain
67% or more fibers.
Field Boring Log – Logged information of a boring prepared during the drilling process.
A typical field log includes all the relevant information for the boring that was completed,
including a unique boring identification number, date of drilling, personnel on-site, boring
advancement method (i.e., auger, rotary wash, direct push, sonic), depths where
samples were obtained, type of samples (i.e., split-barrel and Shelby tube), hammer
type, raw SPT N-values, water level observations, and preliminary estimates of
stratigraphy. If available, the global positioning system (GPS) coordinates should be
included. The field log provides a unique designation of each recovered sample,
whether disturbed or undisturbed, as well as a field visual classification of the sample in
accordance with ASTM D2488.
Fill – Any constructed soil deposit. It can range from soils that are free of organic
matter and that are carefully compacted (controlled fill) to heterogeneous accumulations
of rubbish and debris (uncontrolled fill).
Final boring log – Official engineering record of the drilling and sampling efforts that is
prepared using the information from the field boring log, and lab and field test results.
Fine-grained soils – Soils that contain 50% or more particles passing a No. 200 (75
μm) sieve.
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Finite difference method (FDM) – Numerical method that approximates derivatives by
finite differences to solve differential equations with geotechnical applications, including
consolidation, seepage, and stress-deformation analysis. In many cases, a physical
body (e.g., soil mass, retaining wall, etc.) is discretized by dividing the geometry into
small regions where properties are be assumed to be uniform.
Finite element method (FEM) – Numerical method in which a physical body (e.g., soil
mass, retaining wall, etc.) is discretized by dividing the geometry into small areas, called
elements, where properties are be assumed to be uniform. Adjacent elements in the
body are connected at nodes. Global equations are developed to relate the elements,
the constitutive theory assigned to the elements, and the selected boundary conditions.
FEM is commonly used to solve stress-deformation and seepage problems in
geotechnical engineering.
Fissured – Soils that break along predetermined surfaces with little resistance.
Fissuring in soils may be an indicator of overconsolidation.
Flat plate dilatometer (DMT) or Marchetti Dilatometer – An in situ test that utilizes a
device consisting of a robust steel blade that is pushed into the ground and then
periodically stopped to allow the controlled measured inflation of a flexible steel
membrane. The testing procedures are presented in ASTM D6635.
Floating foundation – see compensated foundation.
Floodplain – Deposits laid down by a stream that within a portion of its valley is subject
to inundation by floodwaters
Flow banding – Layering that is sometimes seen in rocks formed from magma.
Flow line – Paths that water particles take when flowing through a soil. Flow lines are
an element of flow nets.
Flow net – Graphical solution to the La Place equation used to show the spacial
variation of total head. Flow nets are used for seepage calculations in geotechnical
engineering.
Flow slide – Shear failure in which a soil mass moves over a relatively long distance in
a fluid-like manner, occurring rapidly on flat slopes in loose, saturated, uniform sands,
saturated silts, or in highly sensitive clays.
Foliation – Laminated structure of the minerals in a rock created by deformation.
Free swell – Condition in which the soil is allowed to swell with no confining stress
being applied.
Full scale – Loading condition for a sensor where the maximum design load is applied.
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Fuller's earths – Soils having the ability to absorb fats or dyes. These soils have the
capability to decolorize oil or other liquids without chemical treatment. They are usually
high plasticity sedimentary clays.
Fully softened shear strength – The drained shear strength of a clay in its normally
consolidated state.
Glacial soils – Material transported and deposited by glaciers, or by meltwater from

glaciers.
Glacial till – An accumulation of debris, deposited beneath, at the side (lateral

moraines), or at the lower limit of a glacier (terminal moraine). Material lowered to the
ground surface in an irregular sheet by a melting glacier is known as a ground moraine.
See also Boulder Clay.
Glacio-fluvial deposits – Coarse- and fine-grained material deposited by streams of

meltwater from glaciers. Material deposited on the ground surface beyond the terminal
edge of a glacier is known as an outwash plain. Gravel ridges are known as kames and
eskers. Depressions are known as kettles and can be filled with peat.
Glacio-lacustrine deposits – Material deposited within lakes by meltwater from

glaciers, consisting of clay in central portions of lake and alternate layers of silty clay or
silt and clay (varved clay) in peripheral zones.
Glassified sand – Granular deposits at the ground surface occurring after an intense
forest fire.
Goodman jack – See Borehole jack.
Goonies – Cobbles found floating in a soil matrix.
Gravel – Soil particles that pass through a 3-inch square opening sieve but are retained
on a No. 4 (4.75 mm) sieve. Gravels can be divided into: (1) coarse gravels, gravel
particles that are retained on a ¾-inch square opening sieve, and (2) fine gravels, gravel
particles that pass through a ¾-inch square opening sieve.
Grove sand – See Sugar sand
Gumbo – Fine-grained, highly plastic clay of the Mississippi Valley. It has a sticky,
greasy feel and forms large shrinkage cracks on drying.
Gyp or gip soil – Gypsum soil (or soil containing gypsum) or caliche soil.
Hardpan – Soil layers that have become hard as rock due to cementing minerals, and
do not become plastic when mixed with water, and are relatively impervious. It has also
been applied to any hard or overconsolidated layer that is hard to excavate. Because of
this ambiguity, Sower (1979) recommends that engineers should avoid this term
because many lawsuits have centered about definition. The name implies a condition of
a soil rather than a type of soil.
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Hillwash – Fine colluvium consisting of clayey sand, sand silt, or clay.
Hogging – Manifestation of differential settlement in a structure that results in concave

downward shape.
Homogeneous soil – Soils with the same color, appearance, and properties from point
to point. No soil deposit is truly homogeneous.
Humus – Brown or black material formed by the partial decomposition of vegetable or

animal matter. It is the organic portion of soil.
Hydraulic conductivity – Discharge velocity of water through a unit area under a unit
hydraulic gradient. Can also be viewed as a coefficient of proportionality relating
seepage velocity to hydraulic gradient. Often called permeability in geotechnical
Hydraulic gradient – Head loss divided by the length over which the head loss occurs.
Hydraulic head or Total head – Measure of potential energy calculated as the sum of
the elevation head, velocity head, and pressure head.
Hydrodynamic Lag Time – See Lag Time.
Hydrostatic – Condition of equilibrium of fluids for no-flow conditions. Also referred to a

condition where stresses or pore water pressures are equal in all directions.
Hydrostatic pore pressures – Pore water pressures or groundwater pressures exerted

under conditions of no flow where the magnitude of pore pressures increase linearly
with depth below the ground surface.
Igneous rocks – Rocks formed from the cooling and solidification of magma.
Inherent anisotropy – Variation of shear strength as a function of the direction of the

failure plane. It is the result of significant differences in the soil structure which occur
during the formation of the soil.
Intact sample – See Undisturbed sample.
Isotropic soil – A soil mass having essentially the same properties in all directions,
referring primarily to stress-strain or permeability characteristics.
Isotropic – Equal in all directions.
Kaolin – White or pink clay of low plasticity. It is composed largely of minerals of the
kaolinite family.
Karst – Terrain usually formed from the dissolution of rocks such as limestone,
dolomite, and gypsum. It normally contains an underground drainage system composed
of sinkholes and caves.
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Lacustrine – Material deposited within lakes (other than those associated with
glaciation) by waves, currents, and organo-chemical processes; deposits consist of
unstratified organic clay or clay in central portions of the lake and typically grade to
stratified silts and sands in peripheral zones
Lag time – Time required for an instrument to respond to a change in input.
Laminated – Layering consisting of different materials or material colors of less than ¼

inch in thickness.
Lamination – Sequence of fine layers in a small scale (usually less than one centimeter
in thickness) normally observed in sedimentary rocks.
Landslide deposits – Considerable masses of soil or rock that have slipped down,
more or less as units from their former position on steep slopes.
Laterites – Residual soils rich in iron formed in hot and humid climates (tropical
regions). The cementing action of iron oxides and hydrated aluminum oxides makes
dry laterites extremely hard. The high content of iron oxide makes many laterites to be
rusty-red. Laterites are usually developed after significant weathering of the parent
rock.
Ledge – Colloquial name for bedrock in Vermont and New Hampshire.
Lens or Lensed – Small pockets of dissimilar soil scattered throughout the mass of a
clay.
Load increment ratio (LIR) – Variable used to quantify the change in load to a test
specimen. Defined as the ratio of the change in stress to the current stress. A load
increment ratio of unity indicates that the load was doubled.
Loam – Low plasticity sandy silt or silty sand mixed with organic matter that is well
suited to tilling. Mainly applies to the uppermost soil layer and should not be used to
describe deep deposits of parent materials. Major soil type in the USDA system. Not
considered a USCS soil type in conventional geotechnical engineering (ASTM D2487
and D2488).
Loess – A wind deposited, calcareous, unstratified deposit of silts or sandy or clayey silt
traversed by a network of vertical tubes formed by the decay of root fibers. Loess slopes
have the ability to withstand vertical cuts.
Lugeon – Flow of one liter of water per meter per minute under a pressure of 10 bars
(145 psi) in a constant head double packer test.
Marine soils – Material transported and deposited by ocean waves and currents in
shore, near shore, and offshore areas.
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Marl – Calcium carbonate or lime-rich sedimentary rock. It is mainly composed of a
mixture of sand, silt, and/or clay. Marls are often light to dark gray or greenish in color
and sometimes contain colloidal organic matter.
Matric suction – Difference between pore air pressure minus pore water pressure.
Often used in the characterization of partially saturated soils.
Maximum past pressure – See Preconsolidation pressure.
Metamorphic rocks – Rocks transformation by heat, pressure, or both. This

transformation can alter the physical and chemical properties of the rock.
Minimally disturbed sample – See undisturbed sample.
Modified compression index – Parameter which quantifies the compressibility of

normally consolidated soil in one-dimensional compression. Normally, it is the log-linear
slope of the compression curve defined by axial (y-axis) and the logarithm of vertical
effective stress (x-axis).
Modified recompression index – Parameter which quantifies the compressibility of

overconsolidated soil in one-dimensional compression. Normally, it is the log-linear
slope of the compression curve defined by axial (y-axis) and the logarithm of vertical
effective stress (x-axis). Normally obtained by a rebound-reload loop in a consolidation
test. Also called the modified swelling index.
Montmorillonite – A group of very small clay minerals with extreme swelling and
shrinking properties. Normally results from volcanic or hydrothermal activities.
Mucks – Peat deposits which have advanced in decomposition to such extent that the
botanical character is no longer evident.
Muskeg – North American term for peat. According to Sowers (1979), the bogs in
which the peat forms are often called muskegs.
Nominally disturbed sample – See Undisturbed sample.
Normal consolidation or normally consolidated – Condition of a soil where the

current effective stress is the maximum effective stress ever realized, and all excess
pore pressures have been dissipated.
Nested piezometers – Multiple standpipe piezometers that are installed in a single

boring with an impervious seal separating the different measurement zones.
Neutral stress – Synonym for pore water pressure. This is a term that is normally used
in older geotechnical literature.
One-dimensional compression test – A compression test, normally a consolidation

test, in which the soil specimen is confined laterally and deformation occurs in the same
direction as the vertically applied load.
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Open standpipe piezometer – Type of standpipe piezometer, similar to an open well,
except that the screen extends only across a specific stratum of interest. Seals are
installed above and below this zone to only allow water to enter from the stratum of
interest.
Open well piezometer – Type of standpipe piezometer with a full-length screen and a
surficial seal that is best suited to relatively homogeneous soil profiles. In layered soils,
the measured groundwater level will correspond to the layer with the highest total head.
Organic soils – Soil material containing enough organic or vegetable matter as to

influence the engineering properties. .
Osmotic pressure – Pressure in a solution that is the product of the molar

concentration of the solute solution, the universal gas constant, and the temperature, in
degrees Kelvin.
Overconsolidation – The condition that exists if a soil deposit has been fully
consolidated under an effective stress greater than the existing effective stress.
Peat – Organic soil derived from decomposing plant material, normally sedimented in
an anaerobic environment. Peats are considered to have less than 25% ash (mineral
components) per dry weight.
Perched water table – Spatially limited unconfined water table, separated from the
main groundwater regime, caused by the presence of a low permeability layer.
Piedmont soils – Alluvial deposits at the foot of hills or mountains; extensive plains or
alluvial fans
Piezocone test (CPTu) – Cone penetration test where the pore pressures behind the
tip of the cone are measured during penetration.
Piezometer – A device installed for measuring the pressure head of pore water at a
specific point within the soil mass.
Pinnacle – Is an individual and isolated column of rock, often associated with karst
terrain.
Piping – The movement of soil particles as the result of unbalanced seepage forces
produced by percolating water, leading to the development of boils or erosion channels.
Pit run sand and gravel – See bank run.
Plane strain – A strain boundary condition where strains are only allowed in two
directions. Plane strain boundary conditions often result in a three-dimensional stress
state. Many geotechnical engineering analyses that are performed in two-dimensions
assume that plane strain boundary conditions exist in the third dimension.
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Plastic equilibrium – The state of stress of a soil mass that has been loaded and
deformed to such an extent that its ultimate shearing resistance is mobilized at one or
more points. Solutions employing plastic equilibrium assume full mobilization of the
soil’s shear strength within a soil mass or along a specified failure surface.
Pluff Mud – Colloquial term for a very soft, odorous mud encountered in South
Carolina.
Point bar – Alternating deposits of arcuate ridges and swales (lows) formed on the
inside or convex bank of river bends. The ridge deposits consist primarily of silt and
sand, while swales are often clay filled
Positive cutoff – The provision of a line of tight sheeting or a barrier of impervious

material extending downward to an essentially impervious lower boundary to intercept
completely the path of subsurface seepage.
Preconsolidation pressure – Maximum effective stress, under conditions of full pore

pressure dissipation, that has been applied to a soil in the past. Synonym for maximum
past pressure.
Prefabricated vertical drain (PVD) – Plastic strip, normally encased in a filter fabric,
that can be inserted into the soil by a mandrill to facilitate the dissipation of excess pore
water pressures.
Pressure head – Synonym for piezometric head. Component of total head that is equal
to the water pressure divided by the unit weight of water.
Primary consolidation – The time-dependent compression of a soil under the

application of a stress that occurs while excess pore pressures dissipate with time.
Pumice – Porous rock associated with lava flows. May be mixed with nonvolcanic
sediments.
Pyroclastic soils – Soil-like material ejected from volcanoes and transported by

gravity, wind and air.
Radial consolidation – Consolidation of a soil mass where pore water pressures are
dissipated laterally or radially. Radial consolidation occurs when the drainage boundary
is cylindrical (stone column) or a strip or line (PVD).
Recompression index – Parameter which quantifies the compressibility of

overconsolidated soil in one-dimensional compression. Normally, it is the log-linear
slope of the compression curve defined by void ratio (y-axis) and the logarithm of
vertical effective stress (x-axis). Normally obtained by a rebound-reload loop in a
consolidation test. Also called the swelling index.
Reconstituted – Soil sample formed for laboratory testing at a given density and water
content. This term is mainly used for coarse-grained test specimens.
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Recycled concrete aggregate (RCA) – Recycled road or structural concrete. The
concrete is usually processed and screened. The processing consists of crushing the
concrete into smaller pieces. Any leftover steel is removed using a magnet. This type
of material can serve as a replacement for natural stone aggregates.
Recycled or reclaimed asphalt pavement (RAP) – Excavated and processed asphalt

concrete from road wearing surface. When properly processed, it consists of high-
quality and well-graded aggregates coated by asphalt cement.
Recycled or reclaimed asphalt shingles (RAS) – Recycled shingles that are used as
aggregate for hot mix asphalt. Depending on the quality, this can reduce the cost of the
new asphalt mix and the amount of fine aggregate used in the mix.
Recycled pavement material (RPM) – Pulverized mixture of asphalt and base course
material usually forming a broadly graded material.
Relative density – Parameter used to quantify the density of a soil relative to the
loosest and densest states. It is calculated as the ratio of the difference between the
maximum void ratio and current void ratio to the difference between the maximum and
minimum void ratios.
Remolded – Soil sample mixed to a given water content to achieve a desired

consistency. This term is mainly used for fine-grained soils.
Residual shear strength – The lowest drained shear strength of a soil that is achieved
by shear displacement along a failure plane until particle alignment is achieved. This
term is normally reserved for fine-grained soils. Residual conditions are often
associated with slickensides forming on the failure plane.
Residual soil – Material formed by disintegration of underlying parent rock or partially

indurated material.
Response to wetting tests – Tests in which the volume change of the soil is measured
as the soil is given access to water or if the water content is reduced by drying.
Riprap – Boulder-size material normally place to strengthen structures against scour,

wave action, and ice erosion.
Rippability – The characteristic of dense and/or rocky soils that can be excavated by
ripping with a rock rake or ripper.
Riverjack – Alluvial cobbles and boulders.
Rock – Natural solid mineral or aggregate of minerals which is normally classified by

the way it was formed.
Rock borehole shear test – In situ method to measure the strength of relatively weak
rock or rock that is easily disturbed upon drilling and coring (e.g., weathered rock,
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fractured rock, shale, etc.). This test is a modification of the Iowa borehole shear test
originally developed for soil.
Rock dirt combination (RDC) – Local term used in the Harrisonburg, VA, area to
describe material from a quarry consisting of a mixture of overburden soil and rock.
Rock flour – Fine-grained soil, normally with silt-sized particles, formed by the grinding
of bedrock by glaciers or by drilling. Rock flour normally classifies as a nonplastic silt.
Rock mass – A large body containing rock in intact and weathered conditions
accompanied by structural discontinuities like fault, joints, etc., which can be
interbedded with soil material.
Rock Quality Designation (RQD) – Calculated parameter used to quantify the quality
of a rock core. It is equal to the total length of recovered core pieces greater than 4
inches in length divided by the recorded core run.
Rock Mass Rating (RMR) – Rock classification system based on uniaxial compressive
strength, RQD, spacing and properties of the joints, and groundwater conditions.
Sagging – Manifestation of differential settlement in a structure that results in concave

upward shape.
Sand – Soil particles that pass through a No. 4 (4.75 mm) sieve and are retained on a
No. 200 (75 μm) sieve. Sands can be divided into: (1) coarse sands, sand particles that
are retained on a No. 10 (2.00 mm) sieve, (2) medium sands, sand particles that pass
through a No. 10 (2.00 mm) sieve and are retained on a No. 40 (425 μm) sieve, and (3)
fine sands, sand particles that pass through a No. 40 (425 μm) sieve.
Secondary compression – Time dependent settlement of soil at constant effective

stress. Normally considered to be a result of particle rearrangement.
Shale – Fine-grained sedimentary rock made of silt and clay particles. Shale usually
breaks along planes of weakness and can slake when subjected to wet-dry cycles.
Shape factor – Ratio of the number of flow channels in a flow net to the number of
equipotential drops.
Shore deposits – Deposits of sands and/or gravels formed by the transporting, erosion,
and sorting action of waves on the shoreline.
Shot rock – Material from a rock quarry that has not been sorted or screened. It
includes everything (from fine sand to small boulders) that can be loaded after a quarry
blast. It is also a name given to riprap, although riprap is typically sorted and graded.
Sedimentary rocks – Rocks formed by the accumulation and cementation of smaller

particles.
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Seismic CPT (SCPT) – Cone penetration test where the cone contains a geophone or
accelerometer in order to measure the shear wave velocity. A seismic source is applied
at the ground surface in the vicinity of the cone hole.
Settlement – Vertical deformation of a foundation element (footing, mat, or pile). Can

also be used to describe the compression of a soil layer under an applied change in
stress.
Silt – Nonplastic or slightly plastic soil particles passing a No. 200 (75-μm) sieve that
exhibit little or no strength when air dried. Silt-sized soils are normally considered to be
larger than 0.002 mm. For classification of silty soils, refer to ASTM D2487.
Size-corrected Point Load Strength Index – Strength obtained from a point load test
where the data have been corrected for the size of the test specimen.
Smear – Alignment of clay particles along a shear surface that creates a thin layer of
low hydraulic conductivity.
Specific surface area – Surface area of soil particles, usually expressed as the area
(units of L2) per gram.
Slickenside – Condition of a shear surface where considerable displacement has taken

place. The clay particles align in the direct of shear, and the shear surface is usually
polished, glossy, or sometimes striated.
Slickensided clay – Clay that has experienced repeated or accumulated displacement

along a fissure or a failure plane causing the surface to be smooth and shiny.
Split Cylinder Test – A test used to determine the tensile strength of rock cores in
which the test specimen is loaded diametrically via hardened steel end platens.
Splitting tensile strength – Tensile strength obtained from a split cylinder test.
Staged – Condition where loading or shearing is performed in incremental stages.
Standard Penetration Resistance – The number of blows of a 140-pound hammer,

falling 30 inches, required to advance a 2-inch O.D., split barrel sampler 12 inches in a
soil mass.
Standard Penetration Test – An in situ test that measures resistance to the

penetration of a standard, thick-walled drive sampler in an open borehole using a drop
hammer. This test proceeds by driving a thick-walled, split-barrel (a.k.a., split spoon)
sampler into the ground using incremental blows from a drop hammer. The sampler is
driven a total of 18 inches into the ground. The procedure is presented in ASTM D1586.
Standpipe piezometer – Watertight pipe with a screened section installed in a borehole

with one or more seal to allow long-term measurement of groundwater levels. The term
is used to refer to both open wells and open standpipe piezometers.
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Stone – Gravel-size particles manufactured by crushing rock.
Stratified – Earth materials with layers of different material or color of at least ¼ inch in
thickness.
Stress path triaxial test – Triaxial test in which both the vertical and horizontal
stresses, and possibly the pore water pressure, are varied systematically to follow
prescribed loading paths.
Strike – Line representing the linear feature of the intersection of a rock surface with the
horizontal plane.
Sugar sand – Local name used for specific types of sands in various places. It is a fine
sandy soil in New Jersey. In Kansas, it refers to a type of granular calcite found in Ness
and Hodgeman counties. In Florida, it refers to a fine sand that does not hold water or
nutrients very well.
Surcharge – Fill or other material used to apply a temporary stress to a compressible

soil layer. The fill is removed after a predetermined amount of compression or
consolidation has occurred.
Swelling index – See recompression index.
Talus - Deposits created by gradual accumulation of unsorted rock fragments and

debris at the base of cliffs or slopes.
Terrace – Relatively narrow, flat-surfaced, river-flanking remnants of floodplain deposits

formed by entrenchment of rivers.
Till – See Boulder Clay.
Tilt – Outward rotational displacement of a retaining wall or other structure.
Time curve – See Time-deformation curve.
Time-deformation curve – A plot relating the deformation of a foundation element or

laboratory test specimen as a function of time after being subjected to a change in
stress.
Tire derived aggregate (TDA) – Lightweight construction material obtained by

shredding or chipping scrap tires. The particle size usually ranges from 0.5 inches to 12
inches. TDA has been used in a wide range of projects, including lightweight
embankment fill, landslide repair or stabilization, retaining wall backfill, roads, vibration
mitigation, among others.
Topsoil – Upper and outermost layer of soil that supports plant life. Usually contains
considerable organic matter.
545
UFC 3-220-10
1 February 2022
Total stress – At a point in a soil mass, the sum of the net stress across contact points
of soil particles (effective stress) plus the pore water pressure at the point.
Trap – Dark-colored, fine-grained, non-granitic intrusive rock. The most common trap
rock is basalt, but also includes peridotite, diabase, and gabbro.
Triaxial permeameter – Pressure chamber holding a cylindrical soil test specimen in a

flexible membrane used to perform hydraulic conductivity tests. Also called a flexible
wall permeameter.
Tuff – Soft porous rock composed of consolidated volcanic ash.
Uncorrected Point Load Strength Index – Strength obtained from a point load test
that has not been corrected for the size of the test specimen.
Underconsolidation – Condition that exists if a soil deposit is not fully consolidated

under the existing overburden pressure and excess hydrostatic pore pressures exist
within the material.
Velocity head – One of the three components of total head, equal to the flow velocity
squared divided by twice the acceleration of gravity. Flow velocity in earth materials is
often slow enough (laminar flow) that the influence of the velocity head can be ignored.
Vertical drains – Drainage conduits, such as stone columns or prefabricated vertical

drains, that are used to allow radial dissipation of pore water pressures and accelerate
consolidation.
Virgin compression – Compression of a soil at stresses in excess of the

preconsolidation pressure or maximum past pressure.
Water Level Indicator - An electrical device used to measure the distance from the
ground surface or top of the casing to the top of the water surface in open pipe
piezometers. Graduations on the electrical cable are used to measure the depth to the
water surface.
Well resistance – Resistance to flow in a prefabricated vertical drain.
Undisturbed Specimen – Soil sample taken with a thin-walled sampler or block sample
with special attention given to maintaining the volume, density, soil structure, and water
content. Undisturbed is often written in quotes since no soil sample can be truly
“undisturbed.” Recently, this term has been replaced with the term intact specimen.
Varved Silt or Clay – A fine-grained glacial lake deposit with alternating thin layers of
silt or fine sand and clay, formed by variations in sedimentation from winter to summer
during the year.
546

DM 7.1 - Navfac - Soil Mechanics - 2022

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UFC 3-220-10

UNIFIED FACILITIES CRITERIA (UFC)

SOIL MECHANICS (DM 7.1)

U.S. ARMY CORPS OF ENGINEERS

Record of Changes (changes are indicated by \1\ ... /1/)

This UFC supersedes UFC 3-220-10N, dated 8 JUNE 2005.

• Whole Building Design Guide web site http://www.wbdg.org/ffc/dod.

CHRISTINE T. ALTENDORF, PhD, R. DAVID CURFMAN, P.E., SES

NANCY J. BALKUS, P.E., SES MICHAEL McANDREW

Document: UFC 3-220-10, Soil Mechanics

Superseding: UFC 3-220-10N, Soil Mechanics

The lasting value of DM 7.1 is attributed to its success in distilling geotechnical

This Page Intentionally Left Blank

Figure 1-1 Typical Angularity of Bulky Grains (after Sowers 1979)........................ 8

Table 1-1 Principal Soil Deposits in Terms of Origin ............................................ 2

1-2 SOIL DEPOSITS.

1-2.1 Geologic Origin and Mode of Occurrence.

Major Division Principal Soil Deposits Pertinent Engineering Characteristics

Peats: Somewhat fibrous aggregate of decayed and decaying

Ejecta: Loose deposits of volcanic ash, lapilli, bombs, etc.

Residual sands and fragments of gravel-sized material formed by

Transported soils: See Table 1-2

Major Division Principal Soil Deposits Pertinent Engineering Characteristics

Major Division Principal Soil Deposits Pertinent Engineering Characteristics

1-3 SOIL VISUAL DESCRIPTION, IDENTIFICATION, AND

Standardized procedures for visual description, identification, and formal classification

Boulders: Rock particles will not pass a 12-inch square opening.

1-3.2 Visual Description and Identification (ASTM D2488).

1-3.2.1 Visual Description.

1-3.2.1.1 Descriptors for All Soils.

1-3.2.1.2 Descriptors for Fine-Grained Soils.

1-3.2.1.3 Descriptors for Coarse-Grained Soils.

Figure 1-1 Typical Angularity of Bulky Grains (after Sowers 1979)

1-3.2.2.1 Identification of Fine-Grained Soils.

1) None, if no visible change in the specimen was observed,

1) Nonplastic, if a 1/8-in-diameter thread cannot be rolled at any water content,

Table 1-3 Classification of Fine-grained Soils

Soil Symbol Dry Strength Dilatancy Toughness and Plasticity

ML None to low Slow to rapid Low or thread cannot be formed

CL Medium to high None to slow Medium

MH Low to medium None to slow Low to medium

CH High to very high None High

1-3.2.2.2 Identification of Coarse-Grained Soils.

For coarse-grained soils, the identification is only based on visual observations.

1-3.3 Unified Soil Classification System (ASTM D2487).

1-3.3.1 Classification of Fine-Grained Soils.

Figure 1-2 Plasticity Chart

1-3.4 Soil Classification for Highways (AASHTO).

The American Association of State Highway and Transportation Officials (AASHTO)

GI =( F − 35)[0.2 + 0.005( LL − 40)] + [0.01( F − 15)( PI − 10)] (1-3)

1-3.5 Other Classification Systems.

Table 1-4 Other Soil Classification Systems

1-3.6 Common Soil and Rock Names.

Beachrock: See reefrock.

Plasticity Index ≤6 ≤6 NPa ≤ 10 ≤ 10 ≥ 11 ≥ 11 ≤ 10 ≤ 10 ≥ 11 ≥ 11 ≥ 11

Group Index, GI GI =( F −35) 0.2+0.005( LL −40 )  + 0.01 ( F −15) ( PI −10 )

Calculate the group index as:

Blow sand: Term normally used for wind-driven or drifted sands.

Colluvium: Loose soil deposited at the bottom of a slope.