2258-6003 210514 192200
2258-6003 210514 192200
2258-6003 210514 192200
II
22 r; b 00 1
· Prefabrication
of Concrete
Structures
Bibliotheek TU Delft
IIIIIIIIHI
C 1849689
•
Organizing Committee:
Edited by:
AJ. Hogeslag
J.N.J.A Vambersky
J.C. Wa/raven
by order of:
Cover:
Bloemenveiling Westland
Waco-Liesbosch Beton BV, Utrecht, The Netherlands
Prefabrication
Foreword 5
General Introduction 7
Nowadays, high requirements are imposed on the quality of structures and building
processes. On the other hand the last decade has shown a drama tic increase of man-
hour costs especially in the building industry. Owners, developers and construction
companies are demanding a more man-hour efficient way of building.
Designers are being required to meet ever more stringent requirements concerning the
efficiency of construction to cut down its own spiraling costs.
The heavy labour on the construction site, which very of ten takes place under wind,
rain and frost conditions is criticized by the unions and the health organizations.
In this respect prefabrication of concrete structures offers a number of advantages,
which deserve a due consideration.
Prefabrication of structural members stands for high quality with regard to strength,
stiffness and durability. It offers a wide variation in spans, shapes and colors.
Manufacture and erection of prefabricated concrete structures can occur in short
periods, with a small crew of qualified people, who can work in more favourable
working circumstances.
Designing and building in prefabricated concrete however requires other concepts and
strategies than applied for in-situ concrete, not only with regard to strength and
stability but also with regard to organizational aspects. Only when this is realized, the
way is open to successful applications.
The papers presented at this seminar are dealing as weil with the fundamental
principles of prefabrication as with the recent developments in structural design, code
provisions production techniques and products. As the field of prefabrication is very
large, actually too large to be sufficiently covered by a seminar with a time span of
only one and a half day, the subjects dealt with were this time limited to only those
related to buildings.
The seminar is intended to give a survey of concepts which are experienced and known
to be appropriate for actual building practice, to show new trends and developments
and to open the way to face new challenges in structural engineering.
GENERAL INTRODUCTION
m2 pro capita
0 . 70
Finland (O.65)
0.60
0 . 50
0.40
Holland (0.36)
Norway (O.l2)
0.20
Oenmark (0.1 5)
0.10
BelqiulII (0.08)
Italy (0.06) UK (0.05)
Franc e (0.03)
0.00 L -_ __ __ _- " ' =n:L.JlJL..IllJllll--.5.ll.il.lJlcJ..Q.
O,..g
. O.Q!0lL
SESSION 1
DESIGN ASPECTS
-15-
1. INTRODUCTION
The dramatic increase of manhour costs during the past several years has impacted
heavily on the building industry . Owners, developers and construction companies are
demanding more manhour efficient way of building.
Designers are being required to meet ever more stringent requirements concerning the
efficiency of construction to cut down its ever spiraling costs.
The heavy labour on the construction site, which very of ten takes place under wind,
rain and frost conditions is criticised by the uni ons and health organisations.
These are some of the reasons why today more and more owners, developers and
construction companies are choosing precast concrete for their building projects.
2. DESIGN
Every construction system has its own specifics, which more or Ie ss influence the lay-
out, storey height, stability, statical system etc.
To achieve an optimum design and an optimum structure a design should, from the
very outset, respect these specifics and particular demands of the structure and
system aimed at.
Hence, it is very important when designing a precast concrete structure, to realize
that the best result is arrived at if the structure is designed as a precast concrete
structure from the very outset and is not merely adapted from the traditional cast in
situ or masonry design to a pre cast one.
Not respecting this may cause unnecessary faults and problems during the fabrication
and construction of the elements and during the service life of the whole building.
This is a very important point which, strangely enough, is still very of ten neglected.
It should be realized, that it is a good initial pre cast concrete design which to a great
extent determines whether the advantages of precast concrete shall fully develop and
so contribute to cutting down the construction time and costs and ensuring better
quality, as weIl as cutting down life cycle costs by assuring adaptability and openess to
current and future demands or not.
Designers should, therefore, bear in mind the possibilities, restrictions, advantages and
disadvantages of pre cast concrete, including its detailing, manufacture, transport,
erection and serviceability stages before coming to a final design in precst concrete.
As a rule modular co-ordination should be used throughout the building in every design,
and also necessary tolerances should be carefully considered.
~ Detailing
All successful ideas, formulas and inventions are mostly based on strikingly simple
solutions. This is generally valid, but even more for design and detailing of pre cast
concrete structures.
As time saving is one of the most important advantages of precast concrete, care
should be taken that the details used correspond with the principles of precast
concrete design and fit into the short construction time.
Too elaborate or vulnerable details should be as much as possible avoided.
A good design in precast concrete should use as simple details as possible, the
simplicity of the details determining all advantages of precast concrete.
As examples of simple details can be se en the figures below:
-17-
-~l If-- I]
P'eca" colum,
I,
11 ' I1 Erection wedge
I1
111 I~
11
I
11
11' 11
1
Non-shrinking
grout I',
[
I Fully grouted
~ (
-f/~
:>:-
" .~ .= .-.:""'
.: ..-;:j -----j--- Box-shaped
foo tin..g
[ [ [ [ Levelling pad
(a) (b)
1 I
FCJ5
Concrete poured on site
Where possible standardized units, elements and systems should be used, and non-
standard solutions and details should be avoided wherever possible.
This also applies to the grid line distances, storey heights, stair step dimenstions, etc.
Precast concrete elements should be as large as possible, bearing in mind limitations
of manufacture, transport and erection and crane-lifting capacity, to re duce erection
handlings.
Precast concrete columns should preferably be made as one single unit over two or
more storeys to reduce the time taken in manufacture and erection, as weU as the
number of connections. To keep the detailing of the column/beam connections as
simple as possible and to permit the use of columns in one piece over two storeys or
more, the column/beam connection should preferably be realized by simply supporting
the beams by corbels (corbels forming part of the columns).
Cast in situ structural concrete topping should be used as little as possible, since it is
a disturbing element in the erection process consuming time (and money).
The connections between elements should be as simple as possible and in no way
attempts should be undertaken to make the connections similar to the cast in situ
ones.
b. Strength
A connection must have the strength to re sist the forces to which it will be
subjected during its lifetime. Some of these forces are apparent, caused by dead
and live gravity loads, wind, earthquake, and soB or water pressure. Others are not
so obvious and are frequently overlooked. These are the forces caused by restraint
of volume changes in the members and those required to maintain stability.
Volume changes are caused by temperature change, creep and shrinkage of the
-19-
d. Ductility
Ductility is usually defined as the ability to accommodate relatively large
deformations without failure. In structural materials, ductility is usually measured
by the amount of deformation that occurs between first yield and ultimate failure.
Ductility in building frames is usually associated with moment resistance. This is
particularly true in designing for earthquake forces, which is where concerns over
-20-
e. Durability
Evidence of poor durability is usually exhibited by corrosion of exposed steel
elements, or by cracking and spalling of concrete. Connections which will be
exposed to weather should have steel elements adequately covered with concrete,
or should be painted or galvanized. If not, sufficient non corrosive connection
materials should be used.
All to weather or other aggressive environment exposed connections should be
periodically inspected and maintained. Most precast concrete elements are of high
quality, and flexural cracking is seldom a serious problem, provided tensile stresses
are kept within code limits. However, local cracking or spalling can occur wh en
improper details result in restraint of move ment or stress concentrations.
:o!:olo14
_:. . '
d ii
". .
Q k 0'
. . .0
Q~
r----
" (a) Headed studs with wire tie
• • • • • •0 , • 0 °
f. Fire resistance
Many precast concrete connections are not vulnerable to the effect of fire and
require no special treatment. For example, the bearings between slabs and beams
do not generally require special fire protection. If the slabs or tees rest on
elastomeric pads or ot her combustibie materiais, protection of the pads is not
generally needed because deterioration of the pads will not cause collaps. Af ter a
fire, the pads could be replaced. Connections in which weakening by fire would
jeopardize the structures stability should be protected to the same degree as that
required for the structural frame. For example, an exposed steel bracket
supporting a beam may be weakened enough by a fire to cause the beam to
collapse. Such a bracket should be protected. The amount of protection depends on
(a) the stress-strenght ratio in the steel at the time of the fire and (b) the intensity
and duration of the fire. Connections which require a fire resistance rating will
usually have steel elements encased in concrete. Other methods of fire protection
include enclosing with gypsum wallboard, coating with fire protecting coating, or
spraying with fire protection material. There is evidence that exposed steell
-22-
g. Fabrication simplicity
Maximum economy of precast concrete construction is achieved when connection
details are kept as simple as possible, consistent with adequate performance and
ease of erection. Furthermore, complex connections are more difficult to design,
to make and to con trol and will of ten result in poor fitting in the field. This can
contribute to slow erection and less satisfactory performance. Underneath follows
a list of items to consider during the design in order to improve fabrication
simplicity. In many cases, some of these items must be compromised in order for
the connection to serve its intended function, or other functional reasons.
g.9 Allowalternates
Very of ten, a precast concrete manufacturer will prefer certain details over
others. The producer should be allowed to use alternative methods or
materiais, provided the design requirements are met. Allowing alternate
solutions will of ten result in the most economical and best performing
connection.
i. Erection simplicity
Much of the advantage of precast concrete construction is due to the possibility of
fast erection of the structure. To fully realize this benefit, and to keep the costs
within reasonable limits, field connections should be kept simpie. In order to fulfill
the design requirements, it is sometimes necessary to compromise fabrication and
erection simplicity. Underneath follows a list of items that should be considered
during the design and detailing of the connections in this respect.
i.B Allowalternates
As with fabrication, the precast concrete manufacturer or erector may
prefer certain details or procedures not anticipated by the designer. Allowing
alternate solutions will of ten resuIt in more economical and better
performing connections.
k. Economy
The costs of the connection itself depend on the magnitude of the forces to be
carried over and the repetition (number of the same connections) involved. For the
economicaly justified choice it is however important to consider also the influence
which the connection has on the total cost of the prefabricated structure as a
whoie. The direct costs of the connection should be weighed against the costs of
the element manufacturing, storage, transport, erection and finishing.
I. Appearance
When the precast members are exposed, the appearance of the connection is of ten
an important consideration. It is sometimes necessary to compromise fabrication
and erection simplicity, and hence, increase the cost, to provide a satisfactory
appearance.
2.3. Standardization
Standardization has been always a hot item in the precast concrete industry. There are
enough examples known from differènt countries, where standardization has been
carried out to such an extent that stereotype buildings with the same appearance and
character over the whole country were the result. This so far carried standardization
has very of ten and understandably, lead to a certain aversion against prefabrication.
This is ofcourse a wrong approach.
In general, when architecture and the building structure are optimized for each
building - such as always should be the case - the component "standard" should be
limited as to allow wide applicability.
This means that precast concrete production plants should be as versatile as possible
in order to guarantee continuity of production and the standardization should be
pursued for details, cross-sections connections, base-type products and systems, taking
modular coordination into account, rather than 100% standardization of components
let stay the 100% standardization of the whole building.
3. SAFETY
Stability and safety are necessary structural considerations in prefabrication. In cast
in situ traditional buildings, up to a certain height the stability is mostly assured
without applying special provisions, but in precast concrete the stability and safety of
the structure should alsways be considered, even if the height is very smal!.
-28-
Also the possibility of progressive collapse should be carefully studied and, where
necessary, provision should be made to prevent this or at least to reduce the risk to an
acceptable minimum.
Generally spoken properly designed and detailed structures, with proper attention paid
to continuity, structural integrity and anchorages and, most important, ductile
performance of reinforced concrete members and connections, re sist in general
accidental loading without turning into a mechanism af ter the failure of the first link.
BIBLIOGRAPHY
1. STUPRé (uitgave Betonvereniging), "Constructieve verbindingen van
geprefabriceerde betonelementen", Zoetermeer, The Netherlands 1976
2. L.O. MARTIN and W.J. KORKOSH, "Connections for Precast Prestressed Concrete
Buildings including earthquake resistance", P.C .I. March 1982
3. FIP Recommendations "Design of Multi-Storey Precast Concrete Structures"
Thomas Telford London 1986
-29-
I.INTRODUCTION
Short and to the point: To satisfy the requirements as to stability and rigidity of
prefabricated bearing structures the partly restrained column seems to be the most
suitable solution.
By "restrained column" is meant the whole sequence of elements from the real
column, via the shearwall to the core (a system of composed walis), which all can be
seen statically as a partly restrained column.
This type of construction can easily be analysed by the approximate method, shown in
the figure.
FVd
12
~"-I/,,-~
c c=oo
T~~
c
Fig. 1
111 C
-=--+-- FE,2 =i (column)
FE FE , 1 FE , 2
2e
FE
, 2 = g;- (core)
E
o
EI + I = moment of inertia of uncracked cross section E" "3
n ~ Ó
n-l' 0
n ~ 10 o.k.
10 > n > 5 acceptable
n < 5 improve design
n <1 incorrect
2. FRAMES
The monolithic frame does not fit weil in our design-philosophy of prefabricated
structures. To -enable a quick erection, the connections between the elements must
be simple. Moment resistant frames for lateral load resistance need moment-
connections and unfortunately the most effective and best performing moment-
connections usuaily employ some cast-in-place concrete, which is laborious and thus
not in line with the speed of erection.
111 I 1
Fig. 2
floors
e lement
?
brace
~
I
I
/
1lieverarm
0= hinge
Fig. 3
1T f rame
Fig. 4
But still by removing the points of zero bending moment from the middle of the
column to the boHom the rigidity of the frame decreases. On the other hand a possible
moment-connection by means of threaded bars is succeptible to dimensional variation.
Altogether the moment resistant frame is to our opinion a less attractive and out of
date structure.
An exception is the external bearing framework, obtained by joining efficiently rigid
units together, to which I will return later.
-,-
. mI
•-
~;:L-----...J~~
rnl
I
r-tr--tr
,.,
~.I
.. ,
!. ..
I
I
-
Fig. 5
Two principal methods for restraining precast columns in the foundation are:
connection in a foundation-socketj
connection by protruding bars.
Fig. 6
The connection with steel base-plate and extended anchors (a rea I steel-connections)
is attractive from the point of view of erection, but quite costly.
The connection in a foundation-socket is mostly applied if no piles are used. In
Germany and Belgium these sockets are prefabricated and as a half standard product
put on the market.
-33-
If a pile foundation is used the socket will be too laborious, employs much material
and therefore is Ie ss attractive because of the costs.
In The Netherlands the application of foundation on piles dominates and is the
foundation socket Ie ss popular. MosUy the connection by extended bars is used, in
spite of the disadvantages of the wet connection, the need of bracing of the columns
and a large dimensional accuracy. Also the cross-sectional area of the columns is
larger in connection with the required spa ce for the recesses.
The aim is to transfer the lateral loads as much as possible to all columns. Then there
is hardly any difference between the loads on the columns so that Ie ss column-types
will do, or even one type will do without a waste of material.
r~
n colums
Fig. 7
To transmit horizontal forces the floor- and/or roof slabs must possess a certain
amount of rigidity within their own plain. Next speaker will deal with that topic. It
implies also contradictionary demands on the beam-column connection, because on the
one side the build-up forces caused by the restraint of shrinkage, creep and
temperature-change must be prevented and on the other hand the lateral wind forces
must be transferred.
The solution can be found on the ground of the characteristic differences between
these two loads: the slow proces of shrinkage etc. opposite to the fast developing
lateral forces. The recess around the threaded bars in the connection can therefore be
filled up with bitumen: plastic for slowly build-up forces, but rigid for fast build-up
forces. Of course each column must be checked with regard to its own inclination and
geometrical imperfections.
The loads are usually distributed to the columns in proportion to the flexural column-
stiffness (depending on cross-sectional area, available normal force and rotational
stiffness of the foundation). "Exact analysis" has become increasingly complex and
adds considerably to the time and cost of engineering. Uniformity in the design and a
regular basic grid of the columns can simplify the method of analysis to a large
extent.
-34-
11 column
threaded II
bar
onch~~.~__~-'.I
fostener
Fig. 8
This type of structure is very ductile because each column provides his own stability.
Besides, by the simple connections, the structure can be regarded as suitable for
demountable design. The provision of the stability, as defined before, is not limited to
single-storey (industrial) buildings, as it will appear from the following examples:
Extension of the flower-auction-buildings at Aalsmeer and Honselersdijk.
Some data:
Column grid: about 16 x 20 m
Cross-sectional area columns:
1,2 x 1,2 m 2
Loads on floors and roof:
10 kN/m 2
Fig. 9
-35-
o
Fig. 10
In many cases it is taken as a basis of the structural design that the core resists the
entire horizontal loading and thus completely ensures the required stability. In that
case the columns can be detailed as members hinged at top and base. The bending
moments developing at the connections are small, so that simple detailing of the
connections is possible. The statical model of the total structure can be as shown in
the figure.
Fig. 11
This sche me justifies using a si mple method of analysis. To determ ine the second order
effects the core will be se en as a partly restrained column.
The procedure of the calculations is the same as already discussed before for the
single column. The location of the stability elements in the plan has to be as effective
as possible. In this choice the structural arguments must have priority.
In practice it is of ten taken for granted that walls of stairwells and elevators are used
as shearwells and cores. In principle that is wrong, because the location of stairwells
and elevators is determined on the ground of serviceability requirements of the
building. In my opinion the stability structures must be located on the most
structurally-favourabie position in the plan. Perhaps, with a little adaptation, either
structural or architectural a combination of stairwells, elevatorshafts or wet cells can
-36-
be possible.
For the distribution of lateral forces it is assumed, that each floor can be modeUed as
an infinite rigid deep beam, elastically supported by the shear waUs. As already has
been said, the required diaphram action of the floor wil! be discussed in the next
lecture.
In general a building loaded by lateral forces wil! translate as weU as rotate.
r---
I ---
-'-'~'-B
i
, ' -". - '--.- .~'
x 1
x
,- ' x " - -
x
x
X
x
-----
I q,
II 1II I!I~
Fig. 12
Starting from the supposition mentioned and from linear elastic behaviour, the
rotation point (the centre of gravity of the fle xural stiffness EI of the stability
elements) can be determined.
In principle the horizontal displacement of each point of the plan is known and with
that the contribution of each element. Depending on the amount of stabilizing
elements and the way of determination of the flexural stiffness, the calculations can
be complex by which a computer is necessary or can be a simple short manual
calculation. By applyil'lg the latter the flexural stiffness of the elements has to be of
the same magnitude, as far as a statically indetermined situation is concerned (i. e.
minimum EI not less than 8096 of the maximum EI), because of the fact that by the
second order effect redistribution of forces occurs; the stiffer elements are loaded
higher than calculated with a linear elastic method.
In the preliminary design phase, linear elastic calculations to establish the load
distribution wil! do, af ter that for each separate element the second order
excentricities have to be determined. If the additional moments, caused by these
excentricities, are large, the sum of the moments can be redistributed among the
stabilizing elements. In this calculation the use of a fictive EI is necessary, taking into
account the non-linear behaviour of the element and also the rotation-stiffness.
An example is shown in the figure.
-37-
2 3 n
e
z
b,
Fig. 13
Z H.e b. - 8.-Z
1 1
1. 1.. b.
. 1 1
Wind: 1:~ . • H T orslOn: 2 . H.e
1 1:(1. .b. )
1 1
1 b.
Total on wall i: H. 1.. ( - + -----'1- • e)
1 Ll i 1:(1 .• b. 2 )
1 1
The stabilizing element can be a single wall, or a composition of coupled walls or even
a box-shaped element. If the bearing structure is prefabricated it is obvious to
prefabricate the stabilizing structures too.
-----1 H
D
Fig. 14
For mid-rise buildings the application of prefabricated elements for shear walls and
cores is, from the point of view of costs, attractive. Because of the limitation of the
composing elements by transport restrictions or crane capacity, the problem of joints
is introduced. Mr. Straman will discuss this later. Of course it is possible to make a
monolithic structure by means of welding re bars.
-38-
f : ~ : ~ : I f:
acceplable 10 good
II:I
occeplable
acceploble 10 good
t i ::: i I I :
struclural good
orchiteclural nol 50 good
very
• • •
• • •
acceptoble
:I
1~ ~~~~
wrong
l eb very wrong
Fig. 15
In the figure examples are shown of favourable, less favourable and incorrect
locations. A practical choice to irlcrease the vertical force is the use of post-
tensioning.
In some cases, at a ratio between the length and the width of the building < 4, the
facade structure at both ends can cooperate with the longitudinal facades. In such a
case a building structure is stabilized by an "external core", composed of strongly
connected facade elements. The norm al force diagram at the base as a result of
lateralloads perpendicular to the longitudinal facade is shown in the figure.
~ A
11 UJllD"'"
Fig. 16
IJ
"""ïl!!!Jl
By shear lack there is no linear distribution of the normal forces. The calculation of
this type is rather time-consuming. For preliminary design purposes a simple
calculation method has been developed to deal with this problem. The calculation is
based on a experiment by mind by which the facade has been modelled as a series of
equal rectangular units horizontally coupled by a shear force connection. A working-
out is given in the appendix. For a 10-storey building some results are shown in the
figure.
Fig. 17
-40-
By the horizontal loads tension as weU as pressure forces exist. As the end facade
concerned the ten sion forces are only compensated by the own weight of the elements.
However, with astrong connection between the facades, a part of the floor loads, in
the first instance carried by the longitudinal facade, will be transmitted to the end
facades. By this the tension forces are leveled, at a right length-width ratio of the
building. The effective part of the longitudinal facade amount to about half the length
. of the end facade.
The effective width will increase if the height of the building increases. With regard
to the check of the stability the first adoption can be, that all columns (the vertical
parts of the units) in the longitudinal fa ca de are equally loaded. The calculation of the
total building is reduced to a single fixed column. In the case of instability the lowest
I
g -r-"}
l l t l l J t t l l -.. l
Hi~
~ i~;
i ~ !~HJ
endfacad e long i ludinal element
focade
Fig. 18
columns are S-shaped curved. The columns of the end facades are supposed to be
rocker members. The column located on the leeside of the building is loaded with the
largest normal force; its contribution to the stability is small, which will be
compensated by the other, less loaded columns. Partial buckling is in general no
problem because the buckling-length is l/2, as the remaining columns prevent the
horizontal displacement.
-41-
INTRODUCTION
For housing as weil as for appartment and office buildings pre cast concrete elements
have found significant application.
To keep the detailing of the connections as simple as possible, beams and floor slabs
are usually simply supported.
If buildings have more than two or three storeys, the stability is of ten provided by
walls or cores, which already can be present for other (functional) considerations.
In the circumstances it is obvious to prefabricate the stabilizing walls and cores also.
However prefabrication of shear walls is not widely developed. To increase the
application, the following items are important:
The three above mentioned items lead to an investigation, carried out by Delft
University of Technology.
q.
1.1 Horizontal joints t f I IJ I1
these shear forces the connection between the elements deforms, so that the
displacements of these elements on either sides of the joints are not equal (fig. 2).
t L K
6
:u~
- - - ,,
"tI ·',
,
i :
6,
I
6u 6
Fig. 2 Fig. 3
OIIlm
ultimate values can be
distinguished (fig. 4):
which agrees with a
monolithic wall 6
10t'LL
K -> 0, which agrees with
the separa te parts
\J \J
K. O
10
The value of K of aconnection will always range between these ultimate values.
The joints, as the weakest link within the system, are usually the determining factor in
establishing structural behaviour. Therefore in order to judge the structural behaviour
of the entire system, the behaviour of the joint must be known.
2. TYPES OF CONNECTIONS
~ The monolithic connection (fig. Sa)
The monolithic connection is an imaginary connection, existing of the same material
as the material of the elements.
This connection has been investigated by Mattock and others [1,2].
Typicalload-slip curves for specimens without cracks before testing are shown in Fig.
Sb. No movement can be detected in the initially uncracked specimen until tension
cracks become visible at shear stresses.
:~
established by a line from 0 to
the starting point of the curve.
Kl' (the uncracked behaviour)
is approximately -200 3 /
Kl
N/mm 3 o Ku
Ku' (the ultimate shear- o 0.2 OA 0.6 Q~
ij lmml
stiffness) is -20-40 N/mm 3
Fig. 5
---PJ -- J
sides of the elements. Loop rein-
_~ a
:IJ [CEl}
@~}
Fig. 7
ï- - _ .- - '1'
I ,
J
"I I" I
I
I .JlH
I
1. monollllc
I
.r'?~
. I
1,0 2. toolh shapl'd I I
I
~O
j- _ . _- _ .- - + "
IlH I
~O
~O
~ ol
~O
2,0 ~ OMLR =J ol
1,15
~ ~
\0
o~o70.1:-:."-,:': .' -,p- -
" --"-
o Q2 Ol- 0,6 0,8 1,0 1,2
6 1"""1
6 !"""J
Fig. 8 Fig. 9
- cast-in;
- cast-in;
- on site mounted.
For these variants the load-deflection curves have been established. The cast-in plate
(a) gives the best results as shown in fig. 9.
-46-
[kNJ
'000
00'
i '00
_ . ~ -I "
>0,
~
"~',I;-:::-
'.'-:,-:-
,. ""
".'---"
,. -
,.. -
. .. --~ _ _ _ _ .c'
Fig. 10
Numerical simulation of the two last mentioned connections have been made with the
help of the finite element program DIANA.
K
eq
where:
Kdis = shear stiffness of the discrete connection
a center to center distance connections
d wall thickness
a) The shear stiffness of the connection itself with the stiffness of that part of the
wall, of which the deformations are affected by the connection: kv'
b) The behaviour of the wall in consequence of the concentrated loads: kc'
-47-
If the pIate connections as wel! as the tube connections are applied under the
following circumstances:
d=250mm
a = 1500 mm
E = 30000 N/mm 2
Some values of different types of connections applied under the same circumstances,
are given below.
type of connection Kl Ku
monolithic 200 25-40
smooth 2,7 2,7
tooth-shaped 15,0 5,0
steel plate 0,35 center to center 1500 mm
steel tube 2,0 center to center 1500 mm
shear stiffness K
location and amount of joints
modulus of elasticity of the elements E
-48-
100
110
90
80
70
lil
lil
:: ,i \
11
ffil
110
100
90
80
70 TIl]]]@)
-
'0
'0 \\ \ \
JO \\ JO
zo \. "'- 'zo
\'.
'0 "::-":' ::-.:;:;.:::-:-.:::-:-=....-:- -=.. '0
Fig. 11 Fig. 12
For arestraint wall, with one vertical joint, the increase of deflection is shown with
different values of K, Hand E.
It is evident that:
The location of the joints is one of the main parameters as shown in fig. 12.
M has the lowest value if the joints are located as far as possible of the neutral axis.
Bending stresses
The additional stress can be expressed by:
Ok
110 [0- - 1] 100%
m
110
'00
90
80
70
60
115
50
ilO
40
30
20
'0
QllSb Q2Sb Q37!b Q.'b
Fig. 13 Fig. 14
The maximum stress in walls and cores with more joins occur:
Fig. 14 shows that the most unfavourable location of the joint is on one third of the
width of the wal1.
DS
Tm=~
m
Tm has a maximum for x = O. Ir the wall contains a vertical joint, the shear stress in
the joint is given by the two curved lines.
It can be seen that:
Q5H
·[JJl F
INI
7,5
61mml
Fig. 15 Fig. 16
The curve of the shear stress has changed. As soon as Tl is reached, T does not
increase at that point (fig. 17).
-51-
I
Foundation
Fig. 18 Fig. 19
cellar wall;
individual foundation;
continuous foundation.
4. DESIGN METHODS
For the determination of the forces in walls and cores with joints some methods have
been used:
-52-
differential equations; if the amount of joints increases, the formulas become very
complica ted;
fini te element method; a model has been developed, with which the behaviour of
walls and cores can be established. That concerns a linear as weIl as a non-linear
behaviour of the joint;
by deriving an equivalent shear-stiffness of the joint Keq , with which the non-
linear behaviour of the joint can be taken into account. With the help of this
method it is possible to use existing analytical calculation methods;
by means of design charts; the wall with one vertical joint is almost similar to a
wall with lintels. To this problem attention is paid among others by Coull [9]. In
this method the results are shown by means of simple design charts;
if the construction is more complicated, for instance walls and cores with more
than one vertical joint, an other numerical approach has been proposed.
The wall or core can be schematized as a frame, which can be calculated with the help
of computer programs for frames. Fig. 20 shows the deformation of a wall by lateral
loads. The different parts of the wall can be conceived as bars with flexural stiffness
EI and strain stiffness EA. To transmit the shear load rigid bars are necessary.
5. CONCLUSIONS
1) General
Stabilizing structures can be executed as prefabricated structures.
The type of joint and the location of the joint determine the extra deformation,
and stresses in a large measure.
Only with a good design, these extra deformations and stresses can be negligible
smal!.
2) Connections
Continuous connections, as tooth shaped reinforced connections, give good results
with respect of strength and stiffness.
Discrete connections are less stiff. To determine the amount of connections, the
stiffness is the decisive criteria. By applying more connections in the lower floors
than in the upper floors, an economical solution can be reached.
-53-
3) Design methods
a) Walls and cores with vertical joints can be calculated with the finite element
method with the help of DIANA.
REFERENCES
[1] HOFBECK, J.A., IBRAHIM, I.O., MATTOCK, A.H., Shear transfer in reinforced
concrete, PCI Journal, 1969, February.
[2] MATTOCK, A.H., HAWKINS, N.M., Shear transfer in reinforced concrete, PCI
Journal, 1972, March/ April.
[3] POMMERET, M., La résistance aux efforts tangents des joints verticaux entre
grands panneaux préfabriqués coplanaires, Service d'étude des structures, 1971, juin.
[5] MEHLHORN, G., SCHWING, H., BERG, K., Tragverhalten von aus Fertigteilen
zusammengesetzten Schei ben und Versuche zur Schubtragfähigkeit verzahnten Fugen,
Deutscher Ausschuss für Stahlbeton, 1977, Heft 288.
[8] THIJS, J., SCHRAVENDEEL, P., SCHOT, F., De krachtswerking van stalen
verbindingen tussen dragende betonnen wandelementen. TU Eindhoven, 1986.
-54-
[9] COULL, A., Stresses and deflections in coupled shear walls, AC! Journal, 1967,
February.
SESSION 2
PRECASTCONCRETEFACADES
-57-
by C. den Ouden
TU-Delft, Faculty of Civil Engineering,
building science department
O. ABSTRACT
The role of the consultant or advisor in Building Science (or
Building Phy sics advisor) has been changed significantly the
last ten years. Within a larger architectural practice an
architect specialised in Building Physics, or a physicist
familiar with design support to architects played in the past
an important role when discussing about design decisions
related to topics like:
overheating of office spaces;
preventing extra heat losses in thermal bridges;
measures of preventing traffic-noise to enter a building;
minimising of contact noise between rooms;
detailing of moisture barrier within walIs, roofs and
facades.
1. INTRODUCTION
Historically the function of a facade has been to protect the
innerspace and the occupants of these spaces from influences
of the outside climate. During the period of low energy costs
(LI973) most attention was paid to influences of wind, rain,
moisture, sound and sun.
During the period of increasing energy costs (1974-1985 and
very likely > 1990) more and more attention has been paid to
extra thermal insulation of the envelope of the facade of
buildings.
Seen from the eyes of a building scientist or an advisor in
building physics one can look to the quality of a facade from
various points of view.
These points of view are:
ensuring sufficient sound insulation;
proper detailing of thermal bridges and moisture
barriers;
minimising energy costs for heating, cooling and
lighting;
ensuring of the comfort behind the envelope.
The envelope, however, cannot be isolated from the rest of
the building; it has to be considered as an integral part of
the overall design of the building. Within this paper an
overview will be presented of the measures to ensure the
quality for sound insulation; detailing of thermal bridges
and moisture barriers; minimising of energy costs and
ensuring comfort-levels.
Finally this paper will contain some guidelines for future
building design.
2. SOUND INSULATION
Many office buildings are today developed in areas where high
traffic noise exists. A typical noise-level of ..> 70 dB (A) ,
measured 2 meters in front of the facade, is not extreme in
such situations. The sound reduction index of a facade from a
building to be build in high traffic noise areas has to be
-59-
d5
n
dO
35
'"o
~ facaa e ~- 50 oB 25
20
0.1 0.5
- - --=---- - -.
l Ocef'lng
" 9
30 2. 2
m
c
20
10
freQ !HZ) •
Ti : IS"C
Figure 3 Figure 4
Calculated temperatures in facade- Calculated temperatures in
element with insulation inside facade-element with
a thermal bridge. insulation outside a
thermal bridge.
-63-
Checklist
for judging comfort conditions within working spaces in
office-buildings and measures to be taken within facades
to achief sufficient comfort conditions on the working
site within an office building.
-66-
5.3 Ventilation
With respect to ventilation it can be stated that there are
two different approaches. On the one hand the control of the
ventilation is at present possible by using e.g. variable air
volume (VAV) systems and by installing air quality sensors in
various parts of the building. This so-called demand
controlled ventilation (DCV) is thoroughly studied in the
framework of IEA-ECBCS annex 18 11 Demand controlled
ventilation" (3). On the other hand there is a tendency that
buildings can opera te with natural ventilation and only be
helped with top-cooling in extreme summer conditions,
preferably with so-called night-ventilation. This technique
to provide cooling by continuous ventilation into the
nighttime period, in order to cool the building structure
with cool night air, looks very promising. A good overview of
the developments with respect to ventilation is given in (4).
For designers and manufacturers of facades or facade elements
an interesting wall component functioning as air inlet aiming
to eliminate draught problems seems to be a novel concept for
buildings with natural ventilation, see (5).
References:
(1) Publication ISO 140/V of the International Standard
Organisation
(2) BAKER, N.V., "Experiences with simplified design . tools
for daylighting and thermal design", Proceedings of 2nd
European Conference on Architecture, 4-8 December 1989,
Paris, pp 161-164, K1uwer Academic Publishers.
(3) "Demand Controlled Ventilation Systems", state of the art
review IEA-ECBCS Annex 18, 1989.
(4) WOUTERS, P., New ventilation concepts with respect to
indoor air quality and energy conservation", Proceedings
of 2nd European Conference on Architecture, 4-8 December
1989, Paris, pp 119-125, Kluwer Academic Publishers.
(5) WERNER, G., "Outlet air inlet without draught problem",
10th A.I.T.C. conference, Dipoli, Finland, 25-28
September 1989.
(6) DEN OUDEN, C., "Building 2000", Proceedings of the 2nd
European Conference on Architecture, Paris, 4-8 December
1989, pp 665-668.
(7) Various papers in "Proceedings of 2nd European Conference
on Architecture, Paris, 4-8 December 1989, pp 234-253,
Kluwer Academic Publishers.
-69-
1. INTRODUCTION
Prefabricated concrete facade-elements have been applied as cladding for buildings
for some decades.
With precast concrete elements the architect has the opportunity to bring out the full
sculptural potential of the material and to freely express his design-concept for the
building. The architects have used this opportunity to a large extend: as weU as the
external form as the exterior finish and colour is concerned. Some examples are shown
in the slides. The result is highly based on the personal taste and preference of the
architect and on the freedom given by the fut ure proprietor.
Fig. 1
Prior to the design of elements the architect should visit one or more manufacturing
plants to become familiar with the manufacturing process, including the fabrication of
the moulds, the problems in casting and finishing, connections, job site handling etc.
Of ten the facade existed of an external concrete element and an internal filling wall
of brickwork.
-70-
Beside it, the solid pre cast concrete wall panel exists, see the figure.
Fig. 2
With the need to conserve energy it is important for the owner to give attention to
this aspect; he will endeavour to minimize the capitalized value of the building-
investments and the control of energy. Besides he wishes a good sound-insulation.
The shape of the building, the size and location of the glazed parts of the cladding
establish the level of insulation of the building. This is already discussed in the
preceeding Ie ct ure.
1
j: . . . . ' .. '. '
archilec1ural
'finish -
insulalion
Fig. 3
Improvement of the composition of the element can be reached by placing the thermal
insulation inside of the element: creating the sandwich panel in this way. Special
attention has to be paid to solid concrete edges (box-construct ion) which means the
crack
-,----------------,
Fig. 4 Fig. 5
\ /
/ :o~
/ \ "
Fig. 6 Fig. 7
However, even this solution has disadvantages. Different temperature changes (day to
day thermal expansion) between the inner and outer leaf cause different changes in
volume. Depending on the way of attachment of the skins with connectors through the
insulation layer, the elements will warp or bow, and stresses occur by restraint change
in volume of the concrete, of ten resulting in crack ing.
A further development is the application of a ventilated air cavity between the outer
layer and the insulation layer, see the figures.
Fig. 8
With respect to an insulation layer without an air cavity there are the following
advantages:
in average the external leaf stays dryer, the chance and the extend of pollution of
the concrete surface decreases; more surface finishings are possible with less risks;
in respect to building physics a (slightly) better behaviour.
but especially: a simple joint construction is possible. This will be discussed later.
-73-
there is a possibility to replace the outer leaf. The details of the mutual
connections have to be adopted to it. To this subject I will return too.
The different parts of the elements with the last discussed composition have all their
own function.
in side
infernol teal
vertical joint
·PRINCIPLE TRAN$LATION TO T WO·$I<IN El E MH
T WO STAGE JOINT CONSTRUCTIQN
It:rt21mn=tnL__
oirpressure release
drainin
u tet ic:
cJltif=l='-"-''''-'"'''''''''"",er oorr ie r(-roircoo''')
i
horizonlal joint _ .
{
Fig. 9 Fig. 10
The internal leaf can be load bearing, has thermal accumulation and sound reduction
functions and takes care of the air-tightness.
The insulation layer takes care of the therm al insulation. The cavity serves for
settlement of the air-pressure with the outside air and takes care of draining. The
external leaf is only a "raincoat", it stops the rain and of course has an estetic
function.
The composition of the element is, as a matter of course, suitable for the two-stage-
joint-construction of which the principle is shown in the figure. This joint-system
exists of two sealing-screens: the inner screen assures the air-tightness; the outer
screen assures the water-tightness. In the space between both the sealing-screens a
release space is found, which serves for the release of the air pressure and for the
draining-off of eventually penetrated moisture.
The inside part of the joint at the internal leaf can be sealed by means of mast ie,
foamtape, rubber- or synthetic profiles. This part of the joint can be checked and
maintained rather easily.
-74-
By a direct connection to the open air the atmospheric pressure and the vapour
pressure in the cavity and outside close to the facade front are equal.
3. PENETRATION OF RAIN
It is interesting, as an intermezzo, to pay attention now to the mechanism of the
penetration of the rain into a facade. The rain-water can be transported by:
a. cappillarity;
b. attached water (adhesion) carried along by an air current;
c. pumping-acti~n, pulsating-action;
d. kinetic energy of the rain-drop; for the rain-drop has a component of move ment
perpendicular to the facade.
rOlndrop
Fig. 11
The causes a and c are simply to avoid by a sufficient joint- width. Thereby is cause c
decicive; a width of eight mm will do.
Cause b can be taken care of by execution the faces of the joint draining off outwards
and by avoiding an air current directed inwards.
By the equal air pressure in the air cavity and close before the front surface, the air
current will not exist. However, close to the edges of the building in the cavity air
currents can be caused by underpressure by Karmann whirls. For the puropose to avoid
these air currents the cavity had to be closed near the building edges in vertical c.q.
horizontal way.
Only by kinetic energy of the rain-drop, water can penetrate into the cavity, via the
vertical joints. The amount is small by the small measure of the joint so that it is
acceptable.
However, if possible, it is sensible that the insulation layer behind the joints at a
certain width will have water-tight cover, e.g. in the shape of a tape.
-75-
DC lapped joint
I unnece$Qry complicatedJ
80 open joint
Fig. 12
4. CONNECTIONS
Much attention has to be paid to the connections of the externalleaf to the internal
one. These connections must be able to withstand the loads, placed on the outside skin
(i.e. due to own weight, stripping, plant handling, transportation, erection and in-situ
loadings) while difference in movements between external and internal leaf must be
possible.
connedor
PRINCIPLE
"""} ..... / \(
..... I . '
Fig. 13
-76-
-- -t- - - --.:;' ----- - -"'j.- ---- --.r--- - ---; ------ -...- -- - -- -+--
r~= === = =,,, ~ = .==~~====----===_=J
~j i!
~
I ·••
I
~ ,•
~ - - ----- - -
+ +
Lf+-- t"lxi ng
4=-,°1,_"___
Fig. 14
-77-
The flexible and non-flexible fixings must be durable. Metal to be used has to be at
least protected hot-dip galvanized steel, or better, austenitic stainless steel, indicated
as ASTMI 316.
Some designers do not trust these materials used in a cavity and choose for concrete,
thereby introducing and accepting thermal bridges.
condens~ ttm<ml'~
Fig. 15
5. MANUFACTURING
For the manufacturing of the double skin concrete elements are two methods:
1. First the outside leaf is cast in the mould, then synthetic elements (egg-box-
shaped) are placed, af ter that the insulation layer is installed and the inner skin is
cast.
in ternal leef 4
thermal insulation 3
eg9 box shap ed space r 2
external leaf
I J!j 1IIII11 1111111111111111 mouid
sequence of casting
Fig. 16
By this method of manufacturing the exposed side of the internal leaf becomes
the side to be finished in the factory. Mostly this surface will not be accepted by
the architect or principal, which means an architectural finish wil! be necessary
afterwards. The exposed, mould side of the external skin can have all the various
surface textures and finishings, used with precast concrete as discussed by Mr.
van Acker.
2. The inside and outside skine are cast in separate moulds. The inside leaf is cast on
-78-
demoulding
lurning over
p lac i ng in fr esh-ca sI exlernal leaf
faslenings r-"
~ 2 :- - - - - - - - - - - - --:
::: P!:::~
I I
insulalion
inlernal leaf Cf: :::::~-r p~ -- - - - ,~,,~ - -c ~,~ ~;
exlernal leaf
DAY 1 DAY 2
Fig. 17
Fig. 18
-79-
6. LOAD-BEARING OR NON-LOAD-BEARING
The double-skin elements can be applied as load-bearing or non-Ioad-bearing elements.
Usually the structural leaf is placed on the inside. For what I wanted to say about the
double-skin elements, the difference in function does not matter.
I have not paid attention to variation in dimension and sizes, to fitting, to horizontal
joints and to connections to the bearing structure or the floor slab, although there is
much to teIl about these subjects. Perhaps next time.
-31-
Arto Suikka
Development Manager
M. Sc. Eng.
Partek Concrete Industry
Finland
1. GENERAL
At the moment all the following products are in the product ion
programme, and they are briefly explained here.
-82-
L..J
Bevestiging boven
r Bevestiging onder
3. VARIAX-SANDWICH
The panels are made on the same 90-110 m long casting lines
as hollow-core slabs, the inner layer being 150 mm thick
prestressed hollow-core.
1200
6. PALAZZO APPLICATIONS
7. CONCLUSION
with weIl designed solutions the total costs can be lowered and
simultaneously freedom of architectural design, e.g. such as
combinations of materiaIs, colours and surfaces, can be added.
-89-
Arie Gerritse
Hollandsche Beton Groep; R&D Department
Rijswijk The Netherlands
1. INTRODUCTION
1.1 Scope
Fibre reinforced cement (or concrete) indicates a range of very versati-
Ie materiais, which do enable us to produce th in (although strong),
relative light (although large), shaped elements. A vast range of high
strength fibres is available (fig. 1). Cost arguments and ease of
handling (spray technique) led to chopped glass fibres as being commonly
used. This is clearly expressed by the vast developments in nice and
representative facade elements manufactured with one of the GRC (Glass
Reinforeed Cement) variants. Prefabricated glass-reinforced cement
elements have given a remarkable impulse to today's architecture.
Since new developments inherently bring about some cases of less
adequate behaviour due to insufficient experience or insufficiently
(r~garded) knowledge a Dutch Stupré-committee*) presented in 1988 a
report on Design rules for architectural panels in GRC [1].
This paperwill mainly report on the study that carried out by the
Stupré-committee 30 "Glass fibre reinforeed cement". The recommendations
and practical advices differ from the existing guides on the issue:
- In addition to strength and stiffness, the ultimate strain of the
material is introduced as being a critical design parameter.
The report itself is also intended to act as a basic guide for FIP
recommendations, to be prepared in the near future by the FIP Commission
on Prefabrication (working group on Thin Walled Units).
1.2Aims
Suitable and economical application of prefabricated architectural GRC
elements will largely depend up on the designers knowledge of the
material properties, manufacture and erection techniques as weIl as the
time related behaviour of the elements.
Since the material properties are time dependent (ageing), it is
important to know how design values are to be determined. The Stupré
recommendations are intended to provide the designer, in a practical
way, with the basic data necessary and the information about relevant
properties of various systems of GRC.
1.3Field of application
These recommendations concern the application of GRC in prefabricated
elements, utilized inside as weIl as outside a building. The recommen-
dations apply to single skin elements (provided or not with stiffening
ribs or a steelframe), to sandwich elements and also to so called
composite elements. In order to achieve sufficient durability in a
cement matrix, use can be made of alkali resistant glass fibre (AR GRC)
or E-glass with polymer admixtures, thus polymer glass fibre reinforced
cement (PGRC) or alternatively a combination of these materiais.
The recommendations only apply in so far as manufacture is carried out
by means of a spray procedure.
The current version of the Stupré report does not contain subjects like
manufacture procedures or erection. For the time being it was assumed
that these items are sufficiently discussed in other literature, e.g.
SBR publication nr. 168 - April 1988 - "Gevelelementen van Glasvezel-
versterkt cement" [2]. (In Dutch).
l)n_HS I I I SI
3000 ' _ , _ _ .rc0rbo~ _ _ /9\055_
I
.
oromid / oromidI
HM ARh stondo rd
~h
prestressing s~~
2 3 I. 5
strain(%) -
20
/'
relt I
I
1 1.5
Direct tension
M.a.n.
3U 3U fibres in vol. %
M.a.H. 28 days
2U
401--l-____-=I=32~%~A~rn~mêid·=FGO~'C~.~N.Wg;.==4==:'"'_ _ _+-
o
MOR ~~~~==~-SAr~•.!l!.~I! .: .:.0.0%60 +2O C
30
...
E
~201-+--------~--~~~~--~~==~~---
z .........
10
1 2 10 20 50 years
2 5 lQ 20 50 100 200 500 10 104 days
Percentage in vol. %
Forton with E glass
Long term behaviour of individual cement composites
10
2 10 20 50 years
2 5 10 20 5U 100 200 500 lOl 10 4 days
equal to 2.5 for GRC, respectively 1.5 for PGRC) and the long term
ultimate strain capacity with the short term ultimate strain capacity
(factor 2511 for GRC, resp. factor 2 . 5 for PGRC) See fig. 2 and table 1.
This substantially greater decline of ultimate strain capacity of an
ageing material has appeared in practice - when not properly looked at -
to cause various damages.
It i s therefore of essential importance to design the elements taking
long term ultimate strain into account (7), (9).
-93-
Bending tensile
strength 26 ± 5 13 ± 2 22 ± 3 18 ± 2 N/mmZ
LOP in bending 9 ± 1,5 10 ± 2 13 ± 2 17 ± 2 N/mmZ
Breakage strain
in bending 10 ± 2 0,7±0,1 4 ± 1 2 ± 0,5 0/00
Tensile strength 9,5± 1,5 5,5±0,5 10± 1 9 ± 1 N/mmZ
LOP in tension 6 ± 1 5,5±0,5 6 ± 1 8 ± 1 N/mmZ
Breakage strain
in tension 10 ± 2 0,4±0,1 3 ± 1 1 ± 0,5 0/00
Modulus of
elasticity 15 ± 5 25±3 13± 1 16± 2 kN/mmz
Density 2 ± 0.1 2±0.1 2.0 2.0 kg/dm3
2. BACKGROUND
2.1Glass fibres in cementitious matrix
The lower strain , capacity af ter ageing, and thus the consequently lower
ductility of aged GRC elements, has caused several cases of less
adequate behaviour (7). The decrease in strain capacity is mainly due to
2 main causes:
a)- Glass itself is vulnerable to alkaline attack -> loss of strength
b)- Continued hydration of cement will induce a strong bond between a
slightly etched fibre surface and the matrix.
The Stupré report does not enter this in detail, however the author
wishes to express his view that cause b) is more relevant.
ad a) The stresses in the fibres up to cracking of concrete are
negligable and SEM photographs show, also af ter prolonged
exposition, nearly never serious harm to the fibres.
ad b) A reinforcement can only take force (be stressed) if it is
given the necessary elongation. Any transfer of force (also in
the well known steel reinforced concrete) from matrix to
reinforcement requires some lenght of slip to accomodate for
some free length to reach the elongation, which belongs to the
stress. Compare the stress-strain curve. But if, due to high
bond (no slip) this elongtion (at cracking) is made impossible ',
than brittle failure will occur (fig. 4 and 5)
Basic advantage of fibre reinforcement in the post-cracking behaviour.
-94-
..
hold slip I ol slip • I
ol
hold
elongation possible
3.DESIGN CRITERIA
3.1 General .
For architectural panels made of GRC a number of design criteria are of
particular importance:
1. strength (in bending andfor direct tension)
2. deformation f flexure
3. freedom of movement
4. durability and fracture strain
5. resistance to pollution
6. watertightness
7. behaviour in fire
8. thermal resistance
9. dimensions f tolerances
10. resistance to impact
In this paper special attention is given to the criteria about strength,
deflection. freedom of movement and durability. The latter expressed as
strain at failure.
-95-
3.2Strength
A GRC component must, exactly like other structures, be designed accor-
ding to the valid national recommendations.
Account must be taken of stresses in relation to prevented deformations
for example in connection with variations in the temperature and
moisture movement in an element. According to this criterion, it can be
established under what conditions no fracture or collapse of the element
should occur or what conditions any possible cracking behaviour must
satisfy.
Class Stro 2:
Strength requirement is equal to class Stro 1; however plastic behaviour
at fai1ure at the aged stage is not required.
C1ass Stro 3:
The requirement is simi1ar to that in Stro 1, it applies however only
for the young materia1.
3.3 Deformation/Deflection
For the time being the limit to deformations due to the combined effects
of the design values of external loading and thermal and hygric influen-
ces are restricted by a standards requirement of l/300.
-96-
3.4Freedom of movement
The elements must be designed in such a way, that movements due to
variations of temperature and moisture can take place as freely as
possible (inner restraints includedl).
This induces some restrictions to the design of especially sandwich ele-
ments and composite elements, but also of single skin elements having a
deep relief because of their stiffness. Connections must be designed in
such a way, that possible movements of the bearing construction are not
transmitted to the GRC element.
3.6Classification
Table 2 specifies the required classes for field of applications
frequentlyoccurring.
Table 2: Classification
- --,---- -
field of use Strenght Deformation Durability
4. CALCULATION
4 . 1General
The most critical loadings for designing of GRC elements are tension and
flexure, due to external loadings andjor restrained movements . Table 3
shows some indicative characteristic values of the parameters relevant
to this types of material and loading.
These values determined for young and aged material -where in the table
3 the lowest value of the two is shown- provide some indications as to
the properties. The bending properties are determined by test at four
points.
Other load cases such as compressive strength and inplane, interlaminar
and punching shear are usually less limitating. If design calculations
show, that one of these load conditions does really become limitating,
additional information will then be necessary.
4.2Partial coefficients
In more recent structural codes (Eurocode), the following relation has
tobe satisfied in the ultimate limit states to be considered:
Design for aged GRC should therefore be based for practical reasons upon
the following design rules:
This implies:
for BI: 1. 1.8 and for B2: 1m 1 (see 3.4)
EUk
The latter equation imp1ying : Etot S 1.8
in which
E~t - the strain to be calculated due to external loads and res-
trained deformations.
EUk = the characteristic strain (capacity) of the material
The latter value can be taken from values measured on the aged material.
This may imply, that GRC (and especially AR-GRC) may find their applica-
tion in elements, with very slight restraint of deformations. PGRC
presents a bit more favourable behaviour thanks to its greater strain
capacity.
4.5Composite elements
Such elements have one skin made of GRC and the other bearing skin of
prefabricated concrete.
As air space, insulation or both air space and insulation are usually
placed between these two skins.
For this type of elements the same recommendations apply as indicated
above for sandwich elements.
-102-
5. CONCLUDING REHARKS
Glass-fibres are extensively used producing attractive thin-walled
(light weight) prefabricated facade relements. The specific consequences
of glass-fibres in connection with a cement matrix are to be considered
carefully in design, e.g. by imposing restrictions to the required
strain capacity. The FIP-commission on Prefabrication (working group on
Thin-Walled Unites) will welcome any comment.
REPERENCES
1. STRUPRE-report studie groep 30.
Glass fibre reinforced cement (GRC).
Design rules for . architectural panels.
Dutch edition 1988;
English translation avai1ab1e from Dutch Concrete Society at Gouda
3. GERRITSE, A.
Typen en eigenschappen van vezelversterkte cement composieten.
Cement; 1982 No. 8 (In Dutch).
4. STUPRE-report.
Bevestigingsconstructies van geprefabriceerde gevelelementen.
(Connections of precast facade elements)
Only in Dutch; Dutch Concrete Society at Gouda (1975).
7. MOORE, J.F.A.
The use of glass-fibre-reinforced cement in cladding panels.
BRE report- BRE lP 5/84.
8. JACOBS, J.N.
Strain Capacity as a design criterium.
Proceedings of Symposium on GRC Darmstadt Oct. 1985.
9. VAMBERSKY, J.N.A.J.
Bemessungsregeln fUr architectonische elementen am Glasfaser beton
Beton + Fertigteiltechnik Heft 7, 1989
-103-
There was a search for more expressive facade design and the
enormous possibilities of concrete in design, colour and
texture were discovered. A not to be despized detail was that
quality was payed, partially because hard competition didn't
exist. It was the time of the goose that laid the golden egg.
Formwork
Smooth concrete surfaces are both the least expensive and the
most difficult to produce finishes.
The aim during demoulding is to obtain a best possible
concrete surface,
without air-bubbles, which means an adapted mix
composition, and an optimal compaction
- prevention of al&y possible damages, e.g. through sufficient
slope (8 to 10%), of vertically aligned adjoining formwork
components and by rounding all corners and edges.
- obtain a surface with uniform glosse
-108-
The polishing method has long been applied for tooling natural
stone. In the sixties it was in rare instances used to obtain
archi tectural concrete and is now ga~n~ng increasing
popularity. Use of automated processing techniques have made
the method less costly.
-111-
Conclusion
SESSION 3
1. INTRODUCTION
The recommended ru1es of design for insitu cast concrete
structures apply also to precast and composite
construction as modified or supplemented by EC2 Part lB.
Particular areas of design which require special
consideration not usually carried out in the design of
cast insitu structures are:
- bearings
- joints
- multi stage construct ion
These and other element types and design areas are
considered in EC2 Part lB.
2. BASIS OF DESIGN
A I FRAMe STRUCTURe'
BI LOAOBEARING WALLS
1 transV~H
L longitud ....
V vertieal
~Ipa<tl
P per ipheral
-122-
Cl SKELETAL STRUCTURE
/
-r- - ,,
/
/
/
,,
/ ,
Ol FLOOR OR ROOF OIAFHRAOIol
TH
J
'bnet
W H Ibne:t
FIG A' VERllCAI. SUSPENSION REINFtJRCeENl FIG 9' INCl.INED SUSPEHSION REIHRlRCEMEHT
3. MATERIALS
4.1.3.3.Concrete cover
The minimum value of cover is determined taking into
account specific parameters of precasting such as:
quali ty control, use of pretensioned tendons , concrete
grade, corrosion resistant reinforcement etc.
-126-
4.5.5 Bearings
The integrity of bearings for precast members shall be
ensured by effectiveness of reinforcement in the element
below and above the bearing, by arestraint against loss
of bearing through movement and by suitable limitation of
the bearing stress.
The different parameters intervening in the design of
bearings are treated in a step by step way in which
allowances for various effects are added to a basic net
bearing surface.
-127-
5. DETAILING PROVISIONS
JA den Vijl
Delft University of Technology
Stevin Laboratory
1.SUMMARY
For pre-tensioned strands the transmission length as weU as the anchorage capacity
shall be verified. For tensile actions in the transmission zone the lower bound value
of the transmission length is decisive, whereas the resistance against moment or shear
failure is estimated by its upper bound limit. Since the bond strength is influenced by
the transverse deformations of the strand, distinction is made between the anchorage
capacity within and beyond the transmission length. To prevent bond deterioration
due to concrete splitting, a minimum concrete cover is required. A proposal for the
new CEB-FIP Model Code, in which these aspects are included, is discussed.
At first, bond between strands and surrounding concrete is due to adhesion and
mechanica I interlocking between cement stone and strand surface, but after a small
relative displacement this initia I bond is broken and another mechanism is activated.
Like for plain bars this is rather based on dry friction than on shear, which plays a
decisive part with rib bed bars; viz. a strand that is displaced through the concrete does
not shear off the concrete ribs moulded by the strand, but it follows the spiral channel.
lack aftit
Between the six outer wires of a strand some space is left because of the larger
diameter of the center wire; see Fig. 1. The distribution of this free space along the
perimeter of the strand differs in subsequent cross-sections. Hence, the shape of those
cross-sections is not similar and, therefore, a relative displacement (slip 0) of a strand
-130-
has a wedging action on the concrete. The magnitude of th is "lack of fit" effect is
proportional to the slip.
Poisson effect
The transverse deformations connected to changes of the longitudinal steel stress ö.ap
have a significant influence on the bond resistance. This is clearly demonstrated by
the different results of a pull-out test and a push-in test, as is discussed hereafter.
Pitch effect
Just like with a curved prestressing tendon, the frictional resistance may be increased
by the contact stress es at the bent; see Fig. 2. As this effect occurs both under
compression and tension, it wil! be proportional to the absolute value of the longitudi-
nal steel stress change.
dPx + cfPxdlj) =0
x= 0 - Px =Po
lIP= Po - Px = Po (l _e - cfIP )
The effect of steel stress changes on the bond resistance of strands was investigated
with the help of pull-out and push-in tests [12]; see Fig. 3. With the latter one the
conditions along the transmission length during release are simulated: the decreasing
steel stress is connected to an increase of the strand diameter. The pull-out test reflects
the situation when the steel stress is increased and, connected to that, the strand
diameter becomes smaller. As both tests were performed with a relatively short
embedment length, additional forces we re superimposed to obtain the required steel
-131-
stress changes. The results of these tests are outlined in Fig. 4. It is c\early shown th at
the bond stress is not defined by the slip alone, but that the steel stress change
corresponding to a specific loading case has to be taken into account.
t,vp.A p
T! i i i
tb·n;·r/l·lb t t,vp·Ap
tb [MPol
8~-----,-------'-----'
t,vp - - 1200MPo
6r-------+------=~~----~
t,vp_+ 1200MPo
2r-------+-------4-------~
fee = 50MPo
o 2 3
5[mml
Fig. 4 - Bond resistance as a function of slip and
steel stress change
The response of the surrounding concrete to the wedging action of a strand, being a
combination of the afore-mentioned lack of fit effect and the Poisson effect, was
studied by means of a two-dimensional numeri cal analysis of the stresses and strains
around an expanding bar [13], taking into account the non-linear behaviour of
concrete loaded in tension, which is characterized by a softening branche after having
reached the tensile strength; see Fig. 5. Although this study was primarily meant to
-132-
estimate the concrete cover required to prevent splitting, it also gave a better insight
in the magnitude of the radial stress es and, connected to th at, in the distribution of
the bond stresses along the embedment length.
2D-model
//"'.",.------ ............. ,
/
/ ,,
.+.__ /'
\
\
~
.~._-_.+.
~Vt
~
'\
I
j
·- x
\, .--/
/
/
'...... -///
c I <IJ c
X ~llWlill4===~llilllliilmw X
o
Er
Fig. 5a - Model for the analysis of stresses and strains around an expanding bar
and stress distribution for linear elastic material properties.
ii - from the Poisson effect alone the maximum radial deformation along the
transmission length can be estimated t~:
Hence, for practical cases a considerable part of the concrete along the trans-
mission length is in the non-linear stage, and this explains why the bond stress
distribution is rather uniform than linear; see Fig. 7;
Fig. 7 - Steel strain distribution along the Fig. 8 - Maximum radial stress in thick wal!
transmission length cilinder with uniform tensile stress
iii - when compared with the response of an internally loaded thick wall cilin-
der with a uniform tensile stress distribution across the wall (ar/fel = 2c/cp ;
see Fig.8), the investigated cases appear to have at least the same resistance.
This means th at the thick wall cilinder approach is conservative because a
larger portion of the surrounding concrete is contributing; see also Section 4.
-134-
It is remarked tb at the aforesaid is only a rough approach of the real situation. On the
one hand not all actions such as "lack of fit" and clamping of crushed cement stone
particles were considered and on the other hand the effect of the "soft" mortar layer
just around tbe strand was neglected. The composition and compaction of tbe latter
is supposed to have a considerable effect on tbe bond resistance, which is also shown
by tbe increase of the transmission length as the distance from the strand to tbe bottom
of the member is larger during casting
The bond of strands is susceptible to varia ti ons of the different parameters involved,
as appears from the scatter th at is connected with e.g. the transmission length. The
characteristic lower and upper bound differ a factor two, in genera1[3].
On the basis of the description of the bond mechanism of strands given in the
preceding chapter, a proposal for the CEB-FIP Model Code 1990 [15] was derived.
This proposal takes into account the following aspects:
- a uniform bond stress distribution;
11 - the effect of the load case (pull-out or push-in);
2000Gp [MPa)
fpd - - +
1 500 ~----~-----+----~~----~~~-+----~
~~~~--~~~------~
o- trans miss ion leng th
40 80 120 240
\ Iq,
Gp [MPa)
2400
1800
0
0 -,
•
o'
=
••
/
~
o .t. •
<Sf .t.
o 00
1200
its Anderso n 151
v - slip
600
'/
(b
T - pull-out
Walraven IBI
• - pull - out. strand !raet.
Te'eni 191
o - pull - oul
strand, see Fig. 9, is parallel to the line that represents the anchorage capacity of a
non-pretensioned one.
It is remarked th at for the calculation of the transfer length the steel stress just after
release is taken into account, but for the anchorage capacity the effective steel stress,
thus including all prestress losses, is used. It is supposed, however, that the trans-
mission length remains constant in the course of time.
Op [MPol
2400r------.-------.-------.------~----~
1200 I--------+'~Hrl_...::y.""-"~__l
IN>
Fig. 11 - Anchorage capacity ofpretensioned strands in beams and beam-ends
For the verification of the proposed design rule for the anchorage capacity of strands,
results of tests on hollow-core slabs and tests on beams and beam-ends were used; see
Fig. 10 and Fig. 11, respectively. It is shown that the proposed rule represents a lower
boundary except for beam tests with large embedment length. Since those results
contradiet the results·with shorter embedment length out of the same test series, this
anomaly is accepted.
The magnitude of the circumferential tensile stresses can be estimated with the
aforesaid thick wall cilinder approach. The internalloading of the cilinder consists of
-137-
l{ t
ribbed bars e. 30°_45.
e strands e" 72·
their resultant and the tendon axis. For
ribbed bars 8 varies from 30° to 45°, but for
strands holds 8 = 12°, as follows from the
assumed coefficient of friction
Cf = 0,32 (8 = 90° - bgtan 0,32); see Fig. 12.
Fig. 12 - Relation bet ween bond stress Tests with ribbed bars we re evaluated with
and radial stress the help of this model, considering different
distributions of the circumferential tensile
stresses across the cover and under the assumption of 8 = 45° [7]. From Fig. 13 foIIows
that the cracked-elastic distribution yields a safe lower bound and th at the plastic one
covers the upper bound of the results. At first sight, the latter seems to contradict the
findings mentioned in section 2.2, where it was stated th at the thick waII cilinder
approach yields conservative results. The difference is, however, that in section 2.2
only the maximum radial stress was considered, whereas in the present case the
ave rage radial stress along the embedded length should be taken into account. In the
foIIowing analysis the latter is done by deriving the radial stress from the average
transmission length.
°r/fc!
6r-~--~---r--,,--.---.
1
"\
D. ordinary concrete
o tightweight conrete
5 6
cl</>
Applied to strands, the plastic stress model can be used to find an expression for the
concrete cover required to prevent splitting. The model assumes that splitting occurs
wh en the ave rage circumferential tensile stress equals the tensile strength:
(1) at = fctj
with:
(2) at = cp/2c . ar
(3) ar = rblCf, with Cf = 0,32 [10]
(4) rb = PJ(nq; lt)
(5) ltm = 10 q; (ap Jfccj)O,5 [3]
(6) fctj = 0,46(fccj)O,5 [4]
(7) Ap = 7136 n q;2 , for strands this yields:
(8) c/q; = 0,066 (ap i)O,5
[:1 [-I
Or Ifct
6
~.J t~l
c/rfJ = 2
s/rfJ = 1 I
5
---~ __ .....1
0
ril [og: Eip! 8J
.__ -.J ---~
9 -
2 3
"
Er [%01
Er=1%0 2%0 3%0
Until now only the situation around a single strand was considered, but the spacing
between strands may be so smal! that the splitting action is increased. This is illustrated
in Fig. 14, where the ave rage radial stress around the tendon nearest to the edge of
the model, is given as a function of the radial deformation of th at tendon. It can be
concluded that both the maximum radial stress and the course after the top are
-139-
ta
Fig. 15 - Occurrence of splitting cracks as a function of concrete cover and clear strand
spacing observed in pretensioned members with 3 strands (a) and 1 or 2 strands (b)
influenced by the number of tendons and their configuration with respect to the edge.
However, there is no model available that adequately describes this interaction and
for that reason design rules can only be based on the limited experimental evidence
available.
Tests to estimate the possibility of splitting cracks as a function of the concrete cover
and the strand spacing are described in [10] and [16] and the results are summarized
in Fig. 15. Fig. 15a displays the resuIts of tests on pretensioned beam-ends with three
strands q; 12,5 mrn and an average cu be strength at release of 44,5 MPa. The results of
tests on beams with one or two strands q;9,3 mm or q;12,5 mm are shown in Fig. 15b.
Release took place after one or after two days at an average cube strength of 33 MPa
and 44 MPa, respectively. This different cube strength at release did not noticeably
influence the occurrence of splitting cracks. Except for cover c/ifJ = 3 and spacing
slq; = 2, both series show similar results. On the basis of these experiments a proposal
for the minimum cover as a function of the clear spacing - for members without
confining reinforcement - was formulated [15].
The foregoing sections focussed on the bond behaviour of smooth seven- wire strands,
and two aspects, th at were mentioned, are found back in a new code proposal: the
effect of the loading case on the average bond resistance (pull-out versus push-in) and
the large scatter with which bond of strands is connected. Now the question may be
raised whether these aspects hold to the same extent for other types of tendons, su eh
as indented wires and indented seven-wire strands.
tm/fee
0.5
~o number of lesls
fee: L2, 55MPa
0.4 r/J :.6mm
/
/
0.3 auolable ronge
/..-;
0.2
rV,.,j
0.1 -I '""-
o
cr~
4 8
~12 ~16 20
profile factor x 10 3 [mml
Fig. 16 - Average bond resÎstance of indented wires
as a function of the profile factor
Because experimental data in this respect are lacking, this question can only roughly
be answered. As the deformations at the steel surface become larger, the part of the
bond delivered by shear increases. Compared to the friction me eh ani sm, the shear
mechanism seems to be less sensible to the loading case (pull-out versus push-in) and
the quality of the th in mortar layer just around the tendon. On the other hand, the
scatter in bond resistance due to the allowable dimensional variations of indented
wires is considerable [11]. Fig. 16 shows th at within the allowable range of profile
depth the bond stress may vary with a factor 2 Cr mis the average of the bond stress at
a relative displacement of 0,01 , 0,1 and 1,0 mm, respectively). Because of these
considerations the effect of the loading case on the average bond stress of indented
wires is estimated lower than in the case of smooth strands, whereas the effect of the
scatter is taken equal in both cases, see [15] .
-141-
REFERENCES
1. HANS ON, N.W., P.H. KAAR, Flexural bond tests of pretensioned prestressed beams,
ACI Journal, Vol. 30, No. 7, Jan. 1959.
2. KAAR, P.H., N.W. HANSON, Bond fatigue tests of beams simulating pretensioned concrete
crossties, PCI Journal, Vol. 20, No. 5, Sept.-Oct. 1975.
3. OLESNIEWICZ, A, Statistical evaluation of transmission length of strands, Bistyp, Warsaw 1975.
4. ACI Standard 318-77.
5. ANDERSON, AR., R.G. ANDERSON, An assurance criterion for flexural bond in pretensioned
hollw core units, ACI Journal, August 1976
6. DEN VIJL, J.A, Force transfer between concrete and prestressing strands (in Dutch),
TV Delft, Stevin Laboratory, Report 5-78-6.
7. TEPFERS, R., Lapped tensile reinforcement splices, J. Struc. Div. AS CE, Vol. 108, No. sn,
January 1982, pp. 283-301.
8. WALRAVEN J.C., W.P.M. MERKS, The bearing capacity of prestressed hollow core siabs,
Heron, Vol. 28, No. 3.
9. Te'eni, M., Flexural bond and face end slip in hollow core units, Spancrete of Israel, Technical
Report, August 1984.
10. DEN VIJL, J.A., Bond proper ties of strands in connection with transmission zone cracks,
Betonwerk + Fertigteil-Technik, Heft 1/1985, pp. 28-36.
11. DEN VIJL, J.A, Comparative study of bond properties of different types of indented wire,
TU Delft, Stevin Laboratory, Report 5-85-10.
12. DEN VIJL, J.A, Transmission length of strands in lightweight concrete (in Dutch), TV Delft,
Stevin Laboratory, Report 5-86-17.
13. DANTVMA, W.F., J.A DEN VIJL, Stresses and crack formation around an expanding bar
(in Dutch), Cement 1988 nr.2, pp. 56-59.
14. VAN DER MAREL, AP.,The shear capacity of pretensioned hollow-core si abs (in Dutch),
TV Delft, Stevin Laboratory, Report 25-88-10, June 1988.
15. CEB-FIP Model Code 1990. First draft. CEB Bulletin nrs. 195 and 196, March 1990.
16. DEN VIJL, J.A, Effect of concrete cover and strand spacing on bond splitting in a pretensioned
member, TU Delft, Stevin Laboratory, Report in preparation.
LOAD DISTRIBUTION AND FAILURE BEHAVIOUR OF PRESTRESSED HOLLOW
CORE SLABS
INTRODUCTION
Prestressed hollow core slabs are nowadays extensively used in prefabricated
buildings. The units can quickly be assembIed to form slab fields, by simply filling
the joints between them. The slab units can be manufactured by various methods:
the most popular methods nowadays are extrusion and slip forming.
The manufacturing process starts with prestressing of the strands on a long line
casting bed. Af ter prestressing, the concrete is cast. As soon as the concrete has
reached a sufficient strength, the prestress is applied and the continuous bed is
sawn-cut to slab units of the desired length. The slab units are special in this
respect that they do not contain either shear reinforcement, nor transverse rein-
forcement, so that weU defined procedures for design and quality control are
necessary. This paper discusses the bearing capacity of slab elements and its con-
trol, and the behaviour of slabs assembIed in larger slab fields.
-144-
M
o = A ( _x + V ) (1)
px p z x
where
stress in the prestressing steel at a distance x from the support
= cross-sectional area of the prestressing steel
inner lever arm
= moment and shear in section x
The formulation (1) takes account of the fact that inclined flexural cracks may
occur.
The stress 0 ,obtained in (1), may not exceed a certain critical value. This value
px
can be obtained by using the limit envelope for anchorage failure. Such a limit
envelope consists of a number of parts (Fig. 3):
- the transmission length g,t ' over which the prestressing force is transmitted
into the memberj
- the development length g,d ' beyond which full anchorage capacity is obtained.
For detailed information on the phenomenon of bond and anchorage, reference is
made to [4].
-145-
}a~x = x,~
: I
- p
As a next step the shear capacity is controlled. Also here the areas cracked and
uncracked in flexure are considered separately.
In the region cracked in flexure the bending shear failure mode is governing: shear
failure occurs af ter a critical propagation of an inclined bending crack. The shear
flexure capacity is formulated in practically all building codes. The proposal for
the new Eurocode is
(2)
where
'Rd = 0.25 fctk/Y c
Ib Is I (3)
w
-146-
where
moment of inertia
bw web thickness
S static moment
cr average prestress in concrete (fully developed)
cp
fctd design tensile strength of the concrete
a reduction factor, taking into account that the prestressing
force in the direct vicinity of the support is not yet
fully developed and is defined as
a =
In order to control the production and simultaneously get areliabie value for the
tensile strength of the concrete, a standard shear test has been defined, details of
which are represented in Fig. 5. The test should be carried out regularly,
depending on the production and the requirements of the local control authorities.
The standard shear test is carried out on a fulI width slab with a length of at least
4 m.
LOADING BEAM B
0000
PASSIVE SUPPORT C
0000 _ . tHlb.am
The standard shear test gives directly reliable information on the shear tension
capacity, which is the most important failure mode for hollow core sI abs.
Furthermore a decrease of the product ion quality in time is discovered in an early
stage. Experience with this control procedure up to now is quite positive, since it
appeared that the shear tension capacity has a relatively low scatter. Fig. 7 shows
a number of results for slabs of different types and origin, size and prestress.
A complete paper with all details of the standard test will be published elsewhere
[5].
-148-
500 Vu (kN)
400
'l:;'
~~
e:.
.
300 qp
ot,
è~
200
~
°
100
I--
I h_c_ slab
---,
h.c. slab I
I i
"-============::j}
f- . re in forced
edge beam
I
confining
act ion by
edge beam
0, °2 °3 °2 0,
70 ,Ii~ear I~adjng, -
\
I
40
"'--" ---
~2
30 I'-
f--
- a:;--=
,,/ .- f--- ----~I
20
.-- °z
°l
" ,"/
f,..--
10
f'
oo A3 4 5 6 7 8 9 1011121314
span: m
Fig. 9. Load distribution factors for linear loadings in hollow core slab floors
without a topping [3]
-150-
Of course those calculations are only valid if the shear capacity of the joint is suf-
ficient.
This was recently investigated in a research program at the Delft University of
Technology [6]. Tests were carried out in a way as shown in Fig. 10.
hydraulic ja ck
lood cell
bali bearing
o
bali bearing
lood cell (2x)
Fig. 10. Test arrangement for a shear test on the longitudinal joint between
two hollow core slabs [6]
Hollow core slabs with a width of 1.20 m we re sawn in units with a length of 0.80
m. Those units were placed parallel to one another and the joint between them
was filled with a mortar with Dmax = 8 mm and a cube compressive strength of 20
N/mm 2• The slab units were at their ends connected with ties, consisting of free
steel bars I> 24 mm, to simulate the confining action of an edge beam. It can be
shown that a steel bar I> 24 mm, with a free length of 2400 m, as used in the test,
gives, per 'm length, the same confining force as an edge beam, reinforced with 21>
12 mm, around a slab assembly with a span of 14 m.
The shear loads were applied at both sides of the joints, in inverse direction, at
the second web from the joint (Fig. 10). Before applying those shear loads, the
joints were cracked, in order to simulate the unfavourable practical situation with
a shrinkage crack.
-151-
Altogether 9 tests were carried out, on slabs with cross-sectional depths of 260
mm and 400 mmo The failure mode observed, was in all the tests the same: failure
always occurred by shearing off of the slab sides (Fig. 11).
preformed crack
The joint mortar was, in spite of its relatively low strength, nev'e r governing for
failure, which can be explained from the multiaxial state of stress occurring
during shear loading.
The shear capacity of the joint region, as found in the tests, varied between 65
and 82 kN/m', which is considerably higher than will occur in practice.
The inclination of , the compression strut (Fig. 8), was calculated from the shear
load and the force in the tie bar. The inclination at failure varied between 42° and
75°.
A safe procedure for design is to dimension each slab, which carries a line - or a
point load, in such a way that it can carry the load on its own. The part which is
transmitted over the joint to the neighbouring slabs, is, additionally, taken into
account for the dimensioning of those slabs.
Af ter 1000 to 2000 cycles the coefficient of friction appeared to stabilize, Fig.
12b.
The tests showed that the coefficient of friction, up to a shear stress of 0.20
N/mm 2, is at least equal to 1.0.
coeff. of friction
2.0
!!! I
joi nt
1.5 ~O.t_ .l
w=O.30 r-n=4000-
/.. IW=O.3d~:3000
[;""
1.0. n=2000
0.5
The shear force, which can actually be transmitted across the joints in a slab
field, depends - for a given coefficient of friction - on the normal force on the
joint. This normal force depends on the bending moment, which is exerted on the
slab field (Fig. 13).
As a result of "arch-action" and "wedging-action" of the joints, the distribution of
the normal force on the joints can be schematically represented as shown in Fig.
13.b. This representation is based on strain measurements on the tensile ties in a
large scale experiment (Research Project "Demountable Building").
For IJ = 1 the shear capacity of the joint is equal to the normal force, transmitted
across the joint.
For an inner lever arm of 0,8 H, the maximum normal force on the joint is
-153-
with
and
wind lood qw
arch action
O.5N max
Since IJ =1 the expressions (4) and (5) also represent the shear capacity of the
joints considered.
Fig. 13.c shows the shear capacity for various values of the slenderness ratio À , in
comparison with the shear force distribution. The diagram shows, in which area
the shear capacity of the joint is sufficient and in which area additional shear
resistance is needed.
The additional shear capacity of the edge beam, which has not yet been taken into
account, mostly reaches out to guarantee a sufficient total shear resistance.
Otherwise shear coupiers across the joints are necessary.
LITERATURE
1. INTRODUCTION
The idea of combining prefabricated concrete with in-situ concrete offers many
possibilities for structural engineering, especially in the area of multistory
buildings and bridges.
In buildings, f.i., structural concrete toppings on prefabricated slabs can be func-
tional for a number of reasons [I}:
increase of the bearing capacity;
increase of the stiffness;
assemblation of deck elements to partially or fully continuing structures;
improvement of diafragm action;
improvement of water tightness and sound isolation;
hidden placement of pipes and ducts;
avoidance of differential deflections between neighbouring slabs.
In order to guarantee composite action between prefabricated sJabs and in-situ
toppings, it should be controlled whether the shear stresses in the contact area
exceed the shear capacity of the interface. In the past many tests have been
carried out in order to define the interface shear capacity. Those tests resulted in
various formuJations, of ten of empirical nature.
In this paper the interface shear capacity is discussed based on recent investiga-
tions.
YlAlEl + y 2A2E2
(1)
AlEl + A2E2
-156-
Y~~--
2
. -- . --
Yc Y
I
.
t i ·
. .
1
Fig. 1
(2)
Y
Yf
max E . dA (3)
lY
To control the shear stress at the interface, the integration is carried out from Ys
to Ymax (Fig. 2).
Fig. 2
-157-
o
Fig. 3
where cr is the stress perpendicular to the interface. The basic idea behind this
formula is that there is a general roughness and alocal roughness (Fig. 4). The
cohesion term c refers to the micro-roughness.
Fig. 4
-158-
T = C + (0 + pf
sy
) tan ~ (5b)
where p is the ratio of the reinforcement crossing the crack and fsy is its yielding
strength. Dowel action of the reinforcement can be neglected, because the devel-
opment of the full dowel capacity requires a large shear displacement.
Not only the roughness of the interface, but also the surface treatment plays an
important role with regard to the transfer of shear stresses across interfaces
between pree ast and in situ-concrete, because:
Laitance skin, dust, etc. are concentrated at the bottom zones of the surface,
whereas the tops tend to be less affected (Fig 5): the rougher the surface, the
less susceptible it is to the quality of the workmanship in cleaning and prepara-
tion. Furthermore, pollution can reduce the frictional properties of the sur-
face, so that all the terms involved in Eq. 5b are reduced.
If the surface of the precast member, before casting, is very dry, this member
will absorb water from the in-situ concrete, so that the quality adjacent to the
interface is governing for the capacity of the interface.
Fig. 5
III a surface which has been trowelled or tamped, so that the fines have been
brought to the top, but where some small ridges, indentations or undula-
tions have been leftj
IV a surface which has been achieved by slip-forming and vibro-beam
screeding;
V a surface obtained in a precast unit produced by some form of extrusion
techniquej
VI a surface which has been deliberately textured by brushing the concrete
when wet ideally to a specified depth of ridgej
VII as for (VI), but where the texturing is more pronounced, obtained typically
by brushing, by a transverse screeder, by combing with a steel rake or by
tamping with a former faced with a suitable expanded metal;
VIII a surface where the concrete has been thoroughly compacted but no
attempt has been made to smooth, tamp or texture the surface in any way
leaving a rough surface with coarse aggregate protruding (but firmly fixed
in the matrix)j
IX where the concrete has been sprayed when wet, to expose the coarse
aggregate without disturbing it;
X a surface which has been provided with mechanical shear keys.
The roughness increases from I to X. land II rarely occur in practice. III is com-
monly the most frequently observed type of "smooth" surface, found in practice.
All other surfaces exhibit a certain roughness. Extruded concrete mostly reaches
the roughness level as described under VI. VII can lead to very rough surfaces, but
it should be taken care that the concrete at the top is weil compacted. The figs.
6-9 [2] show some examples.
T
v
d =c + (pf d +
y
0) tan ~ < 0,25 f k
= c
(6)
where
p ratio of the reinforcement (perpendicuIarly) intersecting the inter-
face ~ 0 ,001
-161-
c tan 4>
Lvd (N/mm 2 )
L..O..---..---,--,-~.,
3.01--+---f--r--t----t
fsy .400N/mm 2
2
2.01---+ fee = 30N/mm
l,OI--/-----+.,,..c....-+---t------j
o
min ~
trowelling III 1
beam-screeding IV 1
nail-scra tching VII 2
naturally rough VIII 2
shear keys X 3
Apart from the surface treatment also the yield stress of the steel was varied.
Two steel types were used, one with a yield stress of fsy = 450 N/mm 2 and one
with fsy = 1200 N/mm 2• In this way the ratio yield stress/roughness was also
varied. This is the more interesting, because it might be possible that for the com-
bination of high strength steel and low roughness the yield stress is not reached.
Fig. 11 shows the results for design category 2 (intentionally roughened) in com-
parison with Eq. 6, with c = 0,4 fct and tan <jl = 0,9.
Four of the five combinations considered give results which lie at the safe side.
Only for the combination "naturally rough + high strength steel" a part of the
values give unsafe results.
On the basis of the diagram a number of conclusions can be drawn:
- The design equation is generally not valid for steel with a high yield stress.
- All results for the surf ace categories VII-X are, for fsy = 450 N/mm 2, at the
safe side. Since category VII is the lowest roughness of design category 2, the
validity of the equation for th is category is confirmed.
- The tests on interfaces, roughened by nail-scratching show that a confining
reinforcement is less effective than a compressive stress normal to the inter-
face (the points for pf are nearer to the design line than the points with 0 ).
sy
-163-
This is obvious, since the maximum confining stress generated by the rein-
forcement is only reached at a certain crack opening, whereas 0 is already fully
available in a closed interface: for w = 0 the contact area is therefore larger
than for w > o.
•o .. .. . fsy = '50
Noil screlched (u n I
.
..
..
..
.
..
.. VlII
.. 'lZII
0.30 - ..
t:;. Keyed (uni
.. .. fsy = 500 ..
..
..
..
..
.. ..~~
-~
t:;.
D~
/:)t:;.
8 0 0
l low~r
0
B:J-
t::i.
limil occording
0.20 10 FI P -
0.10 ~-
t~r
tJ
- I
Fig. 12 shows some values for the design category 1 (not roughened). Also here it
is evident that the high strength steel leads to unconservative results.
The tests for the other categories confirm the validity of Eq. 6, also for design
category 1, with respectively tan Ij> = 0,6 and c = 0,2 fct resp. 0,1 fct.
-164-
• ..
• Screeded U
fsy = 450
n =1J" .
..
.
..
n
.. III
.. .
6. Timber (un} .. I .. .
0.30 o Trowe led (un}
6.
. .
n f----- ..
0.20
• / 0
(8 0 0
0
6. 0 0 0
6.
0.10
~)
~
o
're
0
c2= Q2fct
... ~ C2=0.1fct (cat . I.n }
4. CONCLUSIONS
- The resistance of the interface between pree ast and in situ concrete can be
described by a friction equation, containing a cohesion term and a term which
refers to the clamping effect.
- The roughness of the contact surface can be divided into ten possibilities. For
the design and calculation of the interfaces loaded in shear two main design
categories can be distinguished.
- For the contact surfaces between prefabricated elements and in situ concrete,
the roughness as weIl as the surface treatment play an important role. The
classification into surface roughnesses does only make since if all other
requirements concerning curing, cleanliness and consistence of the young
concrete are dealt with.
-165-
5. LITERATURE
[1] FIP Commission 1I,"Horizontal composite structures composite slabs and
beams", Draft for new FIP Recommendation, 1990.
[2] FIP Commission 11, "Shear at the interface of precast and in situ concrete",
Recommendation 1982.
[3] Daschner, F., "Versuche zur notwendigen Schubbewehrung zwischen Beton-
fertigteilen und Ortbeton", Deutscher Ausschuss für Stahlbeton, Heft 372,
Berlin 1986.
-167-
1. INTRODUCTION
The simplest and most frequently used type of connection between two precast
concrete members loaded in compression is the unreinforced mortar joint.
Almost all precast concrete columns, wa lis and load bearing pre cast concrete facade
elements are placed on each other with an intermediate layer of mortar. Also
horizontal pre cast concrete members such as floor slabs, beams, spandrels are very
of ten placed in a mortar bed at the bearings. The bearing capacity of these mortar
joints forms here a very i mportant aspect.
In the precast concrete industry high strength concrete qualities can be realized,
which enable the fabrication of rather slender columns and walls with relatively low
percentages of reinforcement. In practice, ho wever, It appears very of ten that the
bearlng capacity of these precast high concrete quality columns and walls can not be
fully utiIized, due to the limiting influence of the intermediate mortar joint.
Nevertheless the mortar joint is quite popular, because it is cheep and easy to make.
To what extent a mortar joint limits the capacity of the members adjacent to it
depends on a number of influencing factors, such as the ratio between the strength of
the mortar in the joint and the concrete of the precast members at both si des of it,
and the dimensions of the joint.
Fallure of the joint can occur by crushing of the mortar or by splitting of the concrete
in the precast members adjacent to it. SpIitting is caused by an unequal stress
distribution in the joint af ter spalling off of the mortar at the edges of the joint.
The quality of the joint mortar, which directly influences the behaviour, depends on
the quality of execution at the building site. Bad workmanship may lead to air
retainments in the joint, reducing the effective load bearing joint area.
In the early seventies the DutchBuiIding Research Foundation (SBR) carried out an
investigation into the behaviour of unreinforced mortar joints, with the aim to
formulate design recommendations, which consider the influencing factors mentioned
previously. Those recommendations (1) have been adopted in the Dutch Code of
Structural Concrete, VB'74. Those recommendations were of ten decisive for the cross-
sectional dimensions of the precast members, and had in this way a considerable
impact on the overall costs (material, transport).
This on its turn resulted of ten in rather extreme actions, such as counting the number
-168-
of air bubbles in the joint surfaces, even using a microscope to include the smallest
ones, in order to find an "accurate" estimation of the effective joint area. In a
number of cases thick steel plates were applied to the end faces of the columns alming
at avoiding unequal stress-distributions in order to prevent splitting of the precast
concrete, so that the bearing capacity did not need to be reduced.
Now, more than fifteen years af ter the introduction of the recommendations, much
more experience has been gai ned.
Furthermore the quality of the mortar used has been signiflcantly improved.
Since the actual feeling was that some of the influencing factors have been estimated
too conservatively with regard to modern practice, the Dutch Study Commission on
Prefabrication (Stupré) took the initiative to study the new developments and insights,
in order to formulate an updated version of the recommendations. The results are
glven In this paper together with the experience from the past years.
The article is subdivided into two parts. The first part deals with the bearing capacity
of unreinforced mortar joints, loaded in compression by centrical and excentrical
forces. The second part analyses the efficiency of steel plates at both si des of the
mortar joint on the bearing capacity of this joint.
(1)
Where
Sp = the lowest prism-compressive strength of the concrete in the member,
adjacent to the joint
no = a reduction factor for the joint area, defining the effective part of this
area.
a the joint effectivity factor, defined as the ratio between the bearing
capacity of a column with a joint to a similar column without a joint
According to experiments (1), (2) the joint effectivity factor a can be formulated as
2
5(1-K) ... 1i
a = K .::...:..::.......:.:.!.-_=----=- (2)
2
5(1-10 ... K • li
where
-169-
cS = the ratio between the smallest width of the compressed joint area and the
joint thickness b/v or xlv (fig. 1)
K = the ratio between the compressive strength of the joint mortar and the
lowest compressive strength of the precast members adjacent to the joint.
K is formulated as
where
n a reduction factor, by which the difference in quality of mortars prepared
m
under site conditions and mortars prepared under laboratory conditions is
taken into account.
aM the compresslve strength of the joint mortar.
--t- -+- I
I
>1
+ >±
-t- -1--
x
Honzontol sechon
of the Jomt iI
I
J
11111111111114 : Aill ~
fig. 1. Stress distribution in a centrically and in an excentrically loaded joint.
2. 1 The reductlon factor no indicating the filling grade of the mortar joint
In lito (1), (2) the reduction factor no is defined as
no = 0,7 ror fluid colloidal mortars or dry packed mortars, poured or packed in
the joint space af ter the precast members have been erected.
no = 0,3 for the case of precast members placed in a mortar bed.
-171-
In the first case values higher than 0,7 are aUowcd, if these values are confirmed by
tests, carried out at the building site at regular intervals. However, no criteria have
been given tor those tests, which have resulted in the extremities mentioned before.
Stress concentrations which are able to genera te splitting cracks in the adjacent
pre cast concrete members can only occur due to large, irregularly spaced air
enclosures. Sm aU solitary air bubbles, which are regularly spread over the joint area,
can be disregarded.
The research commission therefore gives the folJowing recommendation:
"When establishing n by regular tests on the building site, solitary air enclosures
o
smaller than 0, 01 b, respectively 0,01 x, but not larger than 5 mm can be disregarded.
Air enclosures between 0,01 b, respectivily 0,01 x, and 0,1 b respectivily 0,1 x, have to
be taken into account. Air enclosures larger than 0,1 band 0,1 x are not allowed.
Furthermore, the last ten years of experience have proven that colloidal pouring
mortars have a more constant qualitly than other types of mortars, independently of
the way of application. Also the quality control of this type of mortars can be carried
out in an casier way.
Based on this experience, the commission recommends the following values for no to
be used if no regular tests on the building site are made.
no 0,9 for a colloidal pouring mortar
no = 0,7 for a dry packed mortar
no = 0,3 if the precast member is placed in a mortar bed.
,
-172-
0,
Bb
Sm -:.-p:cI.l..L.l~ ~Y
Bb
Ct
SI1-KI+ Ó 2
= K '::":"':"":":":-'--'''-;:--
S(1-KI+ Kó 2
o . JL..<~.LJhL--l---~-+--+----r---r-~r-1
O.l~:::""----I----+----t--+----1-_-l......_-t---t---j
-·!!...
Ó -y
2 4 6 8 10 12 14 16
nIO.21-:....--+-I-------,.,L-I-'"7"'c::::....h-,.....,:::=---:!:-.c:::::::::;::::+-=::::~1"':::::::::::_::t=::::::-_::::5===1
I
:1~-4--+---f-t-----t---r--i
:e ci.31--j~~~~q,,-L.:;.....f;....:;;;::±;;:;:::::::::;;;t::::::;;:=-"=:::==--_t_--T_j
~ ~ '~~I:f~jj
Vl.:o o.. ~
o ILI
-'"
QJ'"
I
......................
.......................
.......................
a. b.
fig. 3. Frictional forces caused by pressing out of the joint material (a) and non-
uniform stress distribution in the joint areas.
- due to the mechanism described previously, the mortar at the outer part of the joint
Iocally crushes, so that a redistribution of the stresses occurs, resulting in higher
stresses in the middle part of the joint (fig. 3 b). This unequal distribution of
stresses in the joint causes splitting stresses in the column. The stress trajectories
for this case are shown in fig. 4a.
- 174-
k--+-+-(ompression
Tension
a. b.
fig. 4. Stress trajectories in precast concrete columns due to a non-uniform stress
distribution in the joint (a) and joint between steel plates, aiming at reducing the
tensile and splitting tensile stresses in the precast column ends.
To prevent that the li mited bearing capacity of the joint requires a larger joint area
which, as a consequence, leads to unnecessary large column dimensions (material and
transport costs) of ten steel plates have been ap[)lied (fig. 4b). The basic philosophy
behind such a solution is clear: the steel plate has not only to take over the frictional
forces at the contact area but should also create a more uniform distributIon of the
normal stresses, so that the tensile- and splitting stresses are reduced to a harmiess
level. The steel plates, used for this purpose in pra\!tice, had thicknesses up to 50 mmo
The frequent use of such plates asked for a more detailled analysis of this solution.
Fig. 5. shows a cross section of the column, which was chosen for the analysis. The
strip, shade in fig. 5., had a width of 66 mmo
The reinforcement percentage was assumed to be 496. The reinforcement which was
equally distributed along the circumference of the column cross section, was
simplified to a !ine with an equivalent thickness.
Calculated strip
VI
N
Idealised reinforcement
446
This resulted in a steel stress of 74.7 N/mm 2 in the longitudinal reinforcement and a
stress of 12.1 N/mm 2 in the precast concrete. The fini te element mesh, chosen for the
calculation, is shown in fig. 6. All elements contain 4 integration points, in which the
stresses and the strains are calculated. The element size is 66 x 132 mmo For reasons
of symmetry only one half of the column is considered: in the axis of symmetry the
elements are assumed to be supported by roller bearings. The load at the upper edge
agrees with the stress distribution in the undisturbed area of the column.
-176-
3.2 Variables
Calculations have been carried out for various values of the steel plate thickness and
for various values of the effective joint area. The plate thichnesses we re chosen to be
d = 40,25, 10, 3, 1, and 0.1 mmo
The effective joint wldths were 396 mm (joint fully fIlled, concrete cover neglected)
and 264 mm (joint filled to 2/3 of the width, shaded area in fig. 6). By choosing the
last value a joint is simulated, in which the mortar has been partially pressed out from
the outer area, or has partially become ineffective by crushing. This can be considered
as an extreme case.
VI
·x
"'
êl
.31
0,
W
125 1
66 66 66 !
~
z z z z
I
-"" -"" -"" -""
co r-
a-. 0
a- a- -""I.
..,.
a- :Q '" '"
N
VI
N
VI ""
",'
N
lAl
...
E
E
Reinforcement -I-
~I
"
t:S§~§§~§~~~1 Steelplate
1 :1 P'd4P4Y4W4 Joint thickness
fig. 6. Finite element mesh (half column width, half joint width)
-177-
7a 396 40
7b 396 10
7c 264 40
7d 264 10
For a correct interpretation of the stress diagrams it should be realised, that the scale
of the stresses in the vertical and the horizontal direction is different (in the
horizontal direction a factor 10 targer than in the vertical direction).
'l'he drawings show that in the concrete at the column side of the steel plate, parallel
to this plate, always compressive stresses are acting. 'l'his can be explained by the
fact, that the stress trajectories concentrate on the middie part of the joint (fig. 4a.),
wh ere the stresses in the joint are larger. The smaller the effective width of the joint,
the larger are those compressive stresses. On the other hand, the stronger the
curvature of the compressive stress trajectories, the larger are also the splitting
tenslle stresses in the inner part of the column (compare fig. 7c with 7a and 7d wlth
7b).
Comparing the figures 7a with 7b, and 7c with 7d, where the thickness of the steel
plates is the only variabie, a surprising phenomenon is observed:
The splitting tensile stresses have nearly the same value, independent of the thickness
of the stel plates. 'l'his observation was confirmed by the calculations with other plate
thicknesses. This leads to the conclusion, that steel plates, applied in this way, do not
have the effect they have aiways been supposed to have.
Concentrating on the case with the smaller effective joint area, this behaviour can be
explained as follows:
a thin plate has hardIy any influence on the distribution of the stresses, which are
transmitted from the joint to the column. An unequal distrlbution of the stresses In
the joint will therefore lead to splitting tensile stresses in the column.
A thick plate smoothens out an unequal distribution of the stresses, so that the
stresses transmittted to the column are more uniformly distributed. As a secondary
-178-
Stress notation
Compression Tension
a. b.
Scale
r 10 N/mm 2
~ 1 N/mm 2
c. d.
I Joint width Joint width
Fig. 7. Vertical and horizontal stresses for different steelplate thickness and joint
width
-179-
effect, however, a thick plate restrains the lateral deformation of the concrete in
the joint area. Due to this mechanism splitting tensile stresses occur in the
column, which eliminate the positive effect of the plates, mentioned before.
Leteral
Compression Deformation
Compression
Tension
a. b.
fig. 8. Stress trajectories for (a) column with thin steel plate and (b) column with thick
steel plate.
-180-
4. CONCLUSION
In most cases the bearing capacity of an unrelnforced mortar joint is not governed by
the mortar crushing strength, but by splitting of the adjacent column region .
Using the design recommendations for unreinforced mortar joints, which we re used up
till now, the limitation of the mortar joint capacity lead of ten to uneconomically large
column dimensions.
The application of steel plates at the end races of the columns, or splitting tensile
reinforcement inside the column, does not result in a significant increase of the
bearing capacity of the joint region.
The fluid colloidal mortar types which are now available and the improved application
procedures aHow, however, higher values for the design values than have been used up
till now.
The most effective way to increase the bearing capacity of an unreinforced mortar
joint is therefore to use a stronger mortar for the joint and a stronger concrete for the
precast member adjacent to it.
REFERENCES
1. SBR bericht nr. 34, " mortelvoegen in de montagebouw", Samson uitgeverij Alphen
a.d. Rijn/ Brussel 1973
2. Dutch code of Practice VB 1974/1984
3. AHMAD, S.H. and SHAH S.P., "Stress-strain curves of concrete confined by spiral
reinforcement" ACI journal nov.-dec. 1982 pag. 484-490
-181-
SESSION 4
PRODUCTION
-183-
One can compare a lecture only concerning the product ion itself as a story
without a begin and an end. It is not that important to pay attention. to the pro-
duction itself, when all aspects determining:
how an element can be produced
the possibility to do so and to reach
the product quality level and cost level
are neglected.
This lecture is not only concerning structural wall elements but also facade ele-
ments.
The quality of design, details, workshop drawings, time available to prepare the
production, serie coefficient are more or less of main importance for the success
of the production, quality- and costwise spoken.
There are so many small details which seem to be not so important to our clients,
but of importance to us.
An expert advice, from the precaster, is not on the list, but it should, to be sure
that the other mentioned items can be fulfilled, also in the customer's interest.
In one way of another we have to transfer our specific know-how of product and
product ion to designers.
Planning, dealing with every stage of the project, is becoming more and more of
importance to meet the required time of delivery.
In practice all activities preceding the start of production preparations take more
time than planned on forehand and because we have to deliver in time there is
always a lot of pressure on the people involved.
To show you what is going on and what different production possibilities there are
available for walis, I take you with me on a side-trip along several segments of
building practice and different production facilities.
The next one is apart ment buildings, in which the same type of product ion
methods can be seen.
Then office buiJdings, with the application of load bearing and non-Ioad bearing
walls. Attention to design, details, connections and reinforcement.
An extrapolation can be found in high-rise buiJdings.
Elements then have to meet severe requirements.
1. INTRODUCTION
The definition of "high strength", with regard to the material concrete, shows an
ongoing shift to higher and higher values. For instance, the CEB-FIP Recommenda-
tions of 1970 did not yet give an upper limit. In 1978 the Model Code for concrete
structures took, as a highest quality, the class C50 into account (cube strength" 62,5
N/mm 2 ). The concept of the new CEB-FIP Model Code 1990, recently discussed in
Paris, indicates C80 as the highest quality-class, which implies that the cube
compressive strength is about 100 N/mm 2 • In this respect the new CEB-FIP Model
Code is in front of nearly all national building codes. Only the Norwegian Code NS
3473 defines a maximum characteristic cube strength of 105 N/mm 2. The American
Code ACI 318-83, which appeared in 1983, does not give any upper limit. Nearly all
material properties, necessary for design, like compressive- and tensile strength,
modulus of elasticity, creep- and shrinkage values, are simply given as functions of the
uniaxial cylinder strength. Records for the concrete strength are nowadays regularly
reported from the USA. In 1989 a concrete quality of 130 N/mm 2 was mentioned, used
for the construction of the Two-Union Square Building in Seattle [1].
It is not remarkable that special1y in Norway and in the USA great interest exists in
the application of concrete in the higher strength classes.
In the North-Sea, since Ekofisk 1 (1973), 21 offshore platforms were pIaced. The
specified 28-days characteristic cube strength increased in the meantime from 45 to
70 N/mm 2 (Gullfaks C, 1989, Fig. 1, Lit. 2.3). In the case of those platforms, the
quality requirements did not only concern the strength, but also the durability, the
workability and the speed of erection. To meet those requirements special attention
was given to the development of appropriate mixes. Experiments were carried out
with finer cement types, puzzolanes, finer sand fractions, super-plasticizers and other
chemical ad mixtures. A favourable circumstance in Norway is the availability of a
sufficient amount of silica-fume. Silica-fume is a by-product, which is obtained during
the fabrication of ferro-silicum and silicum metals. With 5-10% silica-fume, related to
the cement weight, concrete strength values between 70 and 100 N/mm 2 are obtained.
During the last years, however, not the streng th but the workability seems to obtain
-11)3-
the highest priority. Application of large amounts of silica-fume results of ten in bad
workability and a sticky concrete, which requires additional effort for cleaning the
formwork, etc. By the addition of low amounts of silica-fume (1 to 3% of the cement
weight) however, the pumpability and the stability of the mixtures increases and
excellent workability is obtained. This is an important advantage, because of ten the
densityof the reinforcement is very large (to 1000 kg/m 3 ).
In the USA, concrete with high strength is predominantly applied in high-rise buildings.
Fig. 3 shows the plan of the Two-Union Square Building in Seattle [1]. The co re of this
220 m high building consists of 4 round steel-concrete columns: these are steel ,pipes
with a diameter of 3 m, filled with high strength concrete. The most important reason
to use high strength concrete in buildings is the wish to build higher, but in the
meantime to keep the effective area as large as possible. Furthermore economie
advantages are mentioned. In [4] it is calculated that concrete with a strength of 96
N/mm 2 is 3,1 times as expensive as concrete with a strength of 21 N/mm 2 • Therefore,
however, one gets a strength which is 4,7 times higher.
-189-
steellacade columns
/ / /
concrete - ti lied steel pipe
steel beams
/ ~~ /// steel- concrete tloor
/ Ff'>., /
/ / ./ '~
~
/
f1
Iy,x,xlx,x
IX IXI.XIX
1
H-n
f.I::
I'
P I,
\ J 1/
ti II
With regard to the Two-Union Square Building, a comparison was made with an alter-
native solution, with a bearing steel structure. The volume of the steel, necessary for
the chosen composite structure, was about 5096 lower than for the pure steel variant.
Therefore the building costs were 3096 lower.
What the concrete strength means for the ratio "lost" surface area - total area, was
calculated by den Boer and Mans [5] for various reinforcement ratios of the column.
The results of these calculation are shown in Fig. 4.
C3S. poO% 0%
D 0 D
CSO.poO%
1"----+--:;;;-"9---+-c::?""""--t--C20.Po8%.
I 8%
0.. 0 0
Iy SOON/mrJ 0
"--~4-~-4---,:b--""--+-
C3S. po12%.
Iy: SOONlmm'
CSO. po2S%.
12% [IJ rn [IJ
25% 0 0 0
°so 75 100 125 150 175 200
building height {mI
400 kg/C
---
100 I
wit h out silic ofum e
0--0 with silica fume
80
I c:s =0.50
60
V ..0--:::::: t:::::=-: -- wlc .0.50
/
/ '~ w/c. O.60
40
~? ~
20
~ ------ ------- cem en t conten t: 270kg/m 3 -
c ement: PC45F
og greg ote: gr ovel/so nd S/C32
silica fu me: 10% o f cement weigh t
The mixes contain 270 kg/m 3 cement and 20 mI superplasticizer. Af ter 2 days the
compression strengths are nearly equal, but af ter 7 days the strength of the concrete
with cement + silica-fume is already 7 - 12 N/mm 2 higher than that of the reference-
concrete. Af ter 28 days the differences are 10 - 17 N/mm 2 .
Comparing tests, carried out by Gjorv [7], also at normal temperatures, showed
-193-
differences only af ter 5 - 7 days. Interesting is also the effect of the temperature.
Mijnsbergen [9] studied the strength-development at 10, 20 and 30°C. The basic mix
contained 375 kg/m 3 blast furnace cement A, with 0, 5 and 8% silica-fume respec-
tively. The results are shown in the diagrams of fig. 7 and 8.
It turns out that, at a hardening temperature of 30°C (the C-lines), an addition of 5
and 8% silica-fume gives a contribution to the 1-day's strength of 42 and 59%. If the
concrete hardens at 20°C (the B-lines) the influence of silica-fume after 1 day is much
smaller: values of 3 and 12% are found for 5 and 8% silica-fume addition. Af ter 2
days, those percentages incease to 14 and 30%, whereas after 28 days values of 7 and
21 % are reached. Hardening at 10°C shows another development. The curves intersect
over another. Addition of silica-fume only contributes to the strength af ter two days.
,....,/ /
7'___
3 -- -
1~/~
-
-/'/ 8%
The strength af ter 1 day is even smaller than that of the concrete without silica-fume.
Af ter 3 days addition of 5 and 8% silica-fume shows an increase of the strength of 14
and 63%. This shows that the addition of silica-fume can lead to an increase of the
strength of young concrete, if the hardening temperature is higher than 20°C.
In fig. 8 the results af ter 28 days are presented. This figure shows, that a higher
strength of the young concrete has to be paid with a lower strength at 28-days, if the
hardening temperature is used to increase the young concrete's strength.
-194-
4. MECHANICAL PROPERTJES
Jn the figures 9 - 12 a number of basic properties has been represented. Fig. 9 shows
the behaviour under uniaxial compressive 10ading. Concrete with higher strength
exhibits
a linear O-E behaviour to a relatively high percentage of the maximum stress;
a higher strain at maximum stress;
a steeper falling branche af ter reaching the maximum stress, which is synonymous
to more brittle behaviour.
For conventional concrete distinction is made between short-term and long-term
strength. For normal concrete the long-term strength is about 80% of the short-term
strength. According to [10], this is also valid for high strength concrete.
100~~---+--~F-++--+-~
fct (N/mm2)
8
/ [ 12)
6
~[ 11]
4
~/
." ~
\ Eg .1 [13)
2 ~
o
0.2 0.4
t:
0.6
(%.)
o 20 40 60 80 100
fee (N/mm2)
With re gard to the uniaxial tensile strength of concrete, most investigations state
that the ratio tensile-compressive strength, found in the past fornormal strength
concrete, is also valid for high-strength concrete. Fig. 10 shows the relations proposed
by Carrasquillo [11] and Pliskin [12] for high-strength concrete in comparison with the
relation
-195-
which was principally derived for normal strength concrete [13]. It is seen that the
differences are smal!.
Practically the same eonelusion holds true for the E-modulus. Fig. 11 compares
formulations aceording to four sources [11 , 13 , 14, 15].
(Jc Ifc
1. 6r--,---,----,----,---...
40
1131 ~~
.---
~
..-::-:-:-::
~
~
1.2t--i--r-+--+--+---l
30 ' ~,.;-
~ ~,;:-
~
20 --- O. Bt---tJf-t----"t---I---I----I
10
o
o 10 20 30 40 50 60 70
2
80 2 3 4 5
f cc (N/mm ) Cc Ic co
Fig. 12 shows the behaviour of confined concrete under compression. Results are
shown for high strength (95 N/mm 2) concrete cylinders with different confinement
ratio's [17]. It is shown that an inereasing amount of eonfining reinforcement will
in ere ase the ultimate stresses and eorresponding strains and reduee the slope of the
descending part of the stress strain curve.
Tests on shear resistanee of cracks in reinforced concrete are described in [18]. In [19]
it is shown that a lower bound expression for the shear strength is obtained by
~
fe
= 0.5 I p f
sy
ff
e
< 0,25 (2)
-196-
'l9- u/fc
0.4
0.3
-
0.2 ~
0.0 I
o 20 40 60 80 100
fc(N/mm 2 )
0.5 0 ...
0.01--~-5~0-~~~-~1~00
-~~~-~~15~0~-~....J
Fig. 14 Relation between abrasion resistance and concrete strength according to [20]
-197-
reinforcement as a reinforcement as a
function of concrete strength
funtion of buckling length
13000 M =61;OkNm N' d=3200 kN , Md='1or. 13000
12000~--.----.----r-~.----r--~ N' d=3200 kN . Md=320 kN m
11000 12000
10000
11000
10000 /
-- V
9000
8000~~~~~---r---r---+--~
9000
8000
~
//
7000
6000~~rt-~~~~~--+----r--~
7000
6000
825
/ /
5000
4000~~~~-+~~~~+----r--~
5000
4000
:/ / /'
3000
2000
100gL-~~~~~~=±~=t==~
10 30 50 70 90 110 130
3000
2000
1000
o
o
855
2
---
885/ 8115
..-/'
4
.-/ /
6
/
8
~
10 12
concrete streng th (N/mm2) buckling length (m)
material costs
800~--.----.--~r---.----r---.
200~--~~~---+--~----r-~
30 50 70 90 110 130
concrete strength
LITERATURE
[1] RALSTON, M., KORMAN, R., "Put that in your pipe and cure it", ENR, 16.2.89,
pp.44- 53
[2] RONNEBERG, H., SANDVIK, M., 'High strength concrete for North Sea
platforms", Concrete Internationsl, January 1990, pp. 29-34
[3] MOKSNES, J., 'North Sea oilgas production platforms - Reflections on 15 years'
experience, with special emphasis on post-tensioning and quality assurance
aspects", FIP-Notes, 1989/3, pp. 22-25
[4] MORENO, J., "225 W. Wacker Drive", Concrete International, January 1990, pp.
25-39
[5] DEN BOER, P., MANS, D.G., "Aspects of high-rise buildings in relation to the
structural design", Cement 1988, Nr. 4, pp. 13-20 (in Dutch)
[6] AITCIN, P., METHA, P., "Selecting materials and proportions for high strength
[19] HSU, T., "Discussion of [18], PCI-Journal, January-February 1988, pp. 166-168
[20] GJORV, 0., BAERLAND, T., RONNING, H.R., "Abrasion resistance of high
strength concrete pavements", Concrete International, January 1990, pp. 45-48
[21] BENNENK, H.W ., GALJAARD, J., NIJSSE, R., SOUWERBREN, C., SOEN,
H.H.M., "High strength concrete", Stuvo Report 90, July 1990
-201-
RECEIVED AFTERWARDS
-203-
DIAPHRAGM~CTION
ir. J. Stroband
Delft University of Technology
8ekker en Stroband, Consulting Engineers, Amsterdam
1 DESIGN CONSIDERATIONS
Floors in multi-storey buildings have to transfer horizontal loads to the
bracing elements, provide lateral support to the columns and ensure the
integrity of the overall structure. The actions include wind load and
support reactions from the columns caused by the lateral deflection of the
floor, with due attention to possible second-order effects.
The internal forces acting in the floor diaphragm depend on the location
of the bracing elements (Fig. 1). In a precast floor, the bending moments
can be resisted by means of a tie, formed by interconnection of the edge
beams or by placing a continuous reinforcement in the longitudinal joints.
The shear forces can be tranferred by means of a reinforced topping, but
it is economically advantageous to omit such a topping. The shear forces
then have to be transferred by means of structural connections between the
precast floor units. Under certain conditions it is also possible to take
advantage of the shear capacity of the grouted slab keyways, which is of
special interest for the widely used floors composed of hollow core slabs.
This paper discusses the behaviour of such floor diaphragms.
When a load is applied, cracks will occur in the joints. The tensile force
in the tension chord is a function of the lever arm z. In case of pure
bending, the lever arm is constant and the force in the tension chord is
proportional to the bending moment (Fig. 2b). If, however, the floor is
regarded as a tie and arch structure, the tensile force is constant. The
actual tensile force distribution will lie between these extreme values.
The design strength of the tension chord follows from:
2
1 q 1
N = - where z = internal lever arm
s 8 z
where z = internal lever arm = 0.8 h < 0.5 1
DDDDDDDDh F F F
DDDDDDDD
...
""""'I:lLUO I II
I
M
/
Ql,Q2&wind lood
F '" suppor t reodion
horizonlol aclion
/
[fililll!" Ir--
- - -=--"Ë] T
@ scheme of 0 precost f100r diophrogm
OITIEJ]O OOO[JE]OOO
00000000
BBBB88ffi
Ns = Ne
-- - - -- -~~:=:::::::: ~:--::-----------7- - - --
DOOD,DODO ~
r~--~-~~---------------------------------------
,,.. ......... , pure bending tie ond orch ......... .. .. ..
@ bending (bonded ten sion chord) © bending (u nbonded ten sion chord )
WJ "
>-------1
@ diophrogm sheor
stress distribution section A-A
cv overoge curvotu re k '" ~.
sr
k1
The effect of tension stiffening may be accounted for by assuming that the
curvature in the uncracked sections between the joints is neglible small.
With a crack spacing s ,the average curvature follows from the anchorage
r
length lb of the tie reinforcement (Fg. 2e):
k=k
l
!g
sr
where: k = curvature at the cracked section
1
With an unbonded tie, anchored at the ends of the floor, the tensile force
is constant and a single crack will occur. The deflection of the floor in
this case is proportional to the crack width wand thus proportional to
the total elongation of the tie (Fig. 2c).
Although the average horizontal shear stresses in the joints are low and
rarely exceed the value of 0,1 N/mm2, the analysis of diaphragm shear is
rather intricate. Because of shrinkage and alternating bending stresses,
it must be assumed that the transverse joints are cracked. Thus, the shear
forces will have to be resisted almost entirely by means of friction in
the compression zones of the joints (Fig. 2d). Assuming a coefficient of
friction f = 1, the maximum shear force that can be resisted is equal to
the tensile force in the tension chord. Because the force distribution in
the tension chord can not be evaluated by means of a simple calculation,
the lower bound value according to Fig. 2b should be selected to evaluate
the shear capacity of the joints. In simply supported floors, as in the
example considered above, the shear forces acting in the critical joints
near the supports exceed the shear capacity. Hence, the shear resistance
must be increased. This can be attained by means of internal ties crossing
the joints, in order to make use of the shear friction effect. For the
same reason ties have to be installed crossing the longitudinal joints.
1: =1: +f.O (1 )
u h c
where:
1: the interlocking strength
h
G normal stress
c
f coefficient of friction Fig 3 Schematic representation of a
cracked joint.
2,5
I'\.
~ I'--..
.2 2, 0
~ ~
u w.O,2
26)
~
f
~ -!!:.tO.,Jo
~ f:>...
c
.!!!
l,S
(13)-.
(21)-.
~, ~O w.O,20 r---- f.-<-j a)
7)
~
.~ w · 015
----
(321-- '''0 w.O.JO 12)
Q; "'·tl; 10)
3 w.0 2
-.......
-
1, 0 6)
(JIJ- 0,30
---
w~O,30 W ·
20)
w.O,20 ---{ 17)
I
L,s (31
w.o,JO
Regarding the shear strength of the joints the results also we re subjected
to a considerable dispersion. For joint mort ar with a compressive strength
of about 20 N/mm2, the shear strength in practically all cases , both under
monolitically increasing and repeated loading, exceeded 0.15 N/mm2. Thus,
the design value of the average shear stress in plain joints with a crack
width =< 0.3 mm should not exceed 0. 1 N/mm2. For higher shear stresses
indented joints can be used . The shear strength of indented joints , with
3 mm shear keys, under repeating load varied fr om 0.5 N/mm2 to 1.0 N/mm2,
Finally , it can be concluded from the test results is that there exist s no
unambiguous rel at ion between the shear stress and the longitudinal shear
displacement. At a stress level of 0.1 N/mm2 a maximum shear displacement
of· about 1 mm was found. This means that in practical cases the influence
of shear deformation on the deflection of the floor may be îeglected.
- 209 -
Most floors failed in shear at one of the jOints in grid lines 2 or 13,
with the exception of the prestressed floor (series A), that failed in
bending due to crushing of the hollow core slabs in the compression zone.
.,.
.
-2 1 0-
~ - -~HH~~++~++~4+~4++HH++hH++r~~
~ · _~4+~++H4~H+~++~++HR+rH++H~~
·b
@r -~~~~~~~~~~~kt~~~~~
I--a 13.720 • .'!:93~6~0_---.:.A=---_ _ _ __
1100 1 480 ~~
710 710
A
f . ~I f
B C
. R
A
f ft
B C 0
T•
H I.
~
I fI
r------+c-!~----11 :~!
~-11Q..-.1.t--- feit
I I'I
11
I.
11
I. II
I. I II II
.
•.
, - - testing rig
H. I. II I.
CV elevation of support eI) boundory conditions
adjusting nut
7
Aa· 19,6 mm 2, fa" 1570 N/mm 2 Aa· 50.3 mm 2, fa· 400 NJmm2 Aa " 22,6 mm 2, '0" 1150 N/ mm1
® series A ® seri es B CD seri es eend D
For all tests the forces in the longitudinal ties were calculated by means
of a linear analysis and compared to the test results [4]. In most cases
the calculated values were in reasonable agreement with the test results.
However, in the tests with pre-cracked joints at grid lines 2 or 13, the
test results deviated at higher load levels. Fig 6 shows a characteristic
example of the forces acting in the beam to beam connections at grid lines
o and 7 at various load level s. From fig. 6a it appears that up to a total
load Q = 18 kN, the measured forces reasonably agree with the calculated
values, but close to the ultimate load (Q = 30.3 kN) the floor tends to
u
behave like a tie and arch structure. From figure 6b it appears that also
a redistribution of forces occurs between the ties at grid lines B to D.
This behaviour can partly be explained by the fact that cracks occurred in
the longitudinal joints at grid lines Band D.
From the observed behaviour two design rules may be deduced. First, the
tension chord should be desingned on the basis of a tie and arch model and
second, internal ties should be installed crossing the longitudinal joints
1 2 J 4 5 6 7 e 9 10 11 12 13 ll.+-grid line
o
~- . A
\\\\ / !
I'--- '-....
JJJwlJNJ
t-..,
r0 r;:
Q:30kN ../
'.~:
' ..
,
" '- -.
" , i'" I"-
10~~~~+'-+-+-+-+-f-+-+-~
......
f\-\ Î"--
........ ........ 0 calculated
QlkL .... 12 , ",18 " z~ .~O
'-.....
Î'-
®-o 1 2 3 I. 5 6 7 8 9 10 11 12
-....-tensile force (kNI
The tests with the pre-cracked joints clearly demonstrated the favourable
influence of the shear friction effect on the shear capacity of the joint.
To set an example, Fig. 7 shows the measured and calculated tensile forces
- 212 -
Ë16m r;
z
=.
ë
Ë e
E
t
12
L / -'
.-
.,*
, '
. . . .rl:~
V-
/
---
__ - f,f
- - calculoted
- - - --measured
o
(
\
1 2 J l, ij
I
/
1 2 J o
,V
1 2 J
~ force IkNI ~ force I kNI ~ force IkNI
The test results prove that internal longitudinal ties have a favourable
effect on the shear behaviour of floor diaphragms . Further recommendations
for the design of floor diaphragms are presented in [4].
60,---,,----,---,,---,----,----,----,
- - colculoted
Ê /
501-~!--t+-+- .......... _. c 206
~ V
1 LO~-4-~~~- ~...-_...-....4...-...-....-.-.+..-...--.-
............ ...~...-_.~
..
rn30r-~-r~~-+---+---r--~~
1 V,/"· 20
t 10 l/
V' L
_ deflection Imml
se ries C
~
NO
, ql
No : ql
5 QUAlITY ASSURANCE
The proper functioning of a floor is determined to a great extent by the
probability of human errors in the design process and during construction
and use. The only effective way to prevent such errors is to introduce a
system of quality assurance comprising the entire building process, from
-214-
the design to the final handing over of the completed project. Guidelines
concerning quality management and quality systems are given in CEN codes
EN 29000 - EN 29004. As to floors special attention should be paid to the
design of the tying system and the detailing of the connections. At the
construction site good care should be given to executing the tying system,
the joint cast and the curing of the joints, according to the guidelines
given in [8].
6 REFERENCES
1. WESTERBERG, B, Design of floors with regard to horizontal loads,
Strangbetong, 1990.
2. CUR-report 135, Demontabel Bouwen (Demountable Construction), CUR,
Gouda 1990. (In Dutch)
3. STROBAND, J, and J.J. KOLPA, The behaviour of a demountable floor,
Proceedings Symposium on Demountable Concrete structures, Delft
University Press, Delft 1985, pp 201-216.
4. CUR-report 136, Voegen in geprefabriceerde constructies (Joints in
precast floors), CUR, Gouda 1988. (In Dutch)
5. CEB-FIP Model Code 1990, First Draft, CEB, Lausanne 1990.
6. FIP Guide to good practice, Quality assurance of hollow core slab
floors, Fourth Draft, FIP, London 1989.