2 15P09 PDF

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Part 9
Commentary1
— 2017 —
FOREWORD
The purpose of this part is to furnish the technical explanation of various articles in Part 1, Design, Part 3, Fabrication, Part 4,
Erection, Part 5, Bearing Design and Construction, Part 6, Movable Bridges, Part 7, Existing Bridges, and Part 8,
Miscellaneous and to furnish supplemental recommendations for use in special conditions. In the numbering of articles of this
part, the second and succeeding digits in each article number represent the article being explained.
TABLE OF CONTENTS
1
Section/Article Description Page
Part 1 Design
9.1.1 Proposals and Drawings . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-6 3

9.1.1.8 Design of Public Works Projects (1993) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-6
9.1.2 General Requirements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-7

9.1.2.1 Materials (2016) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-7
9.1.2.2 Welding (2016) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-10
9.1.2.5 Deflection (2013) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-11
9.1.2.6 Clearances (1995) R(2014). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-11
9.1.2.14 Special Trackwork on Bridges (2015) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-11
9.1.3 Loads, Forces and Stresses . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-12

9.1.3.3 Live Load (1995) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-12
9.1.3.4 Distribution of Live Load (1993) R(2014). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-12
9.1.3.5 Impact Load (2007) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-13
9.1.3.6 Centrifugal Force (2002) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-14
9.1.3.7 Wind Forces on Loaded Bridge (2009) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-15
9.1.3.8 Wind Forces on Unloaded Bridge (2005) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-15
9.1.3.10 Stability Check (2005) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-15
9.1.3.11 Bracing Between Compression Members (2017). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-16
9.1.3.12 Longitudinal Forces (2005) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-16
1
References, Vol. 71, 1970, p. 379; Vol. 72, 1971, p. 153; Vol. 73, 1972, p. 176; Vol. 75, 1974, p. 336; Vol. 76, 1975, p. 240; Vol. 77, 1976, p. 249; Vol. 78,
1977, p. 75; Vol. 80, 1979, p. 188; Vol. 81, 1980, p. 132; Vol. 82, 1981, pp. 78-87, incl; Vol. 84, 1983, p. 100; Vol. 90, 1989, p. 98; Vol. 91, 1990, p. 121;
Vol. 92, 1991, p. 80; Vol. 93, 1992, p. 124; Vol. 94, 1993, p. 1; Vol, 94, 1994, p. 145; Vol. 96, p. 74; Vol. 97, p. 177. Reapproved with revisions 1996.
© 2017, American Railway Engineering and Maintenance-of-Way Association 15-9-1

Steel Structures
TABLE OF CONTENTS (CONT)
9.1.3.13 Fatigue (2016). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-18

9.1.3.14 Combined Stresses (2005) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-25
9.1.3.15 Secondary Stresses (1994) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-25
9.1.3.16 Proportioning of Truss Web Members (2004) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-25
9.1.4 Basic Allowable Stresses . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-26

9.1.4.1 Structural Steel, Fasteners and Pins (2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-26
9.1.4.2 Weld Metal (1994) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-29
9.1.4.3 Cast Steel (1994) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-30
9.1.5 General Rules. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-30

9.1.5.4 Thickness of Material (2011) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-30
9.1.5.8 Net Section (2005) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-31
9.1.5.9 Connections and Splices (2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-31
9.1.5.10 Field Connections (1994) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-32
9.1.5.11 Development of Fillers (2015) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-32
9.1.5.12 Combinations of Dissimilar Types of Connections (1993) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-32
9.1.5.13 Sealing (1993) R(2014). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-32
9.1.5.14 Connections of Components of Built-up Members (1993) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-32
9.1.6 Members Stressed Primarily in Axial Tension or Compression . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-32

9.1.6.1 Compression Members (2004) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-32
9.1.6.2 Outstanding Elements in Compression (2004) R(2014). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-33
9.1.6.4 Lacing and Perforated Cover Plates for Tension and Compression Members (2009) R(2014) . . . . . . 15-9-33
9.1.6.5 Effective Net Area for Tension Members - Strength (2007) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-34
9.1.6.6 Effective Area for Tension Members - Fatigue (2007) R(2014). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-35
9.1.7 Members Stressed Primarily in Bending. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-35

9.1.7.1 Proportioning Girders and Beams (2004) R(2014). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-35
9.1.7.2 Flange Sections (1994) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-35
9.1.7.3 Thickness of Web Plates (2004) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-36
9.1.7.4 Flange-to-Web Connection of Plate Girders (2009) R(2014). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-36
9.1.7.8 Web Plate Stiffeners (Intermediate Transverse and Longitudinal) (2012) R(2014) . . . . . . . . . . . . . . . 15-9-37
9.1.7.9 Composite Steel and Concrete Spans (2008) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-38
9.1.8 Floor Members and Floorbeam Hangers . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-39

9.1.8.3 End Connections of Floor Members (1993) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-39
9.1.9 Construction with Fasteners. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-39

9.1.9.3 Edge Distance of Fasteners (2005) R(2011) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-39
9.1.9.5 Fasteners in Indirect Splices (2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-40
9.1.10 Welded Construction. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-40

9.1.10.1 Transition of Thickness or Widths in Welded Butt Joints (2012) R(2014). . . . . . . . . . . . . . . . . . . . . 15-9-40
9.1.10.2 Prohibited Types of Joints and Welds (2016) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-40
9.1.10.4 Welded Attachments (2004) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-42
9.1.11 Bracing . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-42

9.1.11.2 Lateral Bracing (2012) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-42
9.1.11.4 Cross Frames and Diaphragms for Deck Spans (1994) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-43
© 2017, American Railway Engineering and Maintenance-of-Way Association
15-9-2 AREMA Manual for Railway Engineering

Commentary
9.1.12 Pins and Pin-Connected Members . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-43

9.1.12.1 Pins (2015) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-43
9.1.13 Continuous and Cantilever Steel Structures . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-43

9.1.13.6 Uplift (2015). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-43
9.1.13.8 Longitudinal Stiffeners (2010) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-44
9.1.14 Fracture Critical Members . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-44

9.1.14.1 Scope (2015). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-44
9.1.14.2 Definitions (2013) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-44
9.1.14.3 Design and Review Responsibilities (1997) R(2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-45
9.1.14.5 Notch Toughness of Steel in Fracture Critical Members (2006) R(2008) . . . . . . . . . . . . . . . . . . . . . . 15-9-45
Part 3 Fabrication
9.3.1 General . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-47

9.3.1.1 Fabricator Certification (2016). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-47
9.3.1.6 Thermal Cutting, Copes, and Access Holes (2009) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-47
9.3.1.8 Planing Sheared Edges (1994) R(2008) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-47
9.3.1.18 Bent Plates (2000) R(2008) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-47 1
9.3.1.19 Piece Marking (2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-48
9.3.2 Bolted Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-49

9.3.2.1 High-Strength Bolts, Nuts, and Washers (2005) R(2008) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-49
9.3.2.2 Installation of High-Strength Bolts (2005) R(2008) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-49
9.3.2.4 Size and Workmanship of Holes (2017). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-50
9.3.2.5 Preparation of Holes for Shop-Installed Fasteners (2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-50 3
9.3.2.6 Preparation of Holes for Field-Installed Fasteners (2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-50
9.3.2.13 T esting and Documentation of High-Strength Fasteners (2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-50
9.3.3 Welded Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-50

9.3.3.1 General (2015) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-50
9.3.3.3 Flange-to-Web Welds of Flexural Members (1995) R(2008). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-51
9.3.3.4 Tack Welds (1995) R(2008) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-51 4
9.3.5 Inspection . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-51
9.3.5.5 Inspection – Welded Work (2002) R(2008) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-51
9.3.6 Shipment and Pay Weight . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-51

9.3.6.3 Pay Weight (2009) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-51
Part 4 Erection
9.4.6 Plans (2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-51
Part 5 Bearing Design and Construction
9.5.1 Bearing Design. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-51
AREMA Manual for Railway Engineering 15-9-3

Steel Structures
9.5.1.2 General Requirements (2010) R(2017). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-51

9.5.1.5 Bearing Selection Criteria (2010) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-52
9.5.2 Basic Allowable Stresses . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-55

9.5.2.1 Structural Steel, Bolts and Pins (2012) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-55
9.5.2.6 Polyether Urethane Disc Bearings (2007) R(2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-55
9.5.3 Steel Bearing Component Design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-56

9.5.3.2 Materials (2016) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-56
9.5.3.3 Shoes and Pedestals (1997) R(2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-56
9.5.4 Bronze or Copper-Alloy Sliding Expansion Bearing Design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-56

9.5.4.3 Design (2001) R(2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-56
9.5.5 PTFE Sliding Bearing Surface Design. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-56

9.5.5.3 Design (2006) R(2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-56
9.5.7 Multi-Rotational Bearing Design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57

9.5.7.1 Scope (2007) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57
9.5.7.3 Design (2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57
9.5.10 Bronze or Copper-Alloy Sliding Expansion Bearing Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57

9.5.10.1 General (2001) R(2008) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57
9.5.11 PTFE Sliding Bearing Surface Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57

9.5.11.1 General (2002) R(2008) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57
Part 6 Movable Bridges
Foreword. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57
9.6.1 Proposals and General Requirements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57

9.6.1.1 General (1986) R(2010) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57
9.6.1.8 Signals and Interlocking (2012) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-58
9.6.2 General Features of Design. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-59

9.6.2.12 Rail Ends (2015) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-59
9.6.3 Loads, Forces and Stresses . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-59

9.6.3.11 Machinery Supports (2013) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-59
9.6.4 Basic Allowable Stresses and Hydraulic Pressures. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-59

9.6.4.2 Machinery Parts (1993). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-59
9.6.4.8 Hydraulic Systems and Components (1984) R(2010) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-59
9.6.5 General Details. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-59

9.6.5.13 Lubrication (2008) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-59
9.6.5.34 Special Provisions for Swing Bridges (2003). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-60

Commentary
9.6.5.36 Special Provisions for Vertical Lift Bridges (1997) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-60

9.6.5.37 Hydraulic Systems. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-60
9.6.6 Wire Ropes and Sockets . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-60

9.6.6.2 Diameter of Rope (2010). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-60
9.6.6.3 Construction (2010). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-61
9.6.6.7 Wire – Physical Properties (2010). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-62
9.6.6.8 Ultimate Strength (2010). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-62
9.6.9 Erection. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-62

9.6.9.1 Erection of Machinery (1996) R(2002). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-62
9.6.9.4 Lubrication (2008) R(2010) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-62
9.6.9.7 Counterweights (1983) R(2003) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-62
Part 7 Existing Bridges
9.7.2 Inspection . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-63

9.7.2.1 General (2010). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-63
9.7.3 Rating . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-63 1

9.7.3.1 General (2015). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-63
9.7.3.2 Loads and Forces (2016) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-64
9.7.3.3 Stresses (2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-66
9.7.4 Repair, Strengthening and Retrofitting . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-71

9.7.4.2 Plate Girders or Rolled Beams (2012) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-71
3
9.7.5 Maintenance . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-71
9.7.5.2 Maintenance of Structural Elements (2015) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-71
Part 8 Miscellaneous
9.8.1 Turntables . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-71

9.8.1.3 Basic Allowable Stresses and Deflections (2008) R(2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-71 4
9.8.2 Method of Shortening Eyebars to Equalize the Stress . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-71
9.8.2.1 General (2017). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-71
9.8.2.2 Shortening Eyebars with Clamp Plates, Rods and Trammels (2017) . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-72
9.8.2.3 Shortening Eyebars with Ring Heating (2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-72
9.8.2.4 Determining Approximate Dead Load Stress (2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-72
9.8.3 Anchorage of Decks and Rails on Steel Bridges . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-73

9.8.3.1 Foreword (2010) R(2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-73
9.8.3.2 Anchorage of Decks to Bridge Spans (2012) R(2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-74
9.8.3.3 Anchorage of Rail (2011) R(2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-75
9.8.3.4 Rail Expansion Joints (2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-76
9.8.4 Unloading Pits . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-77

9.8.4.1 Scope and Purpose (2015). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-77

Steel Structures
9.8.4.6 Structural Supporting Beams (2015) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-77

9.8.4.7 Concrete Pit (2015) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-77
9.8.5 Walkways, Handrails, and Fixed Ladders on Bridges . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-77

9.8.5.2 Clearances and Minimum Dimensions (2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-77
9.8.5.3 Loads (2014) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-78
9.8.7 Guide to the Preparation of a Specification for the Cleaning and Coating of Existing Steel Railway Bridges
...................................................................................... 15-9-78
9.8.7.4 Coating Systems (2009) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-78
Welding Index (2004) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-85
LIST OF FIGURES
Figure Description Page
15-9-1 Isolines for First-Percentile Minimum Temperatures (USA). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-9

15-9-2 January Design Temperature 1 Per Cent Basis (Canada). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-10
15-9-3 Stress Range vs. Number of Cycles for Various Detail Categories . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-23
15-9-4 Bolted and Riveted Gusset Plates . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-31
15-9-5 Examples of details at intersection of longitudinal attachments and vertical attachments welded to web. . 15-9-41
15-9-6 Examples of welded details at certain locations at the intersection of vertical attachments and gusset plates welded to
web. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-42
15-9-7 Schematic Showing Relation Between Static and Dynamic Fracture Toughness . . . . . . . . . . . . . . . . . . . . 15-9-46
15-9-8 Marking Location on Truss Connection. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-49
15-9-9 Marking Location on Girder Splice . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-49
15-9-10 General Dimensions for Rating of Specific Equipment. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-64
15-9-11 Riveted Bridge Components . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-68
LIST OF TABLES
Table Description Page
15-9-1 Parameters Used to Develop Table 15-1-7 and Table 15-1-10 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-20
15-9-2 Constant A and Thresholds for Detail Categories . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-23
15-9-3 Weight and Axle Spacing of AAR Standard Freight Cars . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-65
PART 1 DESIGN
9.1.1 PROPOSALS AND DRAWINGS
9.1.1.8 DESIGN OF PUBLIC WORKS PROJECTS (1993) R(2014)
a. The purpose of requiring consulting engineers to be familiar with the design of railroad bridges is to ensure compliance
with the Company’s standards and operating procedures with minimum time involvement of the Company’s
engineering staff.

Commentary
9.1.2 GENERAL REQUIREMENTS
The safety and reliability of a bridge is governed by material properties, design, fabrication, inspection, erection and usage.
The following are contributing factors in bridge failures: inadequate inspection and non-destructive testing; design details
resulting in notches or high stresses due to secondary effects; joints which are difficult to weld and inspect; hydrogen-induced
cracks; improper fabrication, welding and weld repair; lack of base metal and weld metal toughness. Excessive attention to a
single preceding item will not overcome the effects of a deficiency in any other item.
The fatigue provisions of the AREMA recommended practices are based on a design loading which minimizes the possibility
of fatigue crack growth under regular traffic (see Article 9.1.3.13).
9.1.2.1 MATERIALS (2016)
a. Twist-Off Tension-Control Bolts: ASTM includes two designations for “twist-off” type tension-control bolts (F1852
and F2280, which have nearly identical properties to the regular F3125 Grade A325 and A490 bolts, respectively).
Only F1852 bolts are included in the Manual, for reasons described below.
F1852 and F2280 bolts are tightened until they achieve a torque that twists off the spline on the end of the bolt.
Although this method of achieving a prescribed torque is quite reliable, the tension in twist-off-type bolts is generally
not as consistent as that which can be achieved with ASTM F3125 Grade A325 and A490 bolts when tightened by the
turn-of-nut method.
Twist-off bolts control the failure torque of the spline through the metallurgy and geometry of the bolt, according to the 1
properties selected by their manufacturers. The achieved bolt tension is controlled by the accuracy of this torque plus
the consistency of the friction developed in the contact between the threads of the nut and bolt. If the threads are
damaged or contaminated, or the viscosity of the lubrication has changed due to evaporation or temperature, the
resulting bolt tension can vary significantly in spite of having a consistent torque. As friction increases, the achieved
bolt tension decreases.
When properly performed, turn-of-nut tightening achieves a prescribed elongation of the bolts to reliably stretch them 3
into plastic yield. Variations in the friction may change the torque required to stretch the bolt, but will not affect the
final stretch and achieved bolt tension because the stretch is controlled by the rotation of the nut.
F1852 bolts are approved by this Manual and require careful control of the installation procedures to achieve consistent
and reliable bolt tensions. Installation procedures should include using fresh bolts and prevention of damage to or
contamination of the bolt threads. It is not feasible to reliably clean and re-lubricate bolts once they are contaminated or
damaged, or if the lubricant has dried, and so re-lubrication by the end user is not permitted. 4
F3125 Grade A490 bolts have a much shorter yield plateau, and the yield stress is much higher in proportion to the
ultimate stress than is the case for Grade A325 bolts. F2280 bolts have similar metallurgical properties to Grade A490
bolts and are therefore less tolerant of minor variations in the thread-to-thread friction.
As a result of the increased ductility and greater ratio of ultimate to yield stress, F1852 bolts tolerate greater variations
in friction (controlling the tightening torque) than F2280 bolts. Therefore, smaller variations in nut friction in F2280
bolts can lead to more frequent failures than is the case for F1852 bolts, which have a longer yield plateau. For this
reason, only F1852 bolts are included in this Manual.
Weathering Steel: Refer to Table 15-1-1, Note 2: There is a potential for atmospheric corrosion rates to increase in
applications that subject weathering grade steels to frequent alternating wet and dry or continuously moist conditions
for prolonged periods of time; or to corrosive chemicals, including deicing salts. Guidelines for proper application of
unpainted weathering steels in bridges may be found in FHWA Technical Advisory T5140.22 “Uncoated Weathering
Steel in Structures”, dated October 3, 1989.

Steel Structures
Impact Test Requirements for Structural Steel: Table 15-1-2 and Table 15-1-14 make provisions for materials with
improved notch toughness. ASTM A709, Grade HPS 70W and Grade HPS 50W steels have such high toughness that
when they were included in the ASTM A709 Specification, the Zone 3 requirements, which are the most severe, were
specified for all zones for both Non Fracture Critical, Table 15-1-2, and for Fracture Critical, Table 15-1-14. Because
of their high toughness it was decided to eliminate the need to choose the appropriate zone when using HPS 50W or
HPS 70W and treat all zones alike.
Refer to Table 15-1-2 and Table 15-1-14: “Service Temperature” shall be taken to be the lowest ambient temperature
expected for the area in which a structure is to be located or to which a structure is to be exposed while in service. The
testing zones correspond with those chosen by AASHTO and imply the service temperatures listed in the tables. Zone
1 implies a minimum service temperature of 0 degrees F; Zone 2 implies a minimum service temperature of –30
degrees F; Zone 3 implies a minimum service temperature of –60 degrees F.
For guidance in determining the Lowest Anticipated Service Temperature for a particular location in the United States
or Canada, Figure 15-9-1 and Figure 15-9-2 may be used. Both figures show temperatures in degrees Fahrenheit.
Figure 15-9-1 (U.S. and Alaska) shows isolines for which there is a 99% chance that the daily minimum temperature
will be no lower than shown. Figure 15-9-2 (Canada) shows isolines for which the temperature during January will be
no lower than shown for 99% of the time.

Commentary
Figure 15-9-1. Isolines for First-Percentile Minimum Temperatures (USA)

4
Refer to Table 15-1-2: The recommended practice is silent on energy requirements for material thicker than 4 inches
even though some of the materials listed are available in greater thicknesses. Table 15-1-2 will normally apply to
welded main load carrying components subject to tensile stress and such applications will be rare for thicknesses
exceeding 4 inches. Nevertheless, if an engineer wishes to use a greater thickness, the notch toughness requirement for
these materials not listed in Table 15-1-2 should be specified.

Steel Structures
Figure 15-9-2. January Design Temperature 1 Per Cent Basis (Canada)
b. Based on commonly accepted approximate values for E and  obtained from test results, the approximate value for G
is derived using the following theoretical Equation:
G = E  21 +  .
f. Fracture Critical Members require additional consideration. This includes increased material toughness as specified in
Section 1.14, Fracture Critical Members.
9.1.2.2 WELDING (2016)
a. Prior to 1990, these recommended practices stipulated that welding of structural steel should conform to the American
Welding Society Structural Welding Code Steel ANSI/AWS D1.1. With the introduction of the Bridge Welding Code
ANSI/AASHTO/AWS D1.5 in 1988, under a joint development effort of the American Association of State Highway
and Transportation Officials (AASHTO) and the American Welding Society (AWS), the AREMA had available a
welding specification that specifically addresses bridge construction. The development of the AWS D1.5 code
represents a landmark of cooperative industry action to address the proliferation of costly and sometimes contradictory
regulations. While some specifications of AWS D1.5 may appear unorthodox to members of the welding community,
AWS has agreed that AASHTO should play a deciding role in determining the specifics of the code.

Commentary
b. Since AWS D1.5 is directed toward the construction of highway bridges constructed in accordance with standard State
Specifications, several terminology and definition substitutions must be made in order to render the bridge code
applicable to the construction of railroad bridges.
9.1.2.5 DEFLECTION (2013) R(2014)
b. Prior to 1969, the deflection limitation was covered by an article headed “Depth Ratios.” Structures built with depth
ratios meeting the requirements of that article were satisfactorily stiff for railroad operations since the stresses allowed
for A36 (or A7) steel were used in the design. An explanation in 1959 stated that depth ratios resulted in deflections of
1/800 for girders and trusses and 1/600 for rolled beams used as girders under E72 loading (Reference 21). A
statement to use 1/640 if shallower depth spans were required appeared in 1962 (Reference 23). Since the 1969 edition
of these recommended practices introduced and permitted the use of a variety of higher strength steels, it became
necessary to define more accurately the degree of stiffness which is desirable in terms of the deflection of the structure
rather than in terms of the depth ratio (Reference 24). Relating deflection to live loading also gives a more appropriate
basis for ballasted deck bridges, for which the live load is generally a lesser percentage of total load than for open deck
bridges.
c. Waddell (1916) recommended a vibration load of 700 lb. per linear foot for loaded chords to ensure sufficient lateral
rigidity in members (Reference 165). Furthermore, observations and tests at CN determined that lateral forces on
bridges from equipment can exceed the forces given in Article 1.3.9. Nevertheless, experience indicates that the
provisions for lateral forces of this Chapter have generally resulted in satisfactory structures with sufficient lateral
rigidity when all the recommended loads and lateral forces are considered.
The lateral deflection limits recommended are 50% of the FRA allowable limits for alignment deviation for Class 5 1
Track (Reference 53). Therefore, at least 50% of the allowable limit remains for variations in track alignment, rail
wear and track fastener wear or movement. For higher Classes of track the allowable limit needs to be reduced
accordingly. The limit is applied over a 62 foot chord for tangent track and curved track and also on a 31 foot chord for
curved track.
9.1.2.6 CLEARANCES (1995) R(2014)

3
The requirements for clearances were changed in the 1983 edition to be slightly more severe than in previous editions
(previously changed in 1969). This was done to be consistent with the recommendations of AREMA Committee 28 to
accommodate the increased dimensions of cars and of higher and wider loads.
9.1.2.14 SPECIAL TRACKWORK ON BRIDGES (2015)

4
The possibility of derailment is higher at locations where there is special trackwork.
Any gap in the rail or rail surface discontinuity also potentially increases the impact generated. Movable bridge joints are
covered in Part 6. Expansion joints are covered in Part 8.
With regard to switches, the worst cases are those where an entire turnout is installed on a bridge that has a frog with a gap
(Rail Bound Manganese (RBM) Solid Self Guarded Manganese (SSGM) or rigid frog). Similar effects occur when only the
frog of a switch is located on a bridge. (Reference 133)
Where the switch (point of switch to heel of switch point) is installed on a bridge and the frog is off the bridge, the vertical
impact is less but the switch induces lateral impact to bridges. This may become a deck fastener issue on open decks and a
ballast issue on ballast deck bridges.
While installing turnouts on bridges is not advisable, it is sometimes unavoidable. There are ways to reduce and possibly
mitigate the effect (impact) by use of spring or moveable point frogs.

Steel Structures
Placing a crossing diamond on a bridge is not recommended.
9.1.3 LOADS, FORCES AND STRESSES
9.1.3.3 LIVE LOAD (1995) R(2014)
a. The recommended live load of Cooper E 80 for the design of steel structures was adopted in 1967 by Committee 15.
While locomotives with weights greater than Cooper E 72, the previously recommended design live load, are not likely
to be found on any railroad in the United States, there is a trend toward heavier locomotives, and some of the heavy
cars produce loads equivalent to Cooper E 80 or greater.
Heavy double stack cars with axle loads of 78,750 lb per axle and 4–axle 315,000 lb gross weight cars (both using the
so-called 120 or 125 ton truck) accepted in regular service on certain railroads produce the equivalent of nearly E 80
Loading on shorter spans. In 1995, an alternate loading was introduced with a spacing similar to coupled typical 4–axle
cars with an axle load 25% higher than the Cooper E 80 load to address problems associated with fatigue on shorter
span lengths.
d. For members receiving load from more than one track, the proportions of full live load on the tracks to be used for
design are determined by use of the theory of probability to determine the frequency with which stresses of various
magnitudes might occur. Consideration was given to the fact that most of the trains which pass over a bridge will
produce lower stresses than the recommended design live load on each track (Reference 149).
9.1.3.4 DISTRIBUTION OF LIVE LOAD (1993)1 R(2014)
9.1.3.4.2 Ballasted Deck Structures
The recommendation for distribution of load to ballasted deck structures is based on tests performed by the AAR and reported
in AREA Proceedings, Vol. 56, 1955, page 45, other prior tests, and Report No. ER-5 of Engineering Research Division of
AAR of February, 1961.
9.1.3.4.2.3 Transverse Steel Beams
a. The above noted studies show the beneficial effects of the concrete slab in distributing the applied load for decks
supported by transverse steel beams without stringers. This is now reflected in Article 1.3.4.2.3.
The equation for D shown for moment has been introduced to account for the load carrying and load distributing
effects of the concrete slab. The first term in parentheses,
 
1 
 ----------------
 d 
 1 + ------ -
aH
indicates the amount of the total load that is carried by the beams. The remainder is assumed to be carried by the slab.
However, for this effect to be obtained, the slab must extend over at least the center 75% of the length of the floorbeam.
If there is no slab, or the slab is less than the center 75% of the length of the floorbeam (and thus essentially ineffective)
then, as designated in Article 1.3.4.2.3c, the effective beam spacing becomes d, the actual spacing, and the equation for
P is essentially the same as specified prior to the 1969 edition. The second term in the parentheses,
1
Reference 132

Commentary
 0.4 + --l- + -------H

-
 d 12 
accounts for the effect of the slab in distributing the load. The effect of beam spacing, and slab beam stiffness is shown
in this term.
In special situations, it may be necessary to design decks with transverse beams without ballast. Although the criteria
outlined in Article 1.3.4.2.3 are intended for use with ballasted deck structures, the criteria are acceptable for use with
decks without ballast.
9.1.3.4.2.4 Longitudinal Steel Beams or Girders
a. For ballasted deck structures with longitudinal beams or girders, the test data are limited. It is, therefore, inappropriate
at this time to attempt to refine significantly the criteria for distribution of live loads to these members.
The data indicate that lateral distribution of live load to longitudinal beams or girders is improved by increasing the
ballast thickness or increasing the floor stiffness, or both. The lateral distribution is also affected by the beam stiffness.
Widely spaced diaphragms consisting of beams or plates and angles are relatively ineffective in improving lateral load
distribution but improve stability and rigidity of the floor support system. For groups of beams, the live load carried by
beams more than approximately 7 feet from center line of track is of relatively low magnitude and difficult to predict
because of several factors involved in addition to those mentioned above. A primary objective of this article is to
ensure the placement of the main track supports where they are most effective.
For design purposes, it is assumed that all supports within a width defined by a line with a 1:1 slope down from the end 1
of tie through the ballast and deck, are equally loaded, even though the slope of such a line is usually limited to 1/2:1,
especially through ballast. Using the total depth and the flatter slope recognizes the additional distribution effect due to
bending and shear of the timber or concrete floor and is reasonably consistent with field test results. For floors of
timber or steel the supports generally will be spaced closer together which will reduce the required floor thickness and
result in concentrating the supports in a narrower width. It is undesirable to complicate the formula by introducing the
modulus of elasticity of the floor material, since the available test data do not justify this refinement at this time.
3
In design, all beams outside of the width defined above are assumed to carry only dead load, live load of off-track
equipment and similar loads. For simplicity of details and construction, and for possible future widening, such
additional beams should be of the same section as the main supports.
9.1.3.5 IMPACT LOAD (2007) R(2014)
a. The impact loads specified are based on investigations and tests of railroad bridges in service under passage of
4
locomotives and train loads. The early tests, prior to 1935, were made with mechanical instruments and included
measurements of deflections and strains. In general and particularly for shorter spans, the instruments were subject to
considerable error due to vibration. Later tests (Reference 129) were made with electrical instruments which permitted
more accurate measurements without disturbance from vibrations.
The impacts calculated by the formulas given in this article do not include the effect of impulsive loads that are not
substantially attenuated between the rail and the structure. For example, direct fixation of the rail to a steel deck
without some appropriately designed attentuation device is not covered. Such impacts have been measured to be as
high as 600%. (References 50, 84)
b. Tests (Reference 22) have shown that the impact load on ballasted deck bridges can be reduced to 90% of that specified
for open deck bridges because of the damping which results from the mass and resiliency of the ballast on a ballasted
deck bridge.

Steel Structures
d. The impact load due to rocking effect, RE, is due to a couple created by 20% of the wheel load acting down on one rail
and up on the other rail, which effect was called roll prior to 1967. By service tests (Reference 28), it was established
that the roll effect was essentially the same for all speeds. In 1967, the term 100/S (S in feet) was introduced as a
downward load only (Reference 27), which approximates the effect of roll used in previous recommended practices. S
was defined as the distance between centers of single or groups of longitudinal beams, girders or trusses; or the length
between supports of floorbeams or transverse girders. Because of inconsistent interpretations of the term 100/S (S in
feet) the term RE was introduced in 1991.
In accordance with Article 1.3.5a, impact load due to rocking effect, RE shall be determined as a percentage of live
load applied vertically at each rail. RE is then added to the impact load due to vertical effects (Article 1.3.5c) to
determine the total impact load expressed as a percentage of the specified live load.
The impact load due to rocking effect, RE, is created by a rocking load equal to the applied force couple of 20% of the
wheel load acting vertically at each rail. Vertical loads in members due to the rocking load can be calculated for steel
bridge span members based on the distribution of the rocking load to members supporting the track. RE can then be
expressed as a percentage of specified live load by determining the ratio of the vertical load due to rocking to the
vertical due to otherwise specified vertical live loads in each member supporting the track (for example, if the
distribution of rocking load to members supporting the track is assumed to be the same as the distribution of vertical
live load, RE expressed as a percentage of live load, will be equal for all members supporting the track).
For spans with one longitudinal beam, girder or truss per rail the impact load due to rocking effect, RE, is (100/S)% of
the vertical live load applied at each rail, where S, feet, is the distance between the centers of the longitudinal beams,
girders or trusses. The constant of 100 represents the effective rail spacing of 5 feet times the load factor of 20 percent.
For spans with more than one longitudinal beam, girder or truss per rail the impact load due to rocking effect, RE,
(expressed as a percentage of live load) depends on the distribution of rocking load and specified vertical loads to the
members supporting the track. Vertical loads shall be distributed to members supporting the track in accordance with
Article 1.3.4. The distribution of rocking loads to members supporting the track shall be based on the configuration
and spacing of members supporting the track.
For floorbeams and transverse girders the impact load due to rocking effect, RE, is (100/S)% of the vertical live load
applied at each rail, where S, feet, is the distance between the supports of floorbeams or transverse girders.
e. The requirements specified for members receiving load from more than one track are based on judgement. For a double
track span, the shortest span for which the impact load for only one track is to be used is 225 feet. For an open deck
through span of this length the use of the impact load for the second track would add approximately 5% to the total
design load of the truss. The probability that full impact load effects will occur simultaneously for both tracks is
remote, but should this happen, the resulting increase in total load is small.
9.1.3.6 CENTRIFUGAL FORCE (2002) R(2014)
a. The centrifugal force defined in Section 1.3.6 is a function of curvature and speed (Reference 27). The centrifugal
force contributes to the horizontal forces applied to the bridge through the outer rail of a curve, and affects the
proportion of the vertical force taken by each rail.
b. In cases where the maximum train speed for the expected life of the bridge on a curve is not limited by other
conditions, it is constrained by a practical maximum superelevation of 6 inches (150) and a maximum underbalance of
3 inches (75), which equates to equilibrium speed for a superelevation of 9 inches (225). At that point, regardless of
the actual curvature and corresponding speed, the proportion of centrifugal to vertical force is very close to 0.15.
Article 1.3.6(b) is based upon the assumptions that, at some time in the life of the bridge, a superelevation of 6 inches
(150) could be applied to the track, and trains could be operated at the corresponding maximum speed, with
superelevation underbalance of 3 inches (75).

Commentary
c. In cases where the maximum train speed for the expected life of the bridge is limited by other factors, the design speed
may be reduced to that specified by the Engineer in accordance with the provisions of Chapter 5 of this Manual, with
the centrifugal force factor and superelevation adjusted accordingly.
d. On superelevated curves, the point of application of the vertical load will be offset horizontally toward the center of the
curvature. Article 1.3.6(d) accounts for this offset.
e. Article 1.3.6(e) accounts for the application of the entire horizontal centrifugal force at the flange of the wheel on the
outer rail, combined with the proportion of vertical load, with full impact, taken by the outer rail. No horizontal force
is assumed at the inner rail, as any horizontal wheel forces applied at the inner rail are normally canceled by the other
wheels at the same side of that truck.
9.1.3.7 WIND FORCES ON LOADED BRIDGE (2009) R(2014)
a. The recommended use of 300 lb per linear foot for wind force on a train on a bridge as contained in Article 1.3.7a is
adequate for use on lines where double stack equipment is not operated. The engineer should consider increasing this
force in areas where double stack equipment, or other equipment with a large vertical projection operates, and strong
wind gusts are known to occur.
b. The specified basic wind pressure of 30 lb per square foot on a structure carrying live load has a long historic
background in railroad recommended practices. It was assumed that the maximum wind velocity under which train
operations would be attempted would produce a load of 30 lb per square foot on a flat surface normal to the wind. The
provisions of Article 1.3.7b (1), (2), and (3) were selected to make provisions for the effect of the wind on the portions
of the structure which are behind, and partly shielded by, the portion of the structure directly exposed to the wind 1
pressure.
c. Article 1.3.7c promotes proper proportioning of affected members in order to attain rigidity for the structure as a
whole. It does not actually address wind loads, but rather a “notional” load which was once termed “vibration load” in
earlier bridge specifications (Reference 165). This load is included in the section on wind load because it is applied as
an alternative to wind load. The affected members are to be proportioned for the greater force of either the wind load
or this “notional” load. 3
9.1.3.8 WIND FORCES ON UNLOADED BRIDGE (2005) R(2014)
The specified basic wind force of 50 lb per square foot on an unloaded structure has a long historic background in railroad
specifications. It was assumed that a hurricane wind, during which train operations would not be attempted, could produce a
load of 50 lb per square foot on such surfaces.
4
9.1.3.10 STABILITY CHECK (2005) R(2014)
a. For wind, nosing, and centrifugal forces, the vertical weight of a train on a tower or pier usually improves the lateral
stability of the structure, so it is prudent to model the least weight train that would be present with the applicable lateral
overturning load. A uniform vertical loading of 1,200 lbs/ft applied to the leeward track represents a consist of empty
cars.
For multiple track structures supported by the same pier(s) only the leeward track is loaded.
b. This stability check is designed to ensure that a load equal to half the full design load on the verge of incipient roll will
not cause the span to roll over. It is not intended to prevent damage to the structure, nor is it intended for deck design.

Steel Structures
9.1.3.11 BRACING BETWEEN COMPRESSION MEMBERS (2017)
The 2.5% proportioning force for the bracing is derived from assuming a point of restraint on a compression member with an
assumed initial out-of-straightness. The intent of the article is to establish an appropriate restraining force imparted on the
bracing members that restrain compression members against out-of-plane deformation.
For through plate girders (TPG), the proportioning force is not an additional design load to be resolved through the structure
(e.g. floorbeams or girder top flange), it is only a value used to ensure that the floorbeam-knee brace-TPG system is sized
sufficiently to act as a frame. Finite element modeling and field instrumentation of TPGs have demonstrated that a floorbeam
directly subjected to an axle load places the knee brace in tension, while unloaded floorbeams and knee braces resist torsion of
the girder by placing the knee braces in compression (Reference 156). It was also observed that relative to an axle load acting
on a deck system, the bracing forces are less on the loaded floorbeam (frame action between the floorbeam and girder) and
more on the unloaded floorbeam. Disregarding secondary forces in the TPG knee brace does not affect the design.
9.1.3.12 LONGITUDINAL FORCES (2005) R(2014)
The longitudinal force used in previous editions of this Manual of recommended practice has changed over time. In the 1905
edition, the force was 20 percent of the specified total live load. By the 1920 edition reductions were permitted for ballast
deck spans and for short structures. In the 1932 edition, the additional force of 25 percent of the driving axles of the Cooper’s
series was introduced, and the braking force of 15 percent of the Cooper’s train was introduced.
The AAR conducted a number of tests with the secondary objective of measuring longitudinal forces in the 1940’s and 1950’s.
None of these tests were conducted under conditions that would have approached the maximum possible longitudinal force
available at that time.
It became the practice of some railroads to use one half the specified force which by this time was 25 percent of the Cooper’s
driving axles or 15 percent of the Cooper’s train on the appropriate loaded length.
In the 1968 edition of the Manual, a factor L/1200 was introduced to be applied to the 15 percent of the Cooper’s train with an
exception for bridges with discontinuous rail (e.g.: movable bridges and those with sliding joints or switches). This resulted in
a vastly reduced longitudinal force requirement. The traction force of 25 percent of the weight on driving axles was
eliminated.
A similar change was made to Chapter 8, Concrete Structures and Foundations. Committee 7, Timber Structures, did not make
changes to the recommended practice in Chapter 7.
With the introduction of high-adhesion locomotives, load/empty brakes, and ECP brakes, concern was expressed that
recommended forces were not high enough. Several railroads have acknowledged component failures in bridges due to
longitudinal forces, and several structures have been replaced due to distress caused by high longitudinal forces.
In 1996, the AAR conducted a test specifically to investigate longitudinal forces under the newly developed AC diesel-electric
freight locomotives. The test demonstrated that a longitudinal force of about 100 kips (440 kN) on a 50-foot (15-meter) open-
deck span was more than 25 times the design force in the 1996 edition of the Manual.
Under direction from the Board of Directors, and with the concurrence of the chairmen of the structural committees (7, 8 and
15) who formed the nucleus of a quickly gathered ad-hoc committee, AREA revised its recommended practice for the 1997
edition to conform to this test result. Chapter 7 was thought to be appropriate and not warrant the emergency change.
The AAR followed this test with further tests, all of which confirmed the much higher longitudinal forces, and the far greater
percentage of those forces which went into the structure. On a four-span, 400-foot (122-meter) open-deck steel bridge,
longitudinal forces up to 330 kips (1470 kN) were measured in the entire structure, with up to 220 kips (980 kN) in a 210-foot
(64-meter) truss, and up to 110 kips (490 kN) in a 42-foot (13-meter) beam span. On a two-span, 121-foot (37-meter) open-
deck steel DPG bridge, forces up to 140 kips (620 kN) were measured in the entire structure, with up to 96 kips (430 kN) in a

Commentary
55.5-foot (17-meter) span. On a single-span, 60-foot (18-meter) ballast-deck steel DPG bridge, forces up to 115 kips (510 kN)
were measured. All tests used sets of two or three AC locomotives, operated near their maximum tractive effort capabilities of
180 kips (800 kN) to 200 kips (890 kN) per locomotive. Further information about these tests can be found in Reference 68,
86, 96, 97, 107, 108, 109, 110, 111, 112, 113, 114, 115, 117, 119, 145, 148, 150, 159, 160 and 161.
The results of these tests indicated the following:
(1) Ballast deck spans do not have lower longitudinal forces
(2) Short spans do not have significantly lower longitudinal forces
(3) Half the force is not always dissipated through the rails
(4) There has been considerable confusion over the difference between the force and its distribution and path
(5) High longitudinal forces are not necessarily grade related
(6) High longitudinal forces are related to lower speeds for tractive effort and dynamic braking situations. When a
train is maintaining a speed that exceeds 15 mph (25 km/h) it cannot exert the maximum tractive effort. To cover
future developments, the recommended practice has used 25 mph (40 km/h).
(7) High longitudinal forces due to braking can occur at any location, particularly if an emergency brake application
occurs
(8) The ability of the approach embankments to resist longitudinal forces from the superstructure is reduced as
1
longitudinal forces are also applied to the approach embankments. This would be the case with several
locomotives passing over a short bridge, or a train braking.
Analytical confirmation of the above behaviors has been done by Foutch et al (Reference 67, 68, 113, 150 and 158), and is also
explained by Fryba (Reference 69). Unfortunately, these formulations are too cumbersome for routine work. The problem can
be envisioned as the rails being continuously supported in the longitudinal direction. The longitudinal stiffness of the 3
connection between the rails and the bridge is similar to the stiffness of the connection between the rails and fixed ground on
the approach embankments.
With regard to braking force, the maximum adhesion between wheel and rail is about 15 percent. This level of braking would
typically be reached with an emergency application of the train air brakes. The equation for train braking is derived using 15
percent of the Cooper live loading.
4
Longitudinal force due to braking acts at the center of gravity of the live load. Center of gravity height is taken as 8 feet (2500
mm) above the top of rail. This force is transferred from vehicle to rail as a horizontal force at the top of rail and a vertical
force couple transmitted through the wheels.
Longitudinal force due to locomotive traction acts at the drawbar. Drawbar height is taken as 3 feet (900 mm) above top of
rail. As with braking, this force is transferred from vehicle to rail as a horizontal force at the top of rail and a vertical force
couple transmitted through the wheels.
Longitudinal forces transmitted by tractive effort of locomotives or the braking action of trains will be distributed to bridge
members in accordance with their relative stiffness and orientation with respect to the force path between the applied
longitudinal force and the supporting substructure.
The length “L” in Article 1.3.12 is to be taken as the appropriate length for the structure or portion of the structure under
consideration. The length selected should be the one that produces the maximum force in the structure or portion of the
structure under consideraton.

Steel Structures
In bridges with stringer and floorbeam floor systems, longitudinal forces are first applied to the stringers. The force must then
be transferred to the members to which stringers are connected, usually the floorbeams. Traction bracing can be used to
directly transfer the longitudinal force from the stringers to truss or girder panel points.
In bridges with transverse floorbeam floor systems (such as through girder spans), traction bracing can be used to transfer the
longitudinal force to the bridge members supporting the floorbeams (typically girders).
It is generally considered good practice to design traction bracing to be the same depth as the member being braced. However,
when traction bracing isn’t used, the floorbeams should be designed for transverse bending and torsion where applicable. See
Reference 73. It is generally considered good practice to provide traction bracing rather than design floorbeams or transverse
members for lateral bending and torsion.
Fixed bearings and their anchorages should be designed to transfer the longitudinal force from superstructure to substructure.
In addition to designing the fixed bearings to take all the longitudinal forces, it is the practice of some engineers, given that
bearings tend to become frozen or stuck with time, to design the area around and below expansion bearings for a percentage of
the longitudinal forces going through those bearings as though the expansion bearings were partially fixed.
Longitudinal forces are of importance in railway trestle bridges and may govern the economic span length considering
requirements for longitudinal bracing and column sizes in towers.
9.1.3.13 FATIGUE (2016)
The following symbols are used within Article 9.1.3.13:
N = Number of occurrences of constant stress cycles which would cause fatigue damage equivalent to
the fatigue damage caused by a larger number, Nv, of variable stress cycles
Ni = Maximum number of stress cycles for each of the stress range values represented in the
distribution being considered
ni = Number of stress cycles for each of the stress range values represented in the distribution being
considered
Nv or n = Total number of variable stress cycles in the distribution or life
SR = Stress range, the algebraic difference between the maximum stress and the minimum stress for a
stress cycle
SRact = Stress range actually created at a given location in the structure by a moving load
SRfat = Allowable fatigue stress range as listed in Table 15-1-10
SRi = Stress range of cyclic stress corresponding to the number of occurrences, ni
SRe = Effective cyclic stress range for the total number of variable stress cycles, Nv .
SRRMC = 3 3
  n i S Ri 
(Root Mean ------------------------
Cube Stress n i
Range)
 = SRact/SR or Eact/Eapplied ratio when SR is calculated by using the same load which was applied
when SRact was measured. Field measurements have shown the measured SR is equal to a factor,
times the calculated SR. This reduction reflects the beneficial effects of participation by the
bracing, floor system, or other three-dimensional response of the structure and, also, the fact that
full impact does not occur for every stress cycle. Since SR at a given location is directly
proportional to the loading used, Eact/Eapplied also equals this ratio.
 i = The ratio of the number of occurrences of SRi to the total number of variable stress cycles, Nv

Commentary
SRmin = stress range or lower limit value for the starting point of the function being considered
SRN = Stress range which corresponds to N constant stress cycles for a given detail
a. Fatigue is now covered in a sufficient number of texts so a basic explanation is no longer needed in this Commentary.
It was removed in 2009.
Between 1910 and 1969, inclusive, this Manual required an increase of each stress by 50% of the smaller stress for
members subject to reversal of stress.
Fatigue damage prior to the introduction of 263,000 lb. cars (100 ton capacity) in the 1960’s was primarily the result of
the passage of heavier locomotives. With 20 trains a day for 60 years, the number of damaging cycles caused by
locomotives would be less than 500,000. Most freight cars were light enough to cause little if any damage due to
fatigue.
In 1969, methods were introduced based on the R ratio, the ratio of minimum to maximum stress, and a distinction was
made between cases of more than 500,000 cycles of load or less (Reference 102). Consideration of the methods used to
fabricate and connect members was included.
Fatigue problems accelerated in the 1970s, with the introduction of heavy and frequent unit train service where the
influence of each car produces a significant effect. With the same 20 trains a day with 60 cars per train causing
damage, 500,000 cycles could accumulate in one year for some members.
Fatigue design for this Chapter has been based entirely on the nominal stress range concept since the 1978 Edition.
Other factors, such as mean stress and steel strength have negligible effect in the types of fabricated structures used in
the railroad industry. The type of load distribution has been revised as new knowledge has been developed. 1
Structures designed to the fatigue criteria of Article 1.3.13 should be adequate for:
(1) continuous unit trains with axle loads not exceeding 80,000 lb for loaded lengths less than 100 feet,
(2) continuous unit trains with equivalent uniform load not exceeding 6,000 lb per foot of track and axle loads not
exceeding 80,000 lb, or other variations of higher load with fewer cycles on longer spans (see Article 9.1.3.13c). 3
This should be adequate for mainlines of Class I railroads, and for most heavy haul lines.
The Chapter recommends special consideration for spans exceeding 300 feet (see Article 9.1.3.13j).
b. The major factors governing fatigue strength are the number of stress cycles (covered in section c), the magnitude of
the stress range (section h), and the type of constructional detail (section g). 4
c. The derivation of the design criteria for fatigue did not consider Rail Transit or other Light Rail facilities. For such
cases, unless demonstrated otherwise, the number of constant stress cycles, N, should be > 2,000,000.
For typical North American freight railroads the number of cycles used for design were derived assuming 315,000 lb.
cars in 110 car trains at a frequency of 60 trains per day over an 80-year period. The number of cycles per train is the
result of extensive work done by G. Oommen, S. Beisler and R.A.P. Sweeney as reported to Committee 15 in 1987 and
1988 (See Table 15-9-1). This criterion will theoretically provide infinite life for all loaded lengths less than 100 feet
and will accommodate longer and more frequent trains.
Existing cars (1988) with gross weights of 315,000 lb and certain double stack cars are approaching E 80 loading
values on short spans. In order to provide sufficient fatigue capacity under solid, or “unit” trains of these types of
vehicles the number of design cycles shown in Table 15-1-7 was derived by prorating the fatigue curve formula,
N = Nv x ( x SE60/SE80)3

Steel Structures
to an equivalent number of cycles of E 80 loading. In this formula N is the number of cycles,  is a constant, and SE60
and SE80 are respectively the stress ranges characteristic of E 60 and E 80 loading. The total projected number of
variable stress cycles, Nv, shown in Column 5 of Table 15-9-1, is obtained by multiplying columns 2, 3 and 4. The
value of alpha is to be taken as one unless a test on the member being evaluated indicates that a lower value is
appropriate.
Table 15-9-1 is based on 110-vehicle train. Critical characteristic load is assumed to be ¾, i.e. 60/80, of design load E
80.
On spans exceeding 100 feet it may be necessary to increase the number of cycles per train if a consistent operating
pattern of loaded cars followed by empty cars is repeated throughout the design train, throughout the service life of the
bridge. It is theoretically possible to get 55 cycles on spans close to 100 feet if the pattern is 2 loaded cars followed by
2 empty cars. Nevertheless, the committee assumed 3 cycles of loaded-empty combinations in its design 110-car train
as a more likely maximum on spans exceeding 100 feet.
Table 15-9-1. Parameters Used to Develop Table 15-1-7 and Table 15-1-10
Classification I
1 2 3 4 5 6 7 8
Stress N N
Span Length Life No. of
Cycles per Projected Nv Alpha Col. 5  used in
L in Days Daily
Train Million () (Alpha x 6/8)3 Table 15-1-7
Ft 80 Yr Trains
Crossing Million Million
L > 100 29,200 60 3 5.3 1.0 2.2 2
100 L > 75 29,200 60 6 10.5 1.0 4.4 >2
75 L > 50 29,200 60 55 96 1.0 41 >2
50L 29,200 60 110 193 1.0 82 >2
It must be recognized that the number of variable cycles leading to the greater-than-2-million category in Table 15-1-
10 is different for each category of detail, varying from 3 to 31 cycles per 110-car train on spans exceeding 100 feet.
For designs exceeding E80, the reduction shown in Column 7 in Table 15-9-1 is not appropriate over the life of the
span. For designs exceeding E80, all span lengths should be designed for the > 2 million category listed in Tables 15-1-
7 & 15-1-10.
d. Impact values used in design are estimated to have a probability of occurrence of 1% or less. Considering that a
railroad bridge is normally designed for an 80-year period, this level of impact is quite likely to occur at least once
during the bridge life and probably more frequently. For fatigue design the mean value of impact is more appropriate.
Nevertheless, the note to Table 15-1-8 covers cases of consistent and continuous poor maintenance practice with regard
to wheel or track maintenance or places where there are joints in the rail due to switches or rail expansion or other
joints where higher impact is a frequent occurrence. This is likely to include but is not restricted to locations where
there is "FRA Excepted Track" or "FRA Class 1 Track."
In locations where a structural member supports or is influenced by a “Conley” or similar style joint or where there are
rail break castings, a rail end connection or similar style joint or switch, the reduction in impact shown in Table 15-1-8
should not be used.
For members supporting end ties on movable spans and at the adjacent ends of fixed spans, use the full impact outlined
in Article 6.3.3, unless test results show a lower permissible impact.

Commentary
Observations on 37 spans with span lengths between 30 and 140 feet, summarized by W. G. Byers (Reference 38),
indicates that mean impact values fall below 65% of the values used for design. Tests included results obtained with
poor wheels and on poor track.
Tests on 15 bridges on Canadian National Railways done between 1975 and 1988 (reported to Committee 15, May
1988, by R. A. P. Sweeney) indicated mean values of 34% on spans less than 80 feet, and 65% on longer spans. A
further presentation by Dr. Sweeney made to the committee in 2002 based on tests on over 100 bridges confirmed the
numbers in revised Table 15-1-8, and confirmed that alpha should be assumed to be 1 unless a particular structure was
tested and alpha proved to be lower for that structure. (References 146, 147)
The mean impact is a function of the geometry of the track and how well it is maintained up to and across the bridge,
along with the maintenance standards for out-of-roundness of wheels and for wheel flats. The more restrictive
limitations placed on short members without load sharing is based on the probability that a single wheel may cause
such values relatively frequently. This is based on a one-year sample of wheel impact data at 10 Wheel Impact Load
Detector (WILD) Sites on CN (Reference 42).
For Light Rail, the Mean Impact Load shown in Article 1.3.13d should be 100% of the impact load specified in
Article 1.3.5 for all member span lengths, and in Table 15-1-7.
e. The fatigue criteria is based on continuous unit trains with equivalent uniform load not exceeding 6,000 lb per foot of
track and axle loads not exceeding 80,000 lb and has been adjusted so that the Standard Cooper load specified in
Article 1.3.3 may be used for design purposes.
f. For the usual design condition of members subjected to bending, only SR derived from bending needs to be considered
for details, such as transverse stiffeners, which are subjected to shear stresses as well. The design detail categories have 1
taken shear into account; therefore, principal stresses need not be considered in the usual design condition. For unusual
design conditions, the principal stresses may need to be considered.
Residual and/or locked-in stresses induced during welding, fabrication or erection shall not be explicitly considered in
investigating fatigue. Residual stresses due to welding are implicitly included through the specification of stress range
as the sole dominant stress parameter for fatigue design. This same concept of considering only stress range has been
applied to rolled, bolted, and riveted components or details where far different residual stress fields exist. The 3
application to nonwelded components or details is conservative.
It has been shown that the level of total applied stress is insignificant to fatigue design for a welded steel component or
detail in structures typically designed using this Manual.
A complete stress range cycle may include both a tensile and compressive component. Only the live load plus impact
stresses need be considered when computing a stress range cycle; dead load does not contribute to the stress range. 4
Tensile stresses propagate fatigue cracks. Material subjected to a cyclical loading at or near an initial flaw will be
subject to a fully effective stress cycle in tension, even in cases of stress reversal, because the superposition of the
tensile residual stress elevates the entire cycle into the tensile stress region.
These provisions shall be applied only to components or details subjected to a net applied tensile stress. In regions
where the permanent loads produce compression, fatigue shall be considered and these provisions applied only if the
tension component of the live load plus impact stress range cycle due to fatigue exceeds the permanent-load
compressive stress in the component or at the detail under consideration.
Fatigue design criteria need only be considered for components or details subject to effective stress cycles in tension
and/or stress reversal. If a component or detail is subject to stress reversal, fatigue is to be considered no matter how
small the tension component of the stress cycle is since a flaw in the tensile residual stress zone could still be
propagated by the small tensile component of stress. Hence, the entire stress range cycle (which may include
compression) is used in computing the stress range. In addition, for fatigue to be considered, the component or detail
must be subject to a net applied tensile stress under an appropriate combination of the permanent loads and the fatigue
live load. The tensile component of the stress range cycle resulting from live load and its appropriate impact

Steel Structures
combination acting in conjunction with the compressive stress due to the permanent loads are used to establish the
presence of a net applied tensile stress in the component or at the detail under consideration.
Cross-frames and diaphragms connecting adjacent girders are stressed when one girder deflects with respect to the
adjacent girder. The sense of stress is reversed depending on which way roll is applied and this usually creates the
largest stress range in these members. To cause one cycle of the stress range so computed requires two vehicles to roll
in opposite direction. This has been observed in practice. For cases where the force effects in these members are
available from an analysis, such as in horizontally curved or sharply skewed bridges, it may be desirable in some
instances to check fatigue-sensitive details on a bracing member subjected to a net applied tensile stress. In no case
should the calculated range of stress be less than the stress range caused by full live load and appropriate impact load.
g. Components and details susceptible to load-induced fatigue cracking have been grouped into nine categories, called
detail categories, of similar fatigue resistance established through full scale testing (Reference 9, 56, 90, 134, 172 and
175).
Table 15-1-9 illustrates many common details found in bridge construction and identifies potential crack initiation
points for each detail. In Table 15-1-9, “Longitudinal” signifies that the direction of applied stress is parallel to the
longitudinal axis of the detail. “Transverse” signifies that the direction of applied stress is perpendicular to the
longitudinal axis of the detail.
Where fasteners and connected material are proportioned in accordance with Article 1.3.13 and Section 1.4, Basic
Allowable Stresses, the fasteners will have greater fatigue life than the connected material (Reference 91). Thus, no
categories for bolts or rivets in shear or bearing are required to replace the 1969 formulas.
For information on Partial Penetration (PJP) joints see Article 1.7.4 and its commentary.
Research on end-bolted cover plates is discussed in Reference 167.
h. The requirement that the maximum stress range experienced by a detail be less than the constant-amplitude fatigue
threshold provides a theoretically infinite fatigue life for all loaded lengths less than 100 feet.
For longer spans see Article 9.1.3.13c.
For cases where different criteria are appropriate, the fatigue resistance above the constant amplitude fatigue threshold,
in terms of cycles, is inversely proportional to the cube of the stress range, e.g., if the stress range is reduced by a factor
of 2, the fatigue life increases by a factor of 23. This is reflected in the equation shown below and shown in
Figure 15-9-3.
(Sr) = (A/N)1/3

Commentary
1
Figure 15-9-3. Stress Range vs. Number of Cycles for Various Detail Categories
Sr-N curves in Figure 15-9-3 were developed (Reference 9, 55, 56, 58, and 59) by using 95% confidence limits for
97.5% survival applied to full-scale test data.
Table 15-9-2. Constant A and Thresholds for Detail Categories 3
DETAIL CONSTANT, A THRESHOLD

CATEGORY TIMES 108 (KSI3) (CAFL) ksi
A 250.0 24
B 120.0 16
4
B’ 61.0 12
C 44.0 10
C’ 44.0 12
D 22.0 7
E 11.0 4.5
E’ 3.9 2.6
F 9.0 8.0
A 325 Bolts
in Axial Tension 17.1 31
A 490 Bolts
in Axial Tension 31.5 38

Steel Structures
Detail Category F is for the allowable shear stress range on the throat of a fillet weld. When fillet welds are properly
sized for strength considerations, Detail Category F should not govern. Fatigue will be governed by cracking in the
base metal at the weld toe and not by shear on the throat of the weld.
i. Detail Category E and E’ details shall not be used on fracture critical members, and Detail Category D details shall be
discouraged and used only with caution. Such details are highly susceptible to fatigue damage.
Eye bars and pin plates are design details which are not recommended except for very long truss spans where live load
stress ranges are very low. In the event of their use, see Article 7.3.3.2 and the appropriate Commentary
Article 9.7.3.3.2.
j. For span lengths exceeding 300 feet an analysis is required for each bridge component using influence lines and the
preceding car types and load frequencies, accounting for the effect of lightly loaded vehicles interspaced within the
design train.
k. When proper detailing practices are not followed, fatigue cracking has been found to occur due to strains not normally
computed in the design process. This type of fatigue cracking is called distortion-induced fatigue. Distortion-induced
fatigue often occurs in the web near a flange at a welded connection plate for a cross-frame where a rigid load path has
not been provided to adequately transmit the force in the transverse member from the web to the flange. These rigid
load paths are required to preclude the development of significant secondary stresses that could induce fatigue crack
growth in either the longitudinal or the transverse member (Reference 60). It is emphasized that the stiffness of this
connection is critical to prevent relative displacement between the components.
9.1.3.13.1 High-Strength Bolts Subjected to Tensile Fatigue Loading
Previous versions of this article were based on the 1969 edition of the Manual and distinguished between connections subject
to less than or more than 500,000 cycles and were based on maximum applied stress.
The current limits are based on applied stress range with a maximum set at the constant amplitude fatigue limit for these bolts.
Prying force was taken as 20% pending further testing.
The formula for the tensile stress area or effective bolt area is: 0.75 * cross-sectional area based on nominal bolt diameter
(Reference 155).
9.1.3.13.2 Anchor Bolts Subjected to Tensile Fatigue Loading
This article is intended for anchor bolts subjected to repeated tension cycles. Examples of anchor bolts with tensile fatigue
include hammerhead piers, outrigger bents, or continuous spans. Tension fatigue failures have been noted for anchor bolts on
hammerhead piers using concrete columns with steel caps. The same situation can control in piers using concrete columns and
steel pier caps such as in outrigger bents. This article is to ensure the assumed distribution of load to all anchor bolts on a pier
cap or abutment that are fastening base plates for bearings intended to resist fluctuating tensile loads. It is not intended for the
general situation of anchor bolts on a pier cap or abutment that are fastening base plates for bearings.
Experience with anchor bolt connections on hammerhead piers demonstrated that fatigue failures occur as a result of
inadequate and highly variable bolt pretension within a group of bolts. Cracked bolts had modest levels of bolt tension whereas
most of the other bolts in the group had little or no pretension, with the partially pretensioned bolt resisting more than the
intended share of live load. This resulted in fatigue crack development and fracture. To accomplish the appropriate pretension,
where leveling nuts are used to position a steel pier cap on the column, steel shims may be inserted and the leveling nuts
backed off before the grout pad is installed and anchor bolts are tensioned. Alternatively, the leveling nut anchor bolts can be
ignored in the design for resisting the applied loads.
If stainless steel is used for anchor bolts, AISI Stainless Steel 316 is the best choice for salt water exposure. This material
should be available in bar stock to fabricate anchor bolts. It is identified in the ASTM A193 Standard Specification for Alloy-

Commentary
Steel and Stainless Steel Bolting Materials for High Temperature or High Pressure Service and Other Special Service
Applications. Other stainless steels in the A193 Specifications are not recommended.
9.1.3.14 COMBINED STRESSES (2005) R(2014)
9.1.3.14.1 Axial Compression and Bending
f f f f
The straight line interaction formula ----a- + -----b  1.0 is acceptable for small values of ----a- , but for values of ----a- greater than
Fa Fb Fa Fa
0.15, the deflection of the column and the resulting increase in bending stresses caused by the axial load being made eccentric
1
must be taken into account. The formula accomplishes this by applying a magnification factor ----------------------------------------------------- to
fa  kl
2
1 – ----------------------- ----- 
2  r
0.514 E
f 1
-----b . This factor is similar in form to the formula ---------------- (Reference 41) in which F e is the elastic (Euler) buckling stress of
Fb f
1 – ------a-
F e
2
0.514 E
the column loaded axially, divided by the applicable factor of safety, or ----------------------- in these recommended practices (see
2
 kl
----- 
r
Article 1.4.1). 1
When a member is braced in the plane of bending, at a panel point for example, there is no column deflection and, therefore,
the magnification factor does not apply. Furthermore, the allowable axial stress here may be based on kl  r = 0 . The
applicable formula then becomes f a  0.55F y + f b  F b  1.0 . It should be noted that this formula does not apply at a
connection point which is coincident with the location of maximum curvature of the deflected column axis, because such a
point is not, in effect, braced. 3
The above remarks cover bending about one axis only. For bending about both axes, the three-term formulas obtained by
expansion are sufficiently accurate for use.
9.1.3.15 SECONDARY STRESSES (1994) R(2014)
This article provides that secondary stresses due to truss distortion usually need not be considered in any member of the width 4
of which, measured parallel to the plane of distortion, is less than 1/10 of its length. An exception to this general provision
should be the effects of secondary truss members, such as floorbeam hangers and subverticals; these may produce excessive
secondary stresses in the chord unless adjustment is made in lengths of the verticals.
9.1.3.16 PROPORTIONING OF TRUSS WEB MEMBERS (2004) R(2014)
In determining whether it is safe to keep an old structure in service, the rules of Part 7, Existing Bridges; Section 7.3 Rating,
govern. Experience with older structures designed for lighter live loads, shows that in such structures the web members of
trusses reach their capacity sooner than other portions. This situation can be remedied either by providing an initial design of
all members for an increased live load at higher stresses or by providing a truss design under which the web members reach
their safe live load capacity at substantially the same increased live load as the remainder of the truss. The latter method is
more economical and is provided by the recommended practices requirements (Reference 77).

Steel Structures
9.1.4 BASIC ALLOWABLE STRESSES
9.1.4.1 STRUCTURAL STEEL, FASTENERS AND PINS (2017)
In determining the allowable stresses, the value of 1.82, which is equal to 1/0.55, has been adopted as the usual factor of safety
in tension, based on the minimum yield point of the material. The same value has been used for such compression applications
as are not affected by axial combined with bending effects.
Yielding of the gross area and fracture of the effective net area are considered the failure limit states. Yielding of the gross
area can lead to excessive elongation of the member. This uncontrolled elongation can precipitate failure of the overall
structural system. Fracture of the effective net area was proposed by Munse and Chesson (Reference 40, and 101) and has
been long since adopted as a limit state by both the AISC (References 13, 14, and 15) and AASHTO (Reference 7). The
allowable stress of 0.47 Fu has been adopted by AREMA to align with AASHTO and to provide an additional factor of safety
due to the sudden nature of this failure state.
The more conservative design approach for pin-connected members is based on the results of experimental research
(Reference 85).
Since there have been more failures in floorbeam hangers with connections that do not have pretensioned high-strength
fasteners than in other members, a greater apparent factor of safety has been adopted for such members (Reference 31).
From 1935 to 1969, the secant formula, and parabolic type formula approximating it, formed the basis for the column formula
of these recommended practices. It has been somewhat difficult to use and an assumed value of ec/r2 such that reasonable
values result for intermediate column lengths makes the allowable stress on short columns less than necessary. For these
reasons, and because long columns and eccentrically loaded columns can be provided for by Euler type formulas and
interaction formulas, respectively, without resort to the secant formula, the use of the secant formula was discontinued.
The column curve of the Column Research Council (now titled Structural Stability Research Council) (Reference 43) which
can be expressed in the symbols adopted in these recommended practices:
2
F y  kl  2
f = F y – ------------
- -----
2  r
4 E
was selected as the basic curve for the development of the formulas used in these recommended practices. Studies were made
which included plots of this curve with variable factors of safety such as that used by AISC (Reference 8), and with constant
factors of safety 1.8, 1.9 and 2.0. Many varieties of column curves were plotted on the chart on which these Column Research
Council curves had been plotted, and it was decided that the most practical form to be used was one involving the three
formulas of the recommended practice.
The difficulty of evaluating k in railroad bridge compression members may lead to allowable stresses that are too high,
especially in the approximate range of kl/r between 40 and 100, where a slight variation in k will have a large effect on the
allowable stress. Some protection against this danger is provided by the adopted straight line formula as compared to the
Column Research Council curve.
The formula to be used in determining the allowable compressive stress in the extreme fibers of welded built-up or rolled
beam flexural members symmetrical about the principal axis in the plane of the web (other than box-type members) is based
on theoretical studies made by Professors George Winter and Bruno Thürliman. In Professor Winter’s discussion of a paper by
Karl de Vries, he developed formula (Reference 173) for fc, the critical stress for failure of the beam. This formula may be
written:

Commentary
 2 2 1  2
 E 2  I y  2 2 KI
E - ---------------------------- l - 
2
f c =  --------------- ------- + -------------- y
-  -----
 2I x   
  l2  l  2 2  1 +  I x 2 d 
 2  ---  2 ---
 d 
d
where:
K = torsional constant
 = Poisson’s ratio
Professor Thürliman (Reference 29) has shown that this formula may be expressed in the form of
2 2
fc = w + v
where:
w = extreme fiber stress resulting from warping torsion, where the compression flange bends and the beam
warps
v= extreme fiber stress resulting from pure torsion.
Thus, the critical extreme fiber stress may be considered to be represented by the length of the hypotenuse of a right triangle, 1
whose sides are w and v , and to be equal to or greater than either of them. Under certain conditions, one or the other may be
negligible, so that the value of fc cannot be less than the greater value.
If v , is assumed negligible (i.e. = 0), then the critical stress is
2
fc = w E
2 Iy 
= --------------  -------
2
E  r y 
= --------------  ---------
-
3
 
 l  2 2I x  l  2  2r 2 
2 --- 2 --- x
 d  d
For I shaped members, rx = 0.4d (approx.), so that
1.56E
2 4
f c = --------------------
2
 l  ry 
Based on a factor of safety of 1.8, the allowable stress becomes
2
0.87E
--------------------
2
 l  ry 
This formula is of the Euler type, and the allowable stress so determined must be modified so that it will be limited by the yield
point of the material involved. A parabolic transition curve of the form Fb = A–B (l/ry)2 from the value Fb = 0.55 Fy at l/ry = 0,
and tangent to the Euler type formula curve, is the most acceptable form for this transition curve. This parabola intersects, and
is tangent to, the Euler curve at
E
l  r y = 5.55 ----- , and the values of A and B are such that
Fy

Steel Structures
2
0.55F y  l  2
- ----
0.55F y – ------------------
2  
6.3 E r y
is the first expression applying to this case in Article 1.4.1. Since Article 1.7.1b limits flexural members to those with an l/ry
E , the Euler type formula is not part of the requirements.
not greater than 5.55 -----
Fy
The second compression formula in Article 1.4.1 applying to this case is based on the Winter formula with the assumption that
w is negligible (i.e. = 0), so that the critical stress is
E 
2 KI y  1  2 l 
f c = ---------------  -----------------------------  ------
 d 
l  2  2  1 +  I x 
2

2 ---
 d
and, with only minor error.
2 3
K = --- bt
3
3
tb
I y = 2 -------
12
2
I x = 2bt  d  2 
 = 0.3
Ebt 0.207E
so that f c = ----------------------------- = ---------------------
l ld  bt
2  ---  2.42d
2
 d
0.115E
and the allowable stress, based on a factor of safety of 1.8 and with bt = Af, is --------------------- which is the second of the formulas
ld  Af
in Article 1.4.1 applying to this case.
Since tests have shown that the pure torsional (v) effect on a riveted member is modified considerably by slip in the riveted
connections, only the first type formula is considered suitable for use with riveted construction, and Article 1.4.1 so limits this
case.
For box type flexural members, the stiffness of the member is usually such that the full allowable stress (= 0.55 Fy) can be used
for both flexural tension and compression, without reduction. However, very slender and deep box type flexural members may
require reduction comparable to that of a single plane I type flexural member, and it is necessary to determine the effective
slenderness ratio (defined herein as (l/r)e) of such members by calculating the (l/r)e) value as defined in Article 1.4.1. This
effectiveness slenderness ratio is also the slenderness ratio determining the critical stress in the formula derived above

Commentary
2
1.56 E
f c = --------------------
2
 -l
 r e
for beams in which the pure torsion effect is negligible. This critical stress for box girders is calculated to be (Reference 43):

f c = ------- JGEl y
lS x
2
4A
where J = torsional constant = -----------
s  t
E
G = ---------------------
21 +  
I, Sx, A, Iy and s/t defined in Article 1.4.1.
Equating these two values for fc:
2
1.56 E 
-------------------- = ------- JGEl y
2 lS
 -l x
1
 r e
and making the indicated substitutions, the value of the effective slenderness ratio shown in Article 1.4.1 is solved to be:
 -l = 1.105lS x s  t
-----------------------------------------
-
 r e
Iy
A -----------------
3
1 + 
The allowable stress in bearing between rockers and rocker pins was adapted from editions prior to the 1969 edition and the
low value of 0.375 Fy was retained to minimize pin wear. Pin wear had historically been a cause of trouble when higher values
for this condition were permitted. Refer to Part 5 for additional information.
4
The allowable shears in F3125 Grade A325 and Grade A490 bolts are based on recommendations of the Research Council on
Structural Connections of the Engineering Foundation. Also see Reference 91 and 122.
The allowable stress in bearing on expansion rollers and rockers was based on static and rolling tests on rollers and rockers
(Reference 20 and 91). The average vertical pressures over calculated contact areas for loads substantially less than allowable
design values are in excess of the yield point, causing a flow of the material. It was concluded that the resulting “spread” of the
roller and base, measured parallel to the axis of the roller at points near the surfaces in contact, was the most satisfactory
phenomenon to use in determining design values. Such “spreads” or deformations were measured in units of 0.001 to, per inch
per 1,000 strokes, each stroke corresponding to a roller movement of 4 inches and an equal movement back. Design values
according to the tests would give total deformations varying from about 3 units to less than 1.
9.1.4.2 WELD METAL (1994) R(2014)
The allowable stresses on weld metal specified in Article 1.4.2, Table 15-1-14 are close to those permitted by AWS D1.5.

Steel Structures
9.1.4.3 CAST STEEL (1994) R(2014)
In the 1969 edition, because of better control over casting practices, the allowable unit stresses for cast steel in bearing or
compression were increased from 0.9 to 1.0 of those for rolled steel, and for all other types of stress, from 2/3 to 3/4 those for
rolled steel.
9.1.5 GENERAL RULES
9.1.5.4 THICKNESS OF MATERIAL (2011) R(2014)
a. The 0.335 inch thickness limitation was introduced in 1969 to accommodate the use of certain wide flange beam
sections as timber stringer replacements given that the assumed life of a timber structure was less than the assumed life
of a typical steel structure. It is not the intention of this article to preclude this application on timber trestles.
b. The usual design checks for a gusset plate and each member framing into the gusset plate are:
• Normal Loads on a Section (t Lw), often referred to as the Whitmore Section as shown in Figure 15-9-4, where Lw is
the effective width and t is the thickness of the gusset plate.
• Block shear, for compression and tension loads.
• Shear on critical shear planes.
• Buckling on the average of L1, L2, L3, as shown in Figure 15-9-4 with buckling factor of k to be evaluated for the
gusset plate (k might be greater than 1.0).
• All edge distance and end distance requirements for fasteners should be followed.
• Gusset plates should be as compact as possible.
• Check various critical sections using an acceptable method.
• Check fastener prying action.
• Each gusset plate is unique and must be designed based on specific forces, geometry and details.

Commentary
Figure 15-9-4. Bolted and Riveted Gusset Plates
3
For further information on bolted and riveted gusset plates see Reference 83. Design of gusset plates using the
procedure in Section 13 of “Structural Steel Designers Handbook” Third Edition, edited by R L Brockenbrough and F S
Merritt, written by Prickett, Leroy and Kulicki, provides a reasonable approach.
9.1.5.8 NET SECTION (2005) R(2014)
Net section is discussed by Chapin (Reference 39). He gives the history of this method of obtaining the net section of a riveted 4
tension member to take account of the weakening effect of staggered open holes. He gives the rather complicated formula
which represents the theoretically correct solution of the problem, and states that the simplified formula used in the
recommended practices gives approximately the same results. The application of the formula to bolted fabrication and the 85%
limit were based on later tests. A chart for use with the formula is included with his discussion.
9.1.5.9 CONNECTIONS AND SPLICES (2014)
e. In 2003, provisions to evaluate block shear were added to the Manual. Tests indicate that it is reasonable to add the
yield strength on one plane(s) of a connection to the rupture strength on the other plane(s) of a connection to predict the
block shear strength of a connection (References 11, 74, 123). Coefficients of 0.35 for shear yield and 0.55 for tensile
yield were used.
Updates in 2014 included changing the coefficient for shear yield to the more conservative 0.30 for closer conformance
with AISC and adding the factor Ubs (Reference 17). Examples illustrating the use of Ubs may be found in
Reference 17.

Steel Structures
9.1.5.10 FIELD CONNECTIONS (1994) R(2014)

Welding under field conditions cannot always be satisfactorily performed and inspected to ensure the high quality needed for
welds in railroad structures. Fasteners are therefore required for all main stress carrying connections made under field
conditions.
9.1.5.11 DEVELOPMENT OF FILLERS (2015)
a. Chapter 15 only permits slip-resistant joints. For slip-resistant joints, loose fillers (plates are solely used as packing
pieces) with surface conditions comparable to other joint components are capable of developing the required slip
resistance. Slip-resistant joints do not require additional fasteners when filler plates are used. The fillers become
integral components of the joint, and filler thickness does not significantly affect the joint behavior (Reference 91).
b. Tests on riveted joints have indicated that tight fillers, as described in this sub-article, are desirable when thick filler
plates are needed and long grips result. This requires additional fasteners and they are preferably placed outside the
connection. As an alternative solution, the additional fasteners may be placed in the main splice (Reference 91).
9.1.5.12 COMBINATIONS OF DISSIMILAR T YPES OF CONNECTIONS (1993) R(2014)
b. Welds are more rigid than fasteners. Where used in combination, the welds will be overstressed before the fasteners
become effective.
9.1.5.13 SEALING (1993) R(2014)
b. The requirements of Article 1.5.13b were adopted in 1943 and were based on experience and judgment. The maximum
gage at which a second line of fasteners is considered effective was arbitrarily made the same as the maximum edge
distance (See Article 1.9.3b) recognizing that the maximum gage should increase somewhat with the thickness of the
material.
9.1.5.14 CONNECTIONS OF COMPONENTS OF BUILT-UP MEMBERS (1993) R(2014)
b. The requirements for stitch fasteners in compression members, that the maximum pitch in a single line shall not exceed
12t nor the gage 24t, had been in force for many years prior to 1943 and were considered satisfactory. However, when
it was not practical to have a gage as large as 24t, because the material was not wide enough, or not so disposed as to
permit it, the requirements often led to an extravagant number of fasteners. The 1943 provisions with respect to
staggered pitch permit the use of a reasonable number of fasteners in such cases. A study of the possible fastener
patterns that might result from these provisions indicated that they would give greater security against buckling than
the permissible pattern without stagger, using pitch of 12t and gage of 24t.
9.1.6 MEMBERS STRESSED PRIMARILY IN AXIAL TENSION OR COMPRESSION
9.1.6.1 COMPRESSION MEMBERS (2004) R(2014)
b. The basic formula for determining the minimum permissible thickness of webs and cover plates of compression
members as stated in Article 1.6.1b was derived by Hovey (Reference 80). This basic formula for the determination of
the minimum thickness, t, of plate of width, b, at which buckling of the plate when the plate is simply supported at both
edges and is stressed to the yield point, Fy in compression is:
Fy
t = b ----------------
-
3.616E

Commentary
Hovey then reduced the constant 3.616 by 25% to provide for small initial buckles in the plate as rolled, and the
resulting formula is:
F
t = 0.61b -----y
E
In order to be conservative, the minimum permissible thickness values in these recommended practices have been
F F
established as 0.90b -----y for webs and 0.72b -----y for cover plates.
E E
Where the calculated stress, f, is less than the allowable, Pe, the denominator of the formula determining the
P
permissible minimum thickness may be increased by ---------e with an arbitrary maximum limit of 2 for the value of this
f
radical.
c. For commentary on Article 1.6.1c regarding the minimum thickness of perforated cover plates, see Article 9.1.6.4.
9.1.6.2 OUTSTANDING ELEMENTS IN COMPRESSION (2004) R(2014)
a. The basic formula derived by Bleich for the thickness-width (t/b) ratio at which buckling of the angle leg will occur
when an equal legged angle is stressed to the yield point, Fy , is (Reference 36):
1
Fy F
--t- = - = 1.61 -----y
------------------
b 0.384 E E
For unequal legged angles, for plates supported on one edge, for stems of tees, and for flanges of beams, the Bleich
formula is conservative.
3
In determining the values specified in Article 1.6.2 conservative modifications in the denominator constant have been
made. These modifications were based on experience, judgement and values used currently in other specifications.
9.1.6.4 LACING AND PERFORATED COVER PLATES FOR T ENSION AND COMPRESSION
MEMBERS (2009) R(2014)
9.1.6.4.1 Shear Force

4
The probable maximum shears on column lacing were analyzed by Hardesty (Reference 78). He listed the causes producing
shear on column lacing as follows:
a. Transverse loads acting on the column.
b. Moments at the end of the column, or eccentric application of loads.
c. Initial curvature of the column.
d. The springing of the column as a result of Causes 2 and 3.
e. Local defects in the column and initial stresses set up in the column during fabrication.

Steel Structures
Analyses in 1935 of Causes 2, 3 and 4 by Hardesty (Reference 78) led to the adoption of a column shear formula that remained
in use until 1993. For derivation of this shear formula, the 1935 analyses used the secant formula which resulted in
unnecessarily reduced allowable axial stresses in short columns, as noted in Article 9.1.4. This use of the secant formula led to
unusually high shear forces for short columns. The subsequent abandonment of the secant formula in these recommended
practices (see Article 9.1.4) permits the use of a uniform percentage of axial load for shear load for Causes 2 through 5. The
AFy /150 expression for minimum shear force is included to keep the shear resisting elements from being too light for columns
of length approaching or in the Euler range. Without such a limit, long columns could be designed with very little relative
shear resisting steel since the column area is greatly increased on account of the L/R stress reduction applied for determination
of the column area for axial load, with no corresponding reduction in allowable stress for the shear steel. Furthermore, the
application of the limit to columns in the Euler range makes the shear resisting steel area requirement the same for steels of all
yield strengths, the same as applies to the axial steel area. The limit will not affect columns having customary L/R ratios unless
the yield strength is unusually high.
The formula represents average conditions. For end conditions not properly covered by the assumptions made in the analyses,
special investigation can be made by means of the appropriate formulas given. This design formula covers only shears due to
accidental eccentricities and usual column imperfections, and does not include shears caused by transverse loads (Cause l) or
by eccentricity of load.
Thürliman and White made a study and conducted tests of columns with perforated cover plates which demonstrated that the
formula given for shear on lacing is also adequate for shear on perforated cover plates of structural steels. (Reference 170)
Other specification requirements given for perforated plates are also based on this study and these tests.
9.1.6.4.3 Perforated Cover Plates
d. The formula given in Article 1.6.4.3d for the determination of the thickness of the perforated cover plate is based on
the calculation of the net area required along the center line of perforations to resist the longitudinal shear. Using the
nomenclature of that article, 3V/2ht is the maximum transverse shearing stress at the center line of the cover plate, and
is also the maximum longitudinal shearing stress at that location. The total longitudinal shear in a length equal to the
distance center to center of perforations which must be resisted by the cover plate is (3 V/2ht) ct = 3cV/2h. The net area
of the plate center to center of perforations is (c – a) t; so that the shearing stress, v, on this area is
v = 3cV 1 - or t = --------------------------
3cV -
----------  -----------------
2h  c – a t 2vh  c – a 
The shearing stress in the transverse section through the center of a perforation is usually not critical and can be
calculated according to accepted methods, taking account of all of the section of the member outside of the perforation.
9.1.6.5 EFFECTIVE NET AREA FOR TENSION MEMBERS - STRENGTH (2007) R(2014)
Research (Reference 101) has shown that failure by rupture through a tension member is a function of the effective net section
of the tension member. The effective net section of a member is a function of the geometry of the member and the
connection(s) transferring load into or out of that member.
In connections using fasteners, due to the presence of holes, the concept of effective net section is characterized by rupture
across the net section (Reference 91, and 92). It is important to note that when evaluating a connection using fasteners, the
shear lag reduction coefficient should only be used in conjunction with the net section rupture failure mode (Fa=0.50Fu) and
not the yielding of the net section failure mode (Fa=0.55Fy). In welded connections, due to the absence of holes, the concept
of effective net section is characterized by rupture across the gross section. In welded connections, both the rupture failure
mode and the yielding failure mode occur across the gross section. Again, the shear lag reduction factor should only be used
in conjunction with the allowable stress against rupture (0.50Fu) and not the allowable stress against yielding (0.55Fy).
It should be noted that shear lag is present only when tension is being transferred into or out of a member. Some members,
such as lower chords of trusses, are incrementally loaded across the length of the member. In the case of a lower chord of a

Commentary
truss, the loads are transferred into and out of the chord at panel points. Engineering judgement should be used in applying the
shear lag reduction coefficient in cases such as this. Depending on the details of the connections, the shear lag factor may or
may not need to be applied to that portion of the tension force transferred through the panel point from chord segment to chord
segment. The shear lag factor should be applied to that portion of the tension force transferred from the diagonal members to
the chord segment.
9.1.6.6 EFFECTIVE AREA FOR TENSION MEMBERS - FATIGUE (2007) R(2014)
For the purpose of calculating the stress range in a member, the effective area (gross or net) of a member that receives load
through a connection shall be the sum of the areas (gross or net) of its component parts which receive load directly through the
connection. An example of this would be an I-shaped member that receives load through a connection to gusset plates
fastened only to the flanges and not the web of the member. The effective area (gross or net) of this member, for fatigue
calculations, should be the sum of the areas (gross or net) of its flanges only. If the connection is made through the use of
fasteners in a bearing type connection, the area of the holes shall be deducted in accordance with Article 1.5.8. If the
connection is made through the use of bolts in a slip-resistant connection or through the use of welds, no deduction for holes
shall be made and the gross section of the component parts shall be used.
The purpose of utilizing the effective area (gross or net) for the calculation of a fatigue stress range is to account for the shear
lag effect which occurs when load is transferred through a connection to a member where not all of the component parts of the
member are directly connected. The effective area that is used in the calculation of fatigue stress range is different than that
used in the calculation of stress for a strength evaluation. The shear lag reduction coefficient described in Article 1.6.5 is
based on a level of stress consistent with fracture of the member. Although limited information is available concerning the
magnitude of shear lag at stress levels less than that associated with fracture, researchers agree that the sum of the areas (gross
or net) of the directly connected parts is an appropriate estimate of the effective area that should be used in a fatigue 1
evaluation. (Reference 101, and 102)
9.1.7 MEMBERS STRESSED PRIMARILY IN BENDING
9.1.7.1 PROPORTIONING GIRDERS AND BEAMS (2004) R(2014)
a. These articles provide for proportioning flexural members, whether rolled or built-up, by the moment of inertia
3
method, using a neutral axis along the center of gravity of the gross section, and using the moment of inertia of the
entire net section for the determination of the extreme fiber stress in tension, and the moment of inertia of the entire
gross section for the determination of the extreme fiber stress in compression.
This procedure is not subject to question in the case of welded built-up or of rolled members. In the case of built-up
members using fasteners for construction, the neutral axis for a section taken through fastener holes in the tension area 4
will not be along the neutral axis of the gross section, but will be somewhat nearer the compression flange. If such a
section is analyzed taking account of the lack of symmetry of the section and consequent differences in distances from
the neutral axis to the two flanges in determining the section moduli for the two flanges, the section moduli for the two
flanges will agree very closely with those prescribed in these articles.
b. The requirement that the ratio of the unsupported distance between points of lateral support and the radius of gyration
E
of the compression flange in paragraph b shall not exceed 5.55 ----- is based on the derivation of the parabolic
Fy
formula for the allowable stress in the compression flange as explained in Article 9.1.4. The parabolic formula
becomes tangent to the basic Euler type formula at that point.
9.1.7.2 FLANGE SECTIONS (1994) R(2014)
The 1969 edition of these recommended practices dropped the requirements for riveted and bolted construction which had
appeared in earlier editions specifying relative thicknesses for flange angles and cover plates, and specifying the maximum

Steel Structures
percentage of the total flange area permitted in the cover plates. These requirements had no theoretical basis, but had been
included because of what had historically been considered good practice. Present requirements for the length of partial length
cover plates in riveted and bolted construction control the stress at the end of the cover plate, which is a critical section for
fatigue.
9.1.7.3 T HICKNESS OF WEB PLATES (2004) R(2014)
a. The specified thickness of web plates for flexural members is based on work done by Hovey (Reference 80). Hovey
showed that the buckling of the web of a flexural member on the compression side of the neutral axis can be prevented
either by the use of horizontal (longitudinal) stiffeners or by making the web of such thickness that stability against
buckling is ensured. Vertical (intermediate transverse) stiffeners are not effective in resisting buckling caused by
bending. Assuming the actual extreme fiber stress in the compression flange is 0.55 Fy, and that the compression stress
in the web adjacent to the flange is less than this by an assumed percentage, the ratio of the thickness of the web to the
clear distance between flanges, for a web without horizontal stiffeners, to ensure the stability of the web against
F
flexural buckling may be expressed by the formula 0.18 -----y .
E
Where the extreme fiber stress in compression is less than the allowable, then the ratio may be modified as specified.
b. For web plates stiffened by a horizontal (longitudinal) stiffener located at 0.20 of the web depth from the compression
flange, work by Rockey and Leggett (Reference 124) has shown that, to ensure the stability of the web against flexural
buckling, the web plate thickness required is only 43% of that required without a horizontal stiffener. It is specified that
the web plate thickness shall not be less than 1/2 that determined for a web plate without a horizontal stiffener.
9.1.7.4 FLANGE-TO-WEB CONNECTION OF PLATE GIRDERS (2009) R(2014)
b. The recommendation to use continuous, full penetration groove welds for the flange-to-web connection of open and
non-composite, non-ballasted decks is to prevent fatigue cracking in the top flange welds from the direct application of
cyclic wheel loads. Full penetration groove welds ensure complete fusion between flange and web. Fillet welds may
crack through their throat from the transfer of repetitive concentrated loads unless they have been designed for that
loading condition.
For ballasted, welded steel plate or composite concrete decks, and through plate girders, either continuous, partial joint
or complete joint penetration groove welds or fillet welds can be used as the concentrated loads are distributed such
that the loading on the flange and web plate is not as critical for fatigue. Any of these types of connections will provide
comparable performance. When webs less than 1/2 inch thick are used, fillet welded connections are preferable as they
result in less web out-of-plane distortion from weld shrinkage. Fillet welds on the bottom flange should never be
subjected to concentrated wheel loads.
Fatigue cracking has been known to originate from the lack-of-fusion plane which exists between the web and flange
joints where fillet welds were used in open and non-composite, non-ballasted decks. The driving force producing the
cracking is from the direct application of wheel loads which apply vertical cyclic stresses to the welded joint
perpendicular to the lack-of-fusion plane. Although the applied vertical cyclic stress ranges are compressive, the welds
actually undergo cyclic tensile stress ranges due to the residual tensile stresses from welding. Previous editions of the
manual required the flange-to-web joints be made using continuous, complete joint penetration (CJP) welds. Partial
joint penetration (PJP) groove welds were acceptable if permitted by the engineer, though no guidance on how to
design the PJP joint for fatigue was provided in manual. As a result, CJP were typically specified.
The provisions to include PJP groove welds at the flange-to-web weld in Article 1.7.4b were added to allow the use of
the more economical weld details in lieu of the CJP welds previously required. When properly proportioned, PJP
groove welds will perform as well as CJP groove welds to resist such cracking if sufficient penetration is achieved
and/or if sufficient fillet reinforcement is provided. In such cases, the fatigue strength of the PJP weld subjected to

Commentary
transverse or vertical loads is controlled by weld toe cracking and not throat cracking. Hence, the performance will be
the same as the CJP but with the lower cost PJP.
This can be achieved by using the equation in the provisions of Section 5.4 in Table 15-1-9. The costs to fabricate CJP
are greater than PJP connections due to the increased requirements on inspection and increased overall fabrication
costs.
If a PJP weld is used for the flange-to-web connection, the connection shall be considered a Fatigue Detail Category B’
for checking the longitudinal bending stresses in the girder.
9.1.7.8 WEB PLATE STIFFENERS (INTERMEDIATE TRANSVERSE AND LONGITUDINAL)

(2012) R(2014)
a. Hovey showed that the ratio of web clear depth to thickness for which stiffeners are not needed is determined by the
4.83E
formula -------------- , where Fys is the yield point in shear of the web material (Reference 80). With Fys = 0.636 Fy , the
F ys
E E
formula became 2.75 ----- . The formula 2.12 ----- , used in Article 1.7.8, makes allowance for lack of flatness in the
Fy Fy
web plate.
Where stiffeners are required, their spacing is dependent on the web thickness and the shearing stress in the web. The
development of the formula is based on work by Moisseiff and Leinhard and is based on a factor of safety of 1.5
against buckling of the web (Reference 100). This factor of safety is lower than the basic factor of safety generally 1
used throughout these recommended practices, but is considered adequate because elastic buckling of the web does not
cause failure. When elastic buckling of the web occurs, its share of additional diagonal compression is transferred to
the flange and vertical (intermediate transverse) stiffeners.
The 96 inches maximum spacing of the stiffeners is specified in order to provide stiffeners at reasonably close intervals
so as to aid in eliminating the effect of any small out of flatness that may exist in the web. The 96 inch maximum
spacing is based on work done by Basler indicating that for fabrication, handling and erection purposes the maximum 3
stiffener spacing should not exceed 260t, where t is the web thickness in inches (Reference 34). The distance between
vertical (intermediate transverse) stiffeners shall not exceed the distance between the flanges (web depth) because the
E
formula for stiffener spacing, 1.95t --- , is developed from the theory of elastic stability with this assumption (i.e.
S
critical buckling coefficient in shear always less than 9.35).
The equation for minimum required moment of inertia “I” of the transverse stiffener is a modification of that developed 4
by Bleich (Reference 36). In editions of the Manual prior to 2005 the term “da”, the actual clear distance between
intermediate transverse stiffeners, was used in the formula for “I” instead of the present “d”. The effect of using “da”
was that the stiffeners were sized to develop the elastic shear buckling capacity of the web for that “da”. For various
reasons, transverse stiffeners are sometimes spaced closer than the spacing required by Article 1.7.8a. This close
spacing results in a web shear buckling capacity (based on elastic behavior) much greater than the web shear yielding
capacity. Hence, for such cases of arbitrary stiffener spacing, the stiffener sizes computed from the formula for “I”
were excessive. The use of “d” in the equation for “I” results in more consistent stiffener sizes for girders having
stiffener spacings dependent on other factors in addition to shear.
When Bleich’s work was incorporated into the Manual the limits were omitted. Under certain circumstances where
stiffeners are required, and the shear stress is low, the value of “I” may be negative for values D/d < 1. The equation is
valid only for D/d between 1 and 5, so these limits were added to the Manual in 2012. It is possible that the ratio of D/d
given the other constraints in the article will result in a value less than 1, in which case the limitation yields a minimum
stiffener.

Steel Structures
E
f. The web plate depth criteria of  4.18 --- t relates to the web plate depth required to preclude flexural elastic buckling
 f
of the web plate without horizontal (longitudinal) stiffeners.
g. The recommended practice for placing the centerline of the longitudinal stiffener at 1/5 the web depth from the
compression flange is from work by Rockey and Legett (Reference 124) showing this to be near optimum location to
resist flexural buckling of a simply supported plate. Where longitudinal and intermediate transverse stiffeners
intersect, the preferred detail is to interrupt the transverse stiffener since terminations in the longitudinal stiffener create
details that are more prone to fatigue. Article 9.1.10.2g provides additional guidance and preferred detailing practices
for intersecting stiffeners.
h. The recommended practice for longitudinal stiffener size is taken as a reasonable upper bound for girders of practical
proportions based on the work by Dubas (Reference 48).
i. The recommended practice for the thickness requirements for longitudinal stiffeners is based on the local buckling
behavior of the stem of a tee section.
9.1.7.9 COMPOSITE STEEL AND CONCRETE SPANS (2008) R(2014)
The two types of shear connectors included, i.e. manually welded channels and automatically welded studs, are those most
commonly used (Reference 138). Other types may be satisfactory.
Recommended practice requirements are generally based on performance, allowing the manufacturer and fabricator
considerable leeway as to details and procedures.
9.1.7.9.2 Basic Design Assumptions (1986) R(2005)
and
9.1.7.9.3.1 Design Force for Shear Connectors
The calculations for the value of the horizontal shear between the steel beam and the concrete slab in Article 1.7.9.2 j involve
the determination of the values Sm of the maximum horizontal shear and the value of Sr of the range of the horizontal shear.
The effect of repeated stress variations was studied at Lehigh University by making fatigue tests on composite spans
(Reference 138). The results indicated that the range of horizontal shear resulting from live load and impact load, rather than
the maximum horizontal shear from dead load combined with live load and impact load, controls fatigue capacity. The
allowable design load per shear connector, based on maximum range as specified in Article 1.7.9.3.1, is therefore less than is
specified in that same article for maximum shear.
It is noted that the fatigue check Article 1.7.9.3.1(c) is different from that required in Articles 1.7.9.3.1(a) and 1.7.9.3.1(b).
The checks in Articles 1.7.9.3.1(a) and 1.7.9.3.1(b) are to ensure that fatigue cracking in the weld used to attach the shear
connector to the flange does not occur through the weld throat due to cyclic shear stresses.
The requirement to check fatigue in the base metal of the member to which the shear connectors are attached (Article 1.3.13)
is to ensure toe cracking does not develop at the weld and lead to cracking of the member due to primary bending stress range.
The shear connectors are considered a short attachment on the flange. It is unlikely this check will ever control, as live load
stress ranges will be very small due to the high location of the neutral axis.

Commentary
9.1.8 FLOOR MEMBERS AND FLOORBEAM HANGERS
9.1.8.3 END CONNECTIONS OF FLOOR MEMBERS 1993 R(2014)
The requirements for the connection angles of stringers were developed by Wilson after a study of the bending stresses in such
angles resulting from the lengthening of the bottom chords of through truss bridges under live and impact loads, and from the
deflection of the stringers themselves under such loadings (Reference 171, 176).
Although the flexural stresses in the stringer and connection angles resulting from the lengthening of the bottom chord of
through truss bridges are small, making these connection angles more flexible reduces the rather large bending stresses in the
floorbeams resulting from bottom chord elongation.
The flexural stress in the top portion of the leg of the stringer connection angles connected to the floorbeam may be high as a
result of the deflection of the stringer under load and in the case of thick angles may cause fatigue cracks. For a given
deflection in the top portion of the angle, the stress induced in the angle leg varies directly with the angle thickness, and
inversely as the square of the gage. This deflection is essentially proportional to the length of the stringer. These three factors
have been combined empirically in the requirements of this article.
9.1.9 CONSTRUCTION WITH FASTENERS
9.1.9.3 EDGE DISTANCE OF FASTENERS (2005) R(2011)
a. Based on a review of various textbooks, specifications, and design guides it has been found that there is not a solid
“engineering” reason for the various limits on edge distance. The limits which have been set are more related to
1
detailing issues. The rationale for various edge conditions is as follows:
• Sheared edges
From the Cyclopedia of Civil Engineering, Vol 3, Steel Construction, 1920, a reason for a limit on edge distance is
given based on the tendency of the material to bulge between the edge of the hole and the edge of the plate due to the 3
punching process. To prevent this, it is stated that a minimum edge distance is required. The limit is decreased for
smaller size rolled shapes only to allow punching in the material rather than for any engineering reason. The third
edition of the AISC LRFD Manual of Steel Construction (2003) indicates similar reasoning in the Commentary
contained in Chapter J of that publication. (References 16, and 30)
• Rolled edges of plates, shapes, bars or thermally cut edges

4
Limitations on the clearance for making the hole, as well as the fact that the rolled and thermally cut edges have a
much better tolerance in terms of workmanship, justifies a lower limit for these type edges. The limit is reduced for
flanges of beams and channels to 1.25 times the diameter of the fastener, but only because there may be clearance
issues with the equipment used to make the hole in these sections with smaller flanges. Similar reductions are also
permitted by the AISC Manual. (Reference 16)
• Fatigue
The difference in the edge distance requirements is not related to fatigue. Although a sheared edge is almost always
a lower quality than a rolled or thermally cut edge, the fatigue strength of a plate that is sheared cannot be improved
by slightly increasing the edge distance. The micro cracks produced by the process are present regardless and under
large enough cyclic stresses will grow into fatigue cracks. For a base metal condition to apply at the gross section of
the element (i.e., Category A), there are specific surface quality standards that must be met as per AREMA
Table 15-1-9, AASHTO LRFD, and others. Rolled edges will typically meet these criteria. For thermally cut edges,
it may be necessary to grind the surface of the cut to meet the required surface quality. However, sheared edges will

Steel Structures
not meet the requirements due to the destructive nature of shearing the plate material, and some surface preparation
will be required. (References 4, and 91)
9.1.9.5 FASTENERS IN INDIRECT SPLICES (2017)
For bolted connections, slip-resistant joints are utilized in all connections.
9.1.10 WELDED CONSTRUCTION
9.1.10.1 T RANSITION OF THICKNESS OR WIDTHS IN WELDED BUTT JOINTS (2012)

R(2014)
The requirements of Article 1.10.1 are similar to those in AWS D1.5 with additions to cover flexural conditions, and relocation
of the weld to be outside the end of transition on the radius of the wider plate.
9.1.10.2 PROHIBITED T YPES OF JOINTS AND WELDS (2016)
Because of fatigue considerations, several types of joints and welds are added to types prohibited by AWS D1.5.
g. Highly Constrained Joints:
Welded structures are to be detailed to avoid conditions that create highly constrained joints and crack-like geometric
discontinuities that are susceptible to Constraint-Induced-Fracture (CIF). Avoid intersecting welds by using a preferred
detail (see Figure 15-9-5) or by using high-strength bolted connections. This article is not intended to apply to the
intersection of:
• Welded flange splices with flange-to-web welds
• Welded web splices with flange-to-web welds
• Welded web splices at longitudinal stiffener-to-web welds
Constraint-Induced-Fracture is a form of brittle fracture that can occur without any perceptible fatigue crack growth
and more importantly, without any apparent warning. This type of failure was documented during the Hoan Bridge
failure investigation (Reference 174) as well as in other bridges that have exhibited very similar fractures (References
44, 45 & 49). Criteria have been developed to identify and retrofit bridges susceptible to this failure mode (References
44 & 98).
Although it is common to start and stop an attached element parallel to primary stress (e.g., gusset plate or longitudinal
stiffener) when intersecting a full-depth transverse member, the detail is more resistant to fracture and fatigue if the
attachment parallel to the primary stresses is continuous and the transverse connection is discontinuous. (See
Figure 15-9-5 and Figure 15-9-6)
High-strength bolted connections are not susceptible to Constraint-Induced-Fracture and should be considered where
practical and economical.

Commentary
Net Stress State

Recommendation
Description at Location of Illustrative Example
for Use
Intersection
Use when vertical

attachments can be placed
Tension,
on opposite side of Highly
Compression, or
longitudinal attachments to Preferred
Reversal
the web (i.e., stiffeners or
gusset plates)
Use when vertical

attachments must be placed
on the same side of the web
as longitudinal attachments Tension,
to the web (i.e., stiffeners or Compression, or Acceptable
gusset plates) Reversal
Note the longitudinal 1

attachment is continuous.
Use at bearing stiffeners

Compression Only
Preferred at 3
Note the vertical bearing Bearing Stiffeners
stiffener is continuous
4
Use in zones where net
stresses due to DL and LL Strongly
result in compression stress Compression Only Discouraged at
at location where the Bearing Stiffeners
components intersect.
Figure 15-9-5. Examples of details at intersection of longitudinal attachments and vertical attachments
welded to web

Steel Structures
Net Stress State

Recommendation Illustrative Example
Description at Location of
for Use (Lateral members not shown for clarity)
Intersection
Use where vertical

attachments intersect with
welded gusset plates
Tension,
Note the vertical attachment is Compression, or Preferred
discontinuous Reversal
(see table 15-1-9 for details on

fatigue)
Use where bearing stiffeners

intersect with welded gusset
plates Preferred at
Compression Only
Bearing Stiffeners
Note the vertical bearing
stiffener is continuous
Figure 15-9-6. Examples of welded details at certain locations at the intersection of vertical attachments
and gusset plates welded to web
9.1.10.4 WELDED ATTACHMENTS (2004) R(2014)
The requirements of Article 1.10.4 are based on fatigue considerations.
9.1.11 BRACING
9.1.11.2 LATERAL BRACING (2012) R(2014)
c. In 2012, the statement that “concrete decks are not to be used in through spans unless the deck is isolated from the
flexural stresses of the main girders or trusses” was added because Article 1.11.2a could be interpreted as a general
endorsement of the use of concrete slabs in all steel spans. An inexperienced engineer might not recognize
unintentional composite action which may occur between the slab and the main girders or trusses that has been known
to lead to failure of concrete decks in through spans.
Concrete slabs are generally much stiffer than the steel girders or trusses that support them and may attract a significant
portion of the flexural stresses. Concrete slabs work well in simple deck plate girder (DPG) or deck truss (DT) spans
because flexing from live loading of the main girders/trusses will cycle the slab through compressive stresses that are
safely within the capacity of the concrete. Composite design techniques intentionally transfer compressive loading to
the slab, reducing the quantity of steel required in the main girders or trusses. In through spans, the concrete deck slab
is below the neutral axis of the girder/truss and can attract live load tensile stresses which may exceed the modulus of
rupture of the concrete, resulting in full depth cracks through the deck slab which open and close with each live load

Commentary
cycle of the main girders/trusses. Reference 156 documents strain gauging a TPG span to investigate the breakup of its
reinforced concrete deck. A subsequent paper (Reference 139) includes design details of a concrete deck slab to
provide lateral bracing for a TPG span while isolating it from the flexural stresses of the main girders.
9.1.11.4 CROSS FRAMES AND DIAPHRAGMS FOR DECK SPANS (1994) R(2014)
a. Paragraph a provides the means to accomplish the lateral distribution specified in Article 1.3.4.2.4 (Reference 104 and
168).
Out of plane bending may result from restraint provided by cross frames or diaphragms where there is differential
deflection of adjacent beams or girders. This may be expected to occur in spans with curved alignment, skews or
multiple tracks and has also been observed in single track spans, without skew, on tangent alignment. Out-of-plane
bending may cause high stresses in non-stiffened web gaps, unless rigid type connections are provided to stabilize
these gaps.
b. Determination of whether a cross frame or diaphragm should be used is covered by paragraph b.
c. Requirements for diaphragms are specified in paragraph c to assure suitable lateral distribution of live load.
d-g. Paragraph d, paragraph e, paragraph f and paragraph g concern the spacing of cross frames and diaphragms for various
types of deck construction. Spacing of 18 feet for cross frames and diaphragms in open deck construction has been
specified since 1920; has been found to be satisfactory; and is used as a guide in specifying the spacing of these
members for spans where steel plate, timber or precast concrete decking is utilized in ballasted deck construction and
no top lateral bracing is used, as well as for spans with poured in place decking. The lack of lateral bracing requires 1
close spacing of these members, whereas poured in place concrete decking will allow greater spacing, as evidenced by
tests conducted at the University of Illinois on diaphragms for highway deck spans (Reference 104 and 168).
h. The diaphragms required in paragraph h are primarily for tying the transverse beams together and to some extent for
distributing loads longitudinally.
9.1.12 PINS AND PIN-CONNECTED MEMBERS 3
9.1.12.1 PINS (2015)
a. Normalizing and tempering is the preferred method although heat treatment by annealing, or normalizing alone, may
be acceptable in some cases.
4
b. The hole may be drilled before or after treatment.
c. Historically, recessed pin nuts were available in cast steel for pins up to 24-inches in diameter. Currently, the
availability of pin nuts for this use is limited to a maximum of ten-inch diameter pins. For pins over ten inches in
diameter, the preferred practice is for the pin to be held in place by a recessed cap at each end and secured by a fastener
passing through the caps and through the bored hole in the pin. (References 12, 93)
9.1.13 CONTINUOUS AND CANTILEVER STEEL STRUCTURES
9.1.13.6 UPLIFT (2015)
Uplift should be avoided whenever possible, however when there is a potential for uplift, adequate anchorage must be
provided. Anchorage includes anchor bolts, anchorage details on the span, and substructure self-weight. If there is insufficient
mass in the substructure consideration may be given to providing a counterweight mass or rock anchors to bedrock.

Steel Structures
9.1.13.8 LONGITUDINAL STIFFENERS (2010) R(2014)
A continuously welded stiffener is best for design and performance, but at the intersection of two stiffeners continuity of one
must be sacrificed. In such cases, it is generally better to interrupt the transverse stiffener since terminations in the
longitudinal stiffener create details that are more prone to fatigue. For continuous or cantilever spans, however, the
intersection between longitudinal stiffeners and transverse bearing stiffeners are an exception since bearing stiffener continuity
is necessary for bearing loads and fatigue demand on the longitudinal stiffener termination detail is lower at bearing locations.
Article 9.1.10.2g provides additional guidance and preferred detailing practices for intersecting stiffeners.
9.1.14 FRACTURE CRITICAL MEMBERS
9.1.14.1 SCOPE (2015)
Section 1.14, Fracture Critical Members is used as an extension of and supplement to the current requirements for materials,
fabrication, welding, inspection and testing throughout Chapter 15, Steel Structures and the AWS Specifications. The
implementation of the AWS D1.5 Fracture Control Plan for Fracture Critical Members helps to ensure that a steel bridge with
critical tension components will serve a useful and serviceable life over the period intended in the original design. Some
bridges do not have fracture critical members. However, it is most important to recognize them when they do exist. The
Fracture Control Plan should not be used indiscriminately by designers to circumvent good engineering practice.
The following commentary applies to the provisions of D1.5 Clause 12 FCP as applied to railroad bridges:
Repair Welding [AWS D1.5]
Repair welding consists of deposition of additional weld metal to correct a surface condition, such as insufficient throat or
undercut, or procedures which require removal of weld or base metal preparatory to correcting defects in materials or
workmanship. The latter are divided into noncritical and critical repairs as determined by type and size of defect.
Documentation of critical repair welding is required. This is to enable these areas to be given special attention when
inspections are made after the bridge is in service.
Because virtually all weld repairs are made under conditions of high restraint, the minimum preheat/interpass temperatures
requirements are generally higher than specified for the original welding.
9.1.14.2 DEFINITIONS (2013)
a. Fracture Critical Members (FCM) are defined as those tension members or tension components of members whose
failure would be expected to result in collapse of the bridge or inability of the bridge to perform its design function.
The identification of such components must, of necessity, be the responsibility of the bridge designer since virtually all
bridges are inherently complex and the categorization of every bridge and every bridge member is impossible.
However, to fall within the fracture critical category, the component must be in tension. Further, a fracture critical
member may be either a complete bridge member or it may be a part of a bridge member.
b. Some examples of critical complete bridge members are girders of two-girder bridges and tension chords in truss
bridges, provided a failure would cause loss of serviceability of the bridge. Some bridges do not depend on any single
member, be it in tension or in compression, for structural integrity. Critical tension components of structures usually
occur in flexural members. The tension flange of a flexural member is a critical component if a failure of the specific
flexural member would cause loss of serviceability of the bridge. The web of a flexural member, adjacent to the tension
flange, can be a critical component. Bearing sole plates welded to the tension flange are exempted because they are
located in regions of low tensile stress. By extension, bearing components welded to the sole plate are likewise
exempted from the requirements for FCMs.
Members or member components whose failure would not cause the bridge to be unserviceable are not considered
fracture critical. Compression members and member components in compression may, in themselves, be critical but do

Commentary
not come under the provisions of this Plan. Compression components do not fail by crack formation and extension but
rather by yielding or buckling. Similarly, built-up members made up of mechanically-fastened components, even
though in tension, may not come under the provisions of this Plan. The Plan provides for additional quality of material
and provides for increased care in the fabrication and use of the materials to lessen the probability of fracture of tension
components from crack formation and extension under static and fatigue loading.
9.1.14.3 DESIGN AND REVIEW RESPONSIBILITIES (1997) R(2017)
a. A critical part of any complete Fracture Control Plan must deal with design and detailing. These two sections are not
addressed in Section 1.14, Fracture Critical Members or in this commentary primarily because they are already
included in other parts of this chapter. Fatigue requirements are extensively covered in Article 1.3.13 and, where
necessary, are made more conservative for fracture critical members (see Article 1.3.13i). Fatigue categories for
various bridge details also are extensively covered. However, it remains a prime responsibility of the designer to
examine each detail in the bridge for compliance with the fatigue requirements and to ensure that the detailing will
allow effective joining techniques and non-destructive testing of all welded joints. It is emphasized that the Fracture
Control Plan must begin with the designer and that without proper design, details and specifications, the Plan will fail.
b. The designer is the only one with sufficient knowledge of the design to determine if fracture critical members are
present and to specifically delineate those members or member components. It is, therefore, his responsibility to
designate on the plans those members or member components which are fracture critical. Further, he also is responsible
for the review of the shop drawings to determine whether the plans and specifications have been properly interpreted
and that the fracture critical members are identified and properly fabricated.
9.1.14.5 NOTCH TOUGHNESS OF STEEL IN FRACTURE CRITICAL MEMBERS (2006) 1

R(2008)
For comments relating to Table 15-1-15 see Article 9.1.2.1.
The notch toughness requirements for steels in railroad bridges are similar to those used in steel highway bridges as specified
by AASHTO. The requirements developed by AASHTO were adopted after considerable research and deliberation between
representatives of the AASHTO Subcommittees on Bridges and on Materials, the Federal Highway Administration, the
3
American Iron and Steel Institute, the American Institute of Steel Construction and various consultants. These requirements
were based on numerous technical considerations that include the following:
a. An understanding of the effects of constraint and temperature on the fracture toughness behavior of steels that were
established by testing fracture mechanics specimens.
4
b. An understanding of the effects of rate of structural loading on the fracture toughness behavior of structural steels.
c. The development of a correlation between impact fracture toughness values (KId obtained by testing fracture toughness
type specimens under impact loading) and impact energy absorption for Charpy V-notch (CVN) impact specimens.
d. Specification of CVN impact toughness values that ensure elastic-plastic initiation behavior for fracture of fatigue
cracked specimens subjected to minimum operating temperatures and maximum in-service rates of loading.
e. A verification of the selected toughness values by the testing of fabricated bridge girders that were subjected to the
maximum design fatigue life, followed by testing at the minimum operating temperature and the maximum in-service
rate of loading.
f. An awareness of the extensive satisfactory service experience with steels in bridges and an understanding of the factors
that have occasionally led to brittle fractures in bridges.

Steel Structures
The safety and reliability of steel bridges are governed by material properties, design, fabrication, inspection, erection and
usage. Both the AASHTO and AREMA Fracture Control Plans recognize that attention to all of these factors is essential and
that excessive attention to any single item will not necessarily overcome the effects of a deficiency in any other item.
Neither the AASHTO nor the AREMA fracture toughness requirements are sufficient to prevent brittle fracture propagation
under certain possible combinations of poor design, fabrication or loading conditions. To accomplish that fact would require
intermediate or upper-shelf dynamic toughness levels (also called crack arrest) and these levels of fracture-toughness are not
needed to ensure the safety and reliability of the steel bridges.
The general difference in initiation and propagation behavior as related to fracture toughness test results is shown
schematically in Figure 15-9-7. The curve labeled “static” refers to the fracture toughness obtained in a KId test under
conditions of slow loading. (The curve for intermediate loading rate tests, which are extremely complex to run, would be
shifted slightly to the right of the static curve. The AASHTO material toughness requirements were developed using an
intermediate loading rate found applicable to actual bridge structures.) The impact curve is from KId or other dynamic test
under conditions of impact loading. The difference between these two is the temperature shift, which is a function of yield
strength for structural steels. In an actual structure loaded at temperature A, initiation may be static and propagation dynamic.
However, there is no apparent difference between the two because both initiation and propagation are by cleavage. If a similar
structure is loaded slowly to failure at temperature B, there will be some localized shear and a reasonable level of static
fracture toughness at the initiation of failure. However, for rate sensitive materials, such as structural steels used in bridges,
once the crack has initiated, the notch toughness is characterized by the dynamic toughness level on the impact curve and the
fracture appearance for the majority of the fracture surface is cleavage. If the structure is loaded slowly to fracture initiation at
temperature C, the initiation characteristics will be full shear initiation with a high level of plane stress, crack toughness Kc.
However, the fracture surface of the running crack may still be predominately cleavage, but with some amount of shear as
shown in the lower impact curve at temperature C in Figure 15-9-7.
Figure 15-9-7. Schematic Showing Relation Between Static and Dynamic Fracture Toughness
The use of impact or dynamic fracture tests in fracture control would predict no difference in actual resistance to fracture
between temperatures A and B and only a modest difference between B and C. In fact, however, there is a considerable
increase in resistance to fracture initiation between A and B and between B and C, as is indicated by slow loading tests such as
KIc or crack opening displacement tests. However, there is essentially no difference in the resistance to fracture propagation

Commentary
(i.e., crack arrest behavior) between A and B, and the difference between B and C is modest. Thus, to prevent brittle fracture
propagation in a structure by using material toughness alone (i.e., without proper control of design, fabrication, inspection and
usage), the impact toughness must be quite high, e.g., approaching full shear propagation behavior temperature D. Even then,
there may be situations where crack growth still occurs.
In summary, application of the AREMA material toughness requirements should provide a high level of elastic-plastic or
plastic initiation behavior for steels with fatigue cracks loaded to maximum in-service rates of loading at the minimum service
temperature. Because the AREMA Fracture Control Plan addresses all aspects that may lead to brittle fractures or fatigue
failure (i.e., material properties, design, fabrication, inspection, erection and usage), these material toughness requirements
should be satisfactory in the context of the total AREMA Fracture Control Plan.
The prime focus on Fracture Critical Members must be on quality of the material and fabrication. Using low fatigue resistant
details should be avoided. Category E and E’ details shall not be used on Fracture Critical Members, and Category D details
shall be discouraged and used only with caution.
PART 3 FABRICATION
9.3.1 GENERAL
9.3.1.1 FABRICATOR CERTIFICATION (2016)
Supplemental requirements to the AISC Certification Program for Steel Bridge Fabricators include Intermediate, Advanced,
and Fracture-Critical. Examples of Intermediate and Advanced bridges are given in the certification standard (AISC 205), and 1
the fracture-critical requirements are required by the certification standard for all fabricators producing fracture-critical
members.
9.3.1.6 THERMAL CUTTING, COPES, AND ACCESS HOLES (2009)
There are a number of thermal cutting techniques that are suitable for steel railroad bridge fabrication including, primarily, 3
oxyfuel cutting and plasma cutting. Plasma cutting is generally preferred by fabricators because it is faster and offers
improved quality (such as squarer corners, less hardness variability, and less distortion), but it is limited to thinner sections.
The thickness limit depends upon the equipment, but typically about two inches is the reasonable thickness limit for plasma
cutting. The provisions of this paragraph are similar to those in the AASHTO/AWS D1.5 Bridge Welding Code.
9.3.1.8 PLANING SHEARED EDGES (1994) R(2008)

4
Any sheared edge may have incipient cracks resulting from the shearing operation, which literally tears the material apart.
Since such cracks might be harmful, the requirements for edge planing of sheared material have been included in these
recommended practices and other specifications for many years.
The planing requirements need not be applied to thin A 36 material because the shearing operation does not produce
structurally damaging defects therein.
9.3.1.18 BENT PLATES (2000) R(2008)
In fabrication, plates are often bent to a radius in a press brake or die. When conducted at room temperature, these processes
are known as “cold bending”. To avoid cracking the plate during bending, it is necessary to adopt a suitable minimum inside
bend radius, which typically varies with plate thickness and grade. Over the years, many new grades of steel have come into
existence. A concern in the steel industry was that current limits dealing with this subject may not have been developed on a
consistent basis. As a result, the American Iron and Steel Institute (AISI) initiated a project to develop rational limits for cold
bending plates.

Steel Structures
Initially, AISI retained Concurrent Technologies Corporation (CTC) to conduct an experimental research program, augmented
by inelastic analysis, to investigate the forming characteristics of five plate steels. At the conclusion of that effort in January
1997, AISI then retained R. L. Brockenbrough & Associates to extend the CTC findings to all steel plate specifications
referenced in ASTM A6. That work was accomplished and reported in the document “Fabrication Guidelines for Cold
Bending” dated June 29, 1998.
ASTM A6 has adopted the recommendations of this work as well as some supplemental workmanship language for achieving
quality bent plates. Article 3.1.18 Bent Plates, is derived from both the “Fabrication Guidelines for Cold Bending” and the
ASTM A6 document.
9.3.1.19 PIECE MARKING (2014)
Use of die stamps for steel marking has been customary since early days of fabrication. “Low stress” die stamps have been
desired and specified since the advent of fatigue awareness, but “low stress” has not been defined. To facilitate the low-stress
condition, die stamp marks should not be too deep but need to be deep enough such that the marks are readily legible under
typical paint systems.
There is no defined radius for a "low stress" die stamp, but accomplishing marks with stamps that have a radius instead of a
sharp point is suitable. Examples of stamps that are considered to be low-stress include dot, vibration, and rounded V stamps.
It is known that surface imperfections compromise fatigue performance of the otherwise smooth plate or rolled section.
However, experience demonstrates that die stamp marks are innocuous for steels in Table 15-1-15, particularly when
precautions are taken to ensure the marks are not sharp. Requirements for fracture critical members are more restrictive in
order to be more conservative, but such members also need effective marking for material control, so marking locations should
be chosen to minimize the effect on the member performance.
Modern computer controlled stamping equipment has demonstrated the capacity to provide markings with fatigue detail
categories no worse than Category B. Where the fatigue resistance for a particular marking technique has been documented
through the results of independent laboratory testing and verification, the technique may be used on non-fracture critical
members, and also on fracture critical members provided that the mark is in a location approved by the Engineer, so as to not
affect the fatigue performance of the structural member.
Acceptable locations are intended to not affect the fatigue performance of a member. Since nearly all members have a
connection detail such as a weld or bolt connection that is Category B or lower, placement of a second Category B detail near
this connection will not affect the fatigue resistance of the member. Placement of the piece-mark should be near a connection
which includes but is not limited to connections such as a web-to-flange weld, a bolted connection, or a stiffener at its welded
or bolted connection to the web or flange. Additionally, these marks should not be so close to existing details as to impart
additional residual stresses or sharp transitions, thus the requirement to maintain a minimum distance from a discontinuity.
Examples are provided to aid the fabricator in establishing acceptable locations for gusseted connections (Figure 15-9-8) and
girder splices (Figure 15-9-9). The figures are provided for guidance only.

Commentary
Shaded Areas
for Piece
Mark End of
Member and Gusset
Figure 15-9-8. Marking Location on Truss Connection
Mark End of Splice Plates and Flange,

Marking Inside of Flange Similar
1
Mark End of
Splice Plate and
Web
Figure 15-9-9. Marking Location on Girder Splice 4
9.3.2 BOLTED CONSTRUCTION
9.3.2.1 HIGH-STRENGTH BOLTS, NUTS, AND WASHERS (2005) R(2008)

AND
9.3.2.2 INSTALLATION OF HIGH-STRENGTH BOLTS (2005) R(2008)
High-strength bolts can be adequately installed by methods that control either the bolt deformation (strain) or the applied
torque. Previous editions of this recommended practice adopted turn-of-nut as the primary installation method because
deformation control procedures are typically more reliable and consistent than control by torque measurement. In addition to
turn-of-nut, the current provisions also permit the use of direct-tension indicators (strain-control), tension-control twist-off
bolts (torque-control) and the calibrated wrench (torque-control) methods of high-strength bolt installation. Further details on
these methods can be found in the Specification of the Research Council on Structural Connections (Reference 33 and 130).

Steel Structures
9.3.2.4 SIZE AND WORKMANSHIP OF HOLES (2017)
a. The increase in hole size for bolts greater than or equal to 1 in. in diameter eliminates the need to field ream holes to fit
large-diameter hot forged bolts, which often have a longitudinal forging seam that interferes with holes 1/16 in. larger
than the bolt diameter.
9.3.2.5 PREPARATION OF HOLES FOR SHOP-INSTALLED FASTENERS (2017)
Holes in members subject to live load stress are required to be drilled or reamed in order to avoid the incipient peripheral
cracking at holes punched full size through thick material and the resultant lower resistance against fatigue failures.
9.3.2.6 PREPARATION OF HOLES FOR FIELD-INSTALLED FASTENERS (2017)
The same comments as for Article 3.2.5 regarding cracks in the periphery of punched holes apply also to holes for field
connections, and, in addition, there must be provision for accurate alignment of field connection holes. This article calls for
field connection holes to be so located that they will register exactly when the structure is in its geometric configuration. This
requires that truss members, as erected under a no stress (or practically so) condition, must be bent and forced to fit the end
connections, thus introducing an initial reverse secondary stress which will theoretically disappear when the structure assumes
the loading for which it is cambered.
9.3.2.13 T ESTING AND DOCUMENTATION OF HIGH-STRENGTH FASTENERS (2017)
9.3.2.13.1 Testing
d. Assemblies:
The purpose of the rotational capacity test is to verify that the bolt, nut, and washer assembly will effectively develop
the desired clamping force when it is tightened in the structure, and has adequate lubrication, installation ductility, and
resistance to stripping. With the exception of galvanized ASTM F3125 Grade A325 bolts, the ASTM standards that
define manufacture and testing requirements only address mandatory testing of the individual fastener components.
The only test that evaluates the bolt, nut and washer assembly is the rotational capacity test. This manual requires
rotational capacity testing for all Grade A325 and A490 bolts to assure the performance of the fastener assembly. A
rotational capacity lot is established so that components manufactured with like characteristics that have been tested as
an assembly are controlled for use as bolt, nut, and washer assemblies in the structure.
In general, the required nut rotation for the rotational capacity test is about twice the rotation required to pretension the
bolts using the turn-of-nut method. The RCSC “Specifications for Structural Joints Using High-Strength Bolts” does
not have a required turn-of-nut installation rotation for bolts longer than 12 times the diameter. The required rotation
for installation must be determined by actual tests using a suitable tension device simulating actual conditions
(Reference 122 and Table 15-3-2 Note 2). Since the rotational capacity test rotation is a function of the turn-of-nut
installation rotation, ASTM exempts bolt assemblies longer than 12 times the diameter because there is not an
established standard for these bolts. The rotational capacity test is applicable only to ASTM F3125 Grade A325 and
A490 bolts, and therefore is not applicable to “twist-off” (Grade F1852 and F2280) bolts.
9.3.3 WELDED CONSTRUCTION
9.3.3.1 GENERAL (2015)
a. It is the intent of this article that a fabricator that has properly completed welding procedure qualification tests in
accordance with AWS D1.5 not be required to repeat the tests for individual railroads unless the railroad has made it a
Contract requirement prior to bidding.

Commentary
d. It is common practice to separate flange splices from web splices for plate girders. Additionally, other welds need to be
placed at a distance away from the splice to facilitate flange weld inspection.
e. Welded connections and other welds need to be placed at a distance away from the splice to facilitate weld inspection.
9.3.3.3 FLANGE-TO-WEB WELDS OF FLEXURAL MEMBERS (1995) R(2008)
Only properly selected machine welding is considered to be acceptable for flange-to-web weld of flexural members. In order
to make such welds having the necessary uniformity and quality by any other method, elaborate and costly inspection
procedures would be required.
9.3.3.4 TACK WELDS (1995) R(2008)
The requirements of this Article are based on fatigue considerations.
9.3.5 INSPECTION
9.3.5.5 INSPECTION – WELDED WORK (2002) R(2008)
The requirements of paragraph b and paragraph c take into account the generally high shear to moment ratio in railway
flexural members and the common circumstance of heavy concentrated direct loading of flanges.
9.3.6 SHIPMENT AND PAY WEIGHT 1

9.3.6.3 PAY WEIGHT (2009)
Editions of the AREA Manual prior to that of 1969 permitted payment for pound-price contracts to be based on either scale
weight or computed weight. Consequently, it was necessary to specify a method of computing the weight which is compatible
with scale weight. Since it is practically impossible in many cases to determine an accurate scale weight, and since the
procedure of computing a weight compatible with a scale weight serves no practical purpose, the 1969 edition of the AREA 3
Manual adopted the procedure reflected in the then current edition of the American Institute of Steel Construction Manual of
STEEL CONSTRUCTION for computing weight.
PART 4 ERECTION
9.4.6 PLANS (2017) 4

The AISC Code of Standard Practice for Steel Buildings and Bridges (Reference 10) provides guidance in developing
appropriate contract language about the responsibilities of the Company and the Contractor.
PART 5 BEARING DESIGN AND CONSTRUCTION
9.5.1 BEARING DESIGN
9.5.1.2 GENERAL REQUIREMENTS (2010) R(2017)
b. Each bridge and span configuration induces unique loads and movements into the bearings. In turn, each type of
bearing with their varying restraint characteristics return unique forces back into the span and/or into the substructure.
Movements include both translations and rotations. The sources of these movements include bridge skew and
curvature effects, initial camber or curvature, construction loads, misalignment or construction tolerances, settlement
of supports, thermal effects, and displacements due to live load deflections. Skewed bridges can have both

Steel Structures
longitudinal and transverse movements. Curved bridges can have both radial and tangential movements which could
occur at differing angles at each substructure unit. A practical form for tabulating bearing load resistance and
movement requirements is presented in Section 14 of Reference 6 or in Appendix H of Reference 142.
c. Spans, of any length, with similar live load deflection to span length ratios will experience similar angular rotations at
the bearings. Short spans will incur substantially more rotation cycles than longer spans (once per axle, truck, or car
vs. once per train).
Many, but not all, short spans (50 feet or less) have historically performed well using simple flat plate on flat plate
bearings. The practice has succeeded because:
• Short spans, particularly deck girder spans, often have small span to depth ratios, and are much stiffer than required
to meet deflection criteria contained in Article 1.2.5.b., reducing rotations at the bearings.
• Flat plate on flat plate bearings often permit adequate rotation to occur, usually through a combination of edge
bearing of the sole plate on the bed plate, use of elastomeric or malleable pads between the bed plate and the bearing
seat, and sometimes soft bearing seats absorb rotations of the bed plate (for example timber caps and/or blocking
beneath the bearings).
Unless extensive experience in similar circumstances has proven the success of flat plate on flat plate bearings in a
particular design, the designer should verify that rotation is adequately accommodated. Best practices for using flat
plate on flat plate bearings include:
• Keeping sole plates as small as possible, and setting them on larger bed plates so that edge bearing stresses from the
sole plate are reduced and are ideally applied within the middle third of the bed plate.
• Installation of elastomeric or malleable pads between the bed plate and the bridge seats.
• Flexible connection of the sole plate to the span (bolted rather than welded).
9.5.1.5 BEARING SELECTION CRITERIA (2010)
a. Table 15-5-1, Bearing Suitability, the bearing selection criteria, and insights into typical movement accommodation
characteristics of each bearing type delineated herein are a composite of that presented in Reference 6, Section 14;
Reference 125; and Reference 142, Appendix H; with appropriate adjustments made for usual railroad bridge bearing
practice. Specific bearing type suitability characteristics listed in the table and their application to railroad bridges are
based on the following:
(1) Resistance to Vertical Loads: All bearing types listed in Table 15-5-1 were chosen for their suitable resistance to
downward vertical loads, except Plain Elastomeric Pad Bearings which have limited application as discussed
below. However, since uplift can occur at the bearings of railroad bridges due to the highly dynamic effect of
railroad live loading, Article 5.1.2.b requires that hold-down devices be provided at all bearings unless waived by
the Engineer. When calculated uplifts occur, additional, more substantial elements, such as link bars or other
heavy hold-down devices are to be designed and incorporated into the bridge bearings.
(2) Fixed Bearings: Fixed Bearings are intended to restrain translations in all directions while allowing rotation on at
least one axis. Since the primary movements in the typical railroad bridge are generally in the longitudinal direction,
all Fixed Bearings are indicated to be suitable for rotations about a transverse axis. Fixed Bearings that restrain
rotations in all directions are generally not practical for railroad bridge applications, thus are not included in Table 15-
5-1.
(3) Flat Steel Plate on Flat Steel Plate Bearings: Steel-on-steel sliding bearings are common in historical railroad
bridge practice for spans less than 50 feet (15 000 mm) in length particularly when utilizing rolled beams. Thus,

Commentary
because of the span length limits, the usual limit on longitudinal translation is 0.5 inches (12 mm) and the usual
limit on rotation about the transverse axis is 0.01 radians. Steel-on-steel sliding surfaces develop a higher
frictional force than Bronze or Copper Alloy and PTFE sliding surfaces. This friction force acts on the
superstructure, substructure, and bearing and is an important design consideration.
Steel-on-steel sliding bearings are still used in modern railroad bridge practice when an economical bearing type is
desired and the span and substructure can accommodate the loads induced by the higher coefficient of friction
between the steel plates. Beam span lengths of 70 feet (21 000 mm) have been used with steel-on-steel sliding
bearings and with a plain elastomeric pad placed under the masonry plate to accommodate rotations.
(4) Rocker and Roller Bearings: Rocker Plate, Pin and Rocker, and Roller Bearings utilize a cylindrical surface which
is generally aligned on a transverse axis to the bridge to accommodate the primary longitudinal movements found
in most railroad bridges. Because of this, all bearings of these types are listed as suitable for rotations about a
transverse bridge axis and unsuitable for rotations about a longitudinal bridge axis.
To maintain stability of the rocker (prevent rocker tip-over), a 4 inch (100 mm) longitudinal translation limit is
commonly considered appropriate for Pin and Rocker Bearings in the typical railroad bridge designed in
accordance with AREMA Manual for Railway Engineering, Chapter 15 which has a limit of applicability to spans
of 400 feet or less. Larger bearings can be designed to accommodate larger translations from longer spans or long
continuous multi-span units, but the rocker becomes very tall so rocker stability must be specifically addressed in
the bearing design. Special restrainers may be required, particularly in high seismic zones.
(5) Bronze or Copper Alloy and PTFE Sliding Surfaces: Bronze or Copper Alloy and PTFE sliding surfaces are
commonly used as components of bridge bearings to accommodate the sliding and/or rotating movements.
Sliding surfaces develop a frictional force that acts on the superstructure, substructure, and the bearing. Friction,
1
thus, is an important design consideration. PTFE Sliding Surfaces generally will have the lowest friction
coefficient resulting in the transfer of the lowest friction forces into the bridge or its supports.
Flat Bronze or Copper Alloy or PTFE Sliding Surfaces can be designed to accommodate very large translations,
but cannot accommodate rotations by themselves. Other elements, such as pins, rocker plates, curved sliding
surfaces, elastomeric pads, etc. must be added to the complete bearing assembly if rotations are to be 3
accommodated. Restrainers, such as guide bars or other devices are frequently added to limit translations in
certain directions and to provide resistance to loads in those directions.
Cylindrical Bronze or Copper Alloy or PTFE Sliding Surfaces can be designed to accommodate very large
rotations, but only in one direction. Thus, as described above in Item 4 for Rocker Plate and Roller Bearings, the
cylindrical surface is generally aligned with the transverse axis of the bridge, which provides suitable
accommodation of the primary longitudinal rotations while preventing transverse rotations about a longitudinal 4
bridge axis. A cylindrical surface alone thus aligned cannot accommodate longitudinal translations and can only
accommodate limited transverse translations.
Spherical Bronze or Copper Alloy or PTFE Sliding Surfaces can be designed to accommodate large rotations in
any direction; thus, they are classified as Multi-rotational Bearings. A spherical surface alone cannot
accommodate translations in any direction.
Combined flat and curved Bronze or Copper Alloy or PTFE Sliding Surfaces can be utilized to accommodate both
large translations and rotations. When this is required the flat surface should be placed at the bottom of the
Bronze or Copper Alloy or PTFE element with the curved surface at the top.
Double Cylindrical (Bi-radial) Bronze or Copper Alloy or PTFE Sliding Surfaces can, in general, be designed to
accommodate large rotations about any horizontal axis and limited translations in any horizontal direction. Even
though rotation about a vertical axis cannot be accommodated, this type of bearing is still classified as a Multi-
rotational Bearing. A 1 inch (25 mm) limit on translations should be used in the typical railroad bridge bearing
since this type of bearing can become unstable with larger translations in one or more directions, particularly when
combined with larger rotations. While adding a separate flat sliding surface would accommodate larger

Steel Structures
translations, the complex configuration of having three sliding surfaces combined with the restrainers required to
prevent or limit translations along the axis of the cylindrical surfaces renders a very difficult and costly design.
Spherical Bearings combined with a flat sliding surface should be used to accommodate large translations and
rotations in multiple directions.
(6) Plain Elastomeric Pads: The three most important properties of Elastomeric Bearings that distinguish them from
other construction materials are: 1) flexibility in shear relative to their thickness; 2) stiffness in bulk or direct
compression relative to their shear flexibility; and 3) ability to undergo very large shear deformations without
damage relative to their thickness. However, a simple block of elastomer subjected to compression expands
laterally due to the Poisson effect and is much softer than other typical railroad bridge construction materials. If
the lateral expansion occurs freely, the resulting compressive deflection is unacceptable. For railroad bridges, the
total compressive deflection is limited by Article 5.6.3.5.e to 0.125 inches (3 mm) to provide acceptable ride
quality.
Plain Elastomeric Pads rely on friction at their top and bottom surfaces to restrain compressive bulging. Friction is
unreliable, however, and local slip results in a larger elastomer strain. The increased elastomer strain limits the
load capacity of the Plain Elastomeric Pad. The allowable stress depends upon the shape factor of the elastomeric
bearing pad. Plain Elastomeric Pads, therefore, must be relatively thin, which leads to the thickness limits of
Article 5.6.3.8 and Article 5.12.1.c.
Thin elastomeric bearing pads can tolerate only small translations; thus, a small 0.25 inch (6 mm) maximum
translation limit is recommended for Plain Elastomeric Pads used in the typical railroad bridge application. Since
rotation is accommodated in Elastomeric Bearings by an increase in compression on one side of the pad and a
reduction on the other side, thin elastomeric bearing pads can tolerate only small rotations also. This leads to the
recommendation that a small 0.01 radian maximum rotation limit be considered in the design of Plain Elastomeric
Pad bearings for railroad bridges.
(7) Steel Reinforced Elastomeric Bearings: Many of the issues with total elastomer thickness, load capacity,
translation and rotation limits of Plain Elastomeric Pads can be addressed by increasing the number of elastomer
layers by adding thin steel reinforcing plates between the layers. The steel reinforcing plates prevent outward
movement of the elastomer at the interface between the two materials so that lateral expansion can take place only
by bulging. Thinner elastomer layers thus lead to less bulging and higher compression strength and stiffness,
which is desirable, but this also results in high rotational stiffness. Larger rotations can be accommodated by
adding more layers. A bearing that is too stiff in rotation leads to lift-off and high local stresses that could cause
damage. Thus, selection of the number and thickness of the elastomer layers is a compromise between the needs
for compressive stiffness and rotational flexibility.
For railroad bridges, the total thickness of the bearing or thickness and number of individual layers is limited by
the total compressive deflection limit of 0.125 inches (3 mm) as defined in Article 5.6.3.5.e. Since
accommodating rotation is an important part of railroad bridge bearing design, being able to utilize the full
available rotation limit of 0.04 radians as listed for highway bridges in Reference 125 is typically considered
important. This recommended maximum rotation limit, however, when considered in relation to the compressive
deflection limit for railroad bridges, dictates a reduction in the maximum translation available for railroad bridges
compared to highway bridges. Thus, it is recommended that a translation limit of 2 inches (50 mm) be considered
unless other accommodation, such as a separate flat sliding surface, is provided.
(8) Disc Bearings: In a Disc Bearing, compressive load is carried by a hard elastomeric (polyether urethane) disc. As
with all elastomeric type bearings, rotations are accommodated by an increase or decrease in compressive
deformations on opposite sides of the disc. The hard elastomer used in Disc Bearings is not flexible in shear so it
cannot accommodate horizontal translations without the addition of a flat sliding surface or other device. To
prevent the disc being overstressed by horizontal loads, a metal pin is placed through a hole in the center of the
disc. Thus, Disc Bearings by themselves are listed in Table 15-5-1 as fixed bearings with no translation allowed.

Commentary
Disc Bearings are classified as a multi-rotational bearing. At low loads, they work like an unreinforced plain
elastomeric pad as described above. The elastomer used in Disc Bearings, however, is very much stiffer than that
used in a typical elastomeric pad. Some slip and some lateral expansion occur. It has been shown that the rotation
and compression stiffness are both related to the square of the shape factor. The shape factor therefore cannot be
too small or the disc would deflect too much under compression, and it cannot be too big, or the bearing would be
too stiff in rotation. The choice of disc dimensions is therefore a compromise between these two design goals.
This need for compromise means that designing for a rotation much larger than 0.02 radians is difficult,
particularly for the typical railroad bridge application. See Reference 142, Appendix H.
The rotation may need to be further limited since, for high rotations under lighter loads, significant uplift can occur
creating potential for damage to the Bronze or Copper Alloy or PTFE Sliding Surfaces that may be used to
accommodate lateral translations. Even for fixed bearings, uplift conditions will cause abrasion of the disc and
raises the possibility of ingress of dirt. See Reference 142, Appendix H.
c. Pot-type bearings are not recommended for support of railroad bridges because of concerns over reduced bearing life
due to large cyclical live load deformations and rotations.
d. As described in Article 9.5.1.5a(6) and Article 9.5.1.5a(7), design of Plain Elastomeric Pads and Steel Reinforced
Elastomeric Bearings is a compromise between the need for compressive stiffness and rotational flexibility. To provide
the minimum rotational flexibility required by typical railroad bridge applications and stay within the rotational limits
recommended in Article 9.5.1.5a(6) and Article 9.5.1.5a(7) without lift-off, further limits on the width of elastomeric
pads or bearings are required.
9.5.2 BASIC ALLOWABLE STRESSES

1
9.5.2.1 STRUCTURAL STEEL, BOLTS AND PINS (2012)
c. The allowable stress in bearing between rockers and rocker pins was adapted from editions of AREMA Manual
Chapter 15, Steel Structures, Section 1.4, prior to the 1969 edition and the low value of 0.375 Fy was retained to
minimize pin wear. Pin wear had historically been a cause of trouble when higher values for this condition were
permitted. 3
The allowable stresses for anchor bolts match the 2005 AISC provisions for Allowable Strength Design (ASD) of
bearing-type connection bolts with threads included in the shear plane and are 1/2 of the 2005 AISC and 2007
AASHTO LRFD nominal capacities. The nominal capacities as listed in the 2005 AISC Steel Construction Manual,
Table J3.2, are divided by the ASD safety factor, Ω, of 2.00.
The allowable stress in bearing on expansion rollers and rockers was based on static and rolling tests on rollers and 4
rockers (Reference 20). The average vertical pressures over calculated contact areas for loads substantially less than
allowable design values are in excess of the yield point, causing a flow of the material. It was concluded that the
resulting “spread” of the roller and base, measured parallel to the axis of the roller at points near the surfaces in contact,
was the most satisfactory phenomenon to use in determining design values. Such “spreads” or deformations were
measured in units of 0.001 per inch per 1,000 strokes, each stroke corresponding to a roller movement of 4 inches and
an equal movement back. Design values according to the tests would give total deformations varying from about 3
units to less than 1.
9.5.2.6 POLYETHER URETHANE DISC BEARINGS (2007) R(2017)
The recommended average allowable compressive stress on polyether urethane discs in Disc Bearings of 5,000 psi matched
AASHTO’s average allowable compressive stress in 2005 when it was recommended by a special Subcommittee 7 Task Force
on Implementation of Higher Allowable Bearing Stresses. Even though there was limited test data in 2005 for disc bearings in
railroad bridges or for disc bearings subjected to high live load to dead load ratios, the Task Force believed that the existing

Steel Structures
data and current testing indicate that an allowable average bearing pressure of 5,000 psi on polyether urethane discs is
conservative for the polymer compound recommended in Article 5.7.2.d.
9.5.3 STEEL BEARING COMPONENT DESIGN
9.5.3.2 MATERIALS (2016)
9.5.3.2.2 Anchor Bolts and Anchor Rods
c. The stainless steel materials in Table 15-5-4 were selected for inclusion based on availability. Availability of specific
materials and sizes shall be verified as part of the selection process. The designer should select material and diameter
based on availability at the time of construction. AISI stainless steel Type 316 has been commonly used in salt water
environments.
Heads are generally added to bolts through an upset process that disturbs the surface resistance of the stainless steel.
Passivation is a process that resets the weather resistant surface of the material. Headed stainless steel bolts should be
passivated, even when embedded, to preclude surface discoloration and corrosion.
9.5.3.3 SHOES AND PEDESTALS (1997) R(2017)
The requirements of Article 5.3.3 provide that the load is uniformly distributed over the entire bearing surface, and that, in the
case of welded bearings, the load is transmitted in bearing.
9.5.4 BRONZE OR COPPER-ALLOY SLIDING EXPANSION BEARING DESIGN
9.5.4.3 DESIGN (2001) R(2017)
b. For design, the static coefficient of friction is specified to be a minimum of 0.10 since it is to be applied when calculating
loads acting on bearing components or the bridge substructure or superstructure due to friction, thermal restraints or the
portion of other horizontal loads transferred through an expansion bearing with bronze or copper-alloy sliding plates. The
provision in Article 5.10.1 that limits the coefficient of friction of the bronze or copper-alloy sliding expansion bearing
plates to a maximum of 0.10 will theoretically assure that Article 5.4.3 produces conservative loads for designing other
elements of the bearing or bridge. At the discretion of the Engineer, when calculating loads acting on other bridge
elements, a higher coefficient of friction, such as 0.25 specified by some railroads, may be used to accommodate the
possibility of future partially frozen bearings.
9.5.5 PTFE SLIDING BEARING SURFACE DESIGN
9.5.5.3 DESIGN (2006) R(2017)
9.5.5.3.1 General
c. For design, the static coefficient of friction is specified to be a minimum, over the range listed, since it is to be applied
when calculating loads acting on bearing components or the bridge substructure or superstructure due to friction,
thermal restraints or the portion of other horizontal loads transferred through an expansion bearing with PTFE sliding
surfaces. The provison in Article 5.11.1(c) that limits the coefficient of friction of the PTFE sliding surfaces to a
maximum of the listed values will theoretically assure that Article 5.5.3.1(b) produces conservative loads for designing
other elements of the bearing or bridge. At the discretion of the Engineer, when calculating loads acting on other
bridge elements, some railroads specify a higher coefficient of friction, such as 0.25, to accommodate future partially
frozen bearings.

Commentary
9.5.7 MULTI-ROTATIONAL BEARING DESIGN
9.5.7.1 SCOPE (2007)
c. Pot type bearings are not recommended for railroad loading due to experiences with seal failures.
9.5.7.3 DESIGN (2014)
9.5.7.3.1 General
h. Differing deflection and rotation characteristics of different types of multi-rotational bearings may result in damage to
the bearings and/or structure.
9.5.7.3.2 Loads
a.
(2) The critical combination may be a larger horizontal load combined with a minimal vertical load occurring
simultaneously.
9.5.10 BRONZE OR COPPER-ALLOY SLIDING EXPANSION BEARING

CONSTRUCTION
1
9.5.10.1 GENERAL (2001) R(2008)
b. See Article 9.5.4.3b.
9.5.11 PTFE SLIDING BEARING SURFACE CONSTRUCTION

3
9.5.11.1 GENERAL (2002) R(2008)
c. See Article 9.5.5.3.1c.
PART 6 MOVABLE BRIDGES 4
FOREWORD
See the Foreword for References for Movable Bridges.
9.6.1 PROPOSALS AND GENERAL REQUIREMENTS
9.6.1.1 GENERAL (1986) R(2010)
The history of movable bridge design specifications can be traced back at least as far as 1901 to the Baltimore & Ohio
Railroad Company specification for swing bridges. C.C. Schneider’s Paper No. 1071 in the June 1908 ASCE Transactions,
Volume LX, Page 258 appears to be the earliest specification giving allowable loads and stresses for individual components.
The basic content of the Schneider specification appears in the first edition (1922) of the AREA Movable Bridge
Specification. Many changes and additions have been made over the years to that specification and this recommended
practice.

Steel Structures
Early movable bridges, designed using the requirements outlined inthe Schneider paper, have proved to be durable. In
contrast, certain proprietary movable bridge designs using less stringent requirements have not been as durable.
It thus appears that the Schneider specification and the succeeding editions of the AREA Movable Bridge Specification have
successfully defined adequate design standards for typical movable bridge machinery.
Nevertheless, failures have occurred in bridges designed to these specifications. Some of these failures may have occurred
because of lack of good engineering judgment in the application of the specifications. Others may have occurred because of
lack of good engineering judgment in using components and/or details not covered in the specifications, as well as errors in
construction, faulty operation and inadequate maintenance.
These recommended practices contain no criteria for the anticipated number of openings expected over the life of the bridge.
Two basic categories of machinery components are covered in the recommended practices.
The first category includes components which always or nearly always operate under maximum design loads. These are the
components which support the dead load of the movable span. Examples of these are counterweight sheaves for vertical-lift
spans, trunnions for bascule spans, treads for rolling lift spans, center pivots, rim bearings and end wedges for swing spans.
The second category includes components whose loading consists of friction, inertia, wind, ice, and other transient loads,
during operation of the movable span.
Machinery in the first category carries maximum or near maximum loads at all times. Machinery in the second category
seldom carries maximum design loads and normally operates at a relatively small fraction of design load.
The basis of the recommended practice is textbook mechanical engineering methods and allowable stresses for the design of
heavy machinery developed prior to 1940.
The bridge machinery design philosophy should be simple and normally not be based on overly-sophisticated methodology for
several reasons. They include:
a. Bridge machinery is heavy, low-speed machinery intended to last 50-100 years.
b. The real loading conditions and number of cycles of operation are difficult to establish.
c. The level of maintenance over the life of the bridge is difficult to establish at the time of design.
d. Future replacement of sophisticated components can be difficult.
9.6.1.8 SIGNALS AND INTERLOCKING (2012)
b. A movable bridge rail joint may be of several different types or styles that provide the transfer of rail traffic from the
fixed spans on and off the movable span. Rail locks are present on some types of movable bridge rail joints. A rail
lock will provide a connection of these rails to each other or to the structure of the bridge. While not common on Class
I railroad properties, there are a number of these devices still in use on movable bridges.
c. There may be preferences desired by the Railroad as to the type, placement and mounting of the detectors. Movable
rail joints can be designed and constructed to provide space and mounting for these devices. The tolerance of ¼” in
Article 6.1.8c is intentionally tighter than the FRA requirement in order to allow the approximately 1/8” of future wear
and/or looseness during times between adjustment and repair of movable bridge rail joints.

Commentary
9.6.2 GENERAL FEATURES OF DESIGN
9.6.2.12 RAIL ENDS (2015)
d. Testing at FAST (Reference 72 and 106) and experience on several railroads has shown that movable bridge rail end
joints which provide a level running surface profile for wheels provide reduced impacts, reduced noise, longer life and
better overall performance compared to those which do not provide a level running surface profile.
Where rail ends are cut square, level running surface profiles are typically achieved by tapering the tops of the running
rails rather than raising the wheels with the alternate running surface. The tapering of the ends of the running rails
prevents rail end batter.
Traffic and train speed, span locks, and bridge conditions need to be considered in the selection of appropriate rail end
bridge joints.
9.6.3 LOADS, FORCES AND STRESSES
9.6.3.11 MACHINERY SUPPORTS (2013)
Excessive deflection of machinery supports may have a detrimental effect on the operation of the machinery. In the past, an
attempt was made to prevent these effects by limiting the depth-to-span ratios of support beams to 1/8, but that standard alone
allows for a wide range of stiffness for the same depth member. In addition, higher yield strengths of steel allow for a less stiff
section for the same load demand. While depth-to-span criteria are useful as a starting point for beam depth, deflection criteria
are clearly necessary to limit the detrimental effects on machinery. 1
9.6.4 BASIC ALLOWABLE STRESSES AND HYDRAULIC PRESSURES
9.6.4.2 MACHINERY PARTS (1993)
Some allowance for stress concentration factors is included in the basic allowable design stresses. Stress concentration factors 3
for unusual configurations are not covered and must be recognized by the designer.
Some counterweight sheave trunnions have failed due to fatigue as the sheaves exceeded 500,000 revolutions. The combined
effects of high-cycle complete reversals, small fillet radii at changes in trunnion diameter and section discontinuities resulting
from termination of grease grooves close to the fillets have produced fatigue cracking in the area of the fillets. Journals with a
length to diameter ratio exceeding 1.2 may result in high bending stresses in the area of the fillets.
4
9.6.4.8 HYDRAULIC SYSTEMS AND COMPONENTS (1984) R(2010)
Consideration should be given in the design of hydraulic systems for the effect of the large inertia of the moving span and the
compressability of the hydraulic fluid.
Provision should be made to contain any hydraulic fluid leakage to avoid contamination of the waterway or surrounding areas.
9.6.5 GENERAL DETAILS
9.6.5.13 LUBRICATION (2008)
j. Due to the variety of lubricants available, this article provides a warning to Designers, Owners, and Erectors to verify
that all lubricants are compatible. Experience has shown that serious damage can occur when lubricants are
incompatible.

Steel Structures
9.6.5.34 SPECIAL PROVISIONS FOR SWING BRIDGES (2003)
9.6.5.34.1 Center Bearing
Center bearing swing spans are generally preferable to rim bearing swing spans because of simpler fabrication and erection,
and more reliable operation.
9.6.5.34.2 Rim Bearing
See 9.6.5.34.1 Center Bearing.
9.6.5.34.5 End Wedges and Center Wedges
Wedges with sliding surfaces which must operate under load may be designed with steel against bronze to minimize galling.
9.6.5.36 SPECIAL PROVISIONS FOR VERTICAL LIFT BRIDGES (1997)
9.6.5.36.10 Welded Sheaves
b. Welded counterweight sheaves must be designed with special care to assure adequate fatigue life in both the sheaves
and the trunnions.
9.6.5.36.12 Operating Ropes
c. The minimum tension in the slack rope should preferably be not less than 10% of the maximum operating tension and
can be determined by measuring the sag in the rope.
9.6.5.37 HYDRAULIC SYSTEMS
9.6.5.37.4 Safety (2015)
9.6.5.37.4.7 Counterbalancing
Counterbalance valves are modulated to develop pressure to counteract the force of an overhauling external load. Properly
sized and adjusted, counterbalance valves dampen the system to prevent dynamic oscillations and create back pressure to
prevent over-speed caused by overhauling loads, typically those resulting from gravity, wind and/or inertia. Counterbalancing
is required for open loop hydraulic systems. Closed loop hydraulic systems by their nature resist overhauling loads through
regenerative braking and therefore do not require counterbalance valves. Counterbalance valves generate significant heat
which must be accounted for in the design of the hydraulic power unit.
9.6.6 WIRE ROPES AND SOCKETS
9.6.6.2 DIAMETER OF ROPE (2010)
a. The diameter of counterweight ropes had been limited by the specification to 2-1/2 inches since 1922. Wire ropes of
larger diameter are now available for use on vertical lift bridges.
Prior editions of the Manual listed tolerances for rope diameters from 5/8 inch to 2-1/2 inches.

Commentary
9.6.6.3 CONSTRUCTION (2010)
a. Improved plow steel (IPS) was the only grade of wire rope permitted by the Manual since 1922. Advances in wire rope
technology have led to the wide use, in other industries, of the higher strength grade ropes listed in ASTM A1023 and
Federal Specification RR-W-410F. The higher strength rope grades are; extra improved plow steel (EIP) and extra
extra improved plow steel (EEIP). Each step upward in grade represents an increase of approximately 10 percent of
minimum breaking force, compared with the next lower grade. Extra improved plow steel wire rope (EIPS) is now
permitted by the Manual, as well as the improved plow steel rope (IPS) that was formerly required. The specifier is
alerted to the fact that there is no known published data regarding the relative performance of extra improved plow
steel rope compared to improved plow steel rope over many years of service in a bridge application.
The Manual now requires that all wire rope for movable bridges be preformed in accordance with the strong
recommendation of the Wire Rope Technical Board (April 2007).
Prior editions of the Manual required wire rope to be made of bright (uncoated) carbon steel wires. The use of rope
made with drawn-galvanized or drawn-zinc mischmetal (Zn5/Al-MM) wire is now permitted. However, wires coated
with zinc or Zn5/Al-MM by hot-dipping are not permitted because the hot-dipping process relieves some of the
residual stresses in the wire from prior cold drawing, thereby reducing the strength of the wire.
Drawn-galvanized and drawn-(Zn5/Al-MM) ropes are used in other industries where long life under adverse
environmental conditions is required. Even if the zinc layer is “partly damaged”, the steel remains protected as the
electro-chemical process results in the zinc corroding first. Zinc is more resistant to wear than Zn5/Al-MM. For these
reasons, drawn-galvanized ropes are preferred over bright (uncoated).
b. Only one classification of counterweight wire rope had been permitted by the specification since 1922. It is 6x19 with 1
a fiber core. Since 1938, or earlier, only the subclass 6x25FW has been specified. This construction has generally
given acceptable service when the ropes are draped over counterweight sheaves with sheave diameter (D) to rope
diameter (d) ratios of approximately 80. The 6x25FW characteristics have been found to be an acceptable
compromise between flexibility and wear resistance for counterweight ropes of 2-1/2 inch diameter or less. For
situations where wear resistance is of importance, the subclass 6x26WS (Warrington Seale) is considered
advantageous. However, for specific situations consultation with a wire rope manufacturer may disclose that other
subclasses of 6x19 are more suitable. 3
Vertical lift bridges have recently been designed with larger diameter ropes and smaller D/d ratios. For these
situations 6x25FW and 6x26WS may be too stiff because the diameter of the wires in the outer layer are larger for the
larger diameter ropes. Hence, the Manual now permits the use of Class 6x36 and Class 6x61 wire rope, which have
greater flexibility. Only one subclass of 6x36 rope is manufactured for most rope sizes.
Counterweight ropes may be stationary for long periods under the design tension. They are subjected to lateral 4
compression along the length draped over the sheaves. These forces tend to deform the rope from a circular cross
section. The deformation is resisted by the core, which supports the strands in the radial direction. Fiber cores
deteriorate if the ropes are not properly maintained and lose their effectiveness as strand supports. Independent wire
rope cores (IWRC) do not deteriorate in the same way as fiber cores and are considered better supports for the strand.
However, IWRC are generally stiffer than fiber cores and there is metal-to-metal contact between the strands and the
core.
In addition, the Manual now permits the use of compacted strand (CS) for wire ropes. The strands of these ropes are
subjected to mechanical work after they have been closed. The mechanical work changes the cross sectional shape of
the outer wires, thereby increasing the contact area between wires and increasing the external metallic surface area.
These ropes also have more metallic cross sectional area than ropes with only circular wires of the same diameter and
class. CS ropes are stronger than circular wire ropes of the same class, grade, and diameter and have much greater
resistance to wear and fatigue.

Steel Structures
9.6.6.7 WIRE – PHYSICAL PROPERTIES (2010)
The prior versions of the Manual covered one grade and subclass of wire rope and listed the required properties of the wire and
described wire tests. Because of the expansion to other grades and classes, reference is now made to ASTM A1023 and
ASTM A1007, which are cross-referenced by Federal Specification RR-W-410F.
9.6.6.8 ULTIMATE STRENGTH (2010)
Prior versions of the Manual listed the required ultimate strengths of the 6x19 IPS ropes with fiber cores. Because of the
expansion to other grades and classes of rope, reference is made to ASTM A1023 for ropes 2-3/8 inch diameter or less. For
rope sizes larger than those listed in ASTM A1023, the designer is referred to Federal Specification RR-W-410F. Although
the term “ultimate strength” has been retained in the heading of this article, and elsewhere in Section 6.6, it should be noted
that the synonym in ASTM is “minimum breaking force” and in Federal Specification RR-W-410F “Minimum Breaking
Strength (force)”.
A new requirement is that rope tests to destruction be conducted per ASTM A931 Test Methods for Tension Tests of Wire
Rope and Strand, in the presence of an inspector designated by the Engineer.
9.6.9 ERECTION
9.6.9.1 ERECTION OF MACHINERY (1996) R(2002)
Bridge machinery erection generally should be started with alignment of the lower speed components and working back to the
prime mover. This gives the best flexibility to correct misalignments.
9.6.9.4 LUBRICATION (2008) R(2010)
c. Due to the variety of lubricants available, this article provides a warning to Designers, Owners, and Erectors to verify
that all lubricants are compatible. Experience has shown that serious damage can occur when lubricants are
incompatible.
9.6.9.7 COUNTERWEIGHTS (1983) R(2003)
b. For satisfactory balance, the movable span should have a slight closing force present when seated and either a neutral
or very slight opening force present when fully open. Balance can be checked in the field by the following procedures:
(1) Compare motor currents during opening and closing of the span.
(2) Compare power meter (kw) readings during opening and closing of the span.
(3) Run a drift test in the mid range of travel in both the opening and closing direction. Compare the drift in each
direction with power off and the brakes released.
(4) Measure the torque in the drive train during opening and closing of the bridge.
(5) Compare the grease patterns on the main pinion teeth.
(6) For vertical lift bridges, weigh the imbalance between the span and the counterweights.
The above tests should be run under minimum wind velocity and with equal speed in the opening and closing direction.
Periodic retesting of the balance of the movable span can reveal changes in operational characteristics.

Commentary
PART 7 EXISTING BRIDGES
9.7.2 INSPECTION
9.7.2.1 GENERAL (2010)
The Bridge Inspection Handbook (Reference 18) provides additional information on the inspection of steel railway bridges.
9.7.3 RATING
9.7.3.1 GENERAL (2015)
c. Details of designs that do not meet the criteria in this Chapter may fail much sooner than expected and may require
evaluation beyond the scope of this recommended practice.
In cases where defects have been found as a result of inconsistent details, the Engineer should monitor the results of
inspections for any changes in number or length of such defects. The frequency of inspection might need to be
increased. Repair or retrofit work might need to be undertaken.
9.7.3.1.1 Normal Rating (2011) R (2017)
a. The intent of the Normal Rating is to limit the stresses in the structure to those for which it would be designed given the
yield strength of the steel in question and the design recommendations of Part 1, Design. Limiting the loading at or 1
below the Normal Rating will prolong the useful life of the structure, providing the intended design factor of safety. For
older structures which were generally not designed for current fatigue criteria, rather than reduce the rating by
requiring use of current fatigue allowables, a remaining fatigue service life calculation may be made. It is then up to the
Engineer to consider the trade-off between the resulting higher Normal Rating and the consequent reduced remaining
fatigue service life.
The correct section for tension yielding has always been the gross section. Nevertheless, it was the practice to use the 3
net section prior to 2006 to introduce an additional factor of safety and provide consistency with certain test practices
particularly in the area of fatigue.
For structures designed starting in 2006, an additional requirement of checking the effective net section against the
ultimate tensile stress was introduced to cover a concern with High Performance Steels (HPS). With the re-
introduction of this criterion (dropped many years ago) it is now possible to be consistent with the actual behavior of
structures in checking tension yielding on the gross section and ultimate tensile strength on the effective net section. 4
Traditionally, bridge structures that have been designed and rated in accordance with AREA and AREMA procedures
have used yielding of the net section for tension calculations. Given that there are thousands of bridges already rated
based on these assumptions and with a desire that there not be a sudden change in the calculated rating of railroad
bridges, the Committee felt that yielding of the net section for tension calculations may continue to be used for
structures designed before 2006. This will tend to give conservative results for traditional steels. Ratings should
indicate the rating criteria used (e.g. AREMA Rating 2005) so as to clarify comparisons between ratings based on
different methodologies.
The allowable rating stresses, when wind forces are included, can be increased to 25% greater than basic allowable
stresses, but in no case greater than the allowable stresses for Maximum Rating. The 25% increase is included so that,
for members such as truss chords where wind forces may be significant, the Normal Rating will not be less than the
loading for which the member was designed.

Steel Structures
9.7.3.1.2 Maximum Rating (2014)
a. Maximum Rating recognizes that loads producing stresses higher than allowable design values may be imposed on a
structure. To maintain an adequate factor of safety and to reduce the effects of fatigue, it is recommended that loads
exceeding the Normal Rating be allowed only infrequently.
Infrequent loading is defined as a small number of relevant loads such that fatigue is not a concern.
Maximum Rating is primarily intended for infrequent clearance moves, such as heavy transformers, loads on depressed
center flat cars, Schnabel cars, and other special equipment. Often these loads are carried in a dedicated train and they
are usually accompanied by special handling instructions.
b. Paragraph b permits the Engineer to authorize more frequent loads up to Maximum Rating with the understanding that
the remaining useful life of the structure may be reduced. If frequent loads exceeding Normal Rating are contemplated,
it is appropriate that either a more detailed inspection be made of fracture critical members or a fatigue analysis be
conducted per Article 7.3.3.2 and Article 9.7.3.3.2 to predict the remaining useful life of the structure and preclude the
continued application of loads beyond the stage where the potential for member failure is high. Another alternative is
to predict the theoretical remaining useful life and when this predicted life has expired, continue using the structure by
making more detailed inspections of fracture critical members.
9.7.3.2 LOADS AND FORCES (2016)
9.7.3.2.2 Live Load (2016)
b.
Figure 15-9-10. General Dimensions for Rating of Specific Equipment
c. Table 15-9-3 may be used in the rating evaluation process when a particular maximum gross weight and minimum car
length-over-pulling-faces is determined for specific routes. The vehicle dimensions, configuration, axle loadings and
train consist for evaluation of actual traffic on the route should be as directed by the Engineer.

Commentary
Table 15-9-3. Weight and Axle Spacing of AAR Standard Freight Cars
Minimum Minimum Axle Axle Axle

Gross Maximum
Axles Length Over Truck Spacing Spacing Spacing
Rail Load Axle Load
(n) Pulling Faces Centers Inboard Outboard Truck
(lb) (P)
(Lo) (TC) (SI) (So) (ST)
263,000 4 65,750 lbs 41 ft 10.5 in 27 ft 4 in 21 ft 6 in 8 ft 8.5 in 5 ft 10 in
Reference: AAR Manual Standards and Recommended Practices (Reference 32)
9.7.3.2.7 Bracing Between Compression Members (2002) R(2017)
b. A lateral bracing force of 1.25% of the total axial force is based on an initial out of straightness of L/500 plus a total
load displacement of L/900 or equivalent combination. These two, when combined, are approximately L/320.
For other cases of greater deviation from the straight, the following formula may be used:
Lateral Bracing Percentage = 400 (Initial Deviation + Total Maximum Deformation Under Load)/L.
1
9.7.3.2.8 Longitudinal Force R(2017)
a. Longitudinal forces due to train traffic on railway bridges are influenced by a number of factors including:
(1) the type of motive power used

3
(2) train tonnage
(3) grades and curves
(4) type of braking system
(5) likelihood of starting or stopping a train at or near a particular bridge 4

(6) individual railroad operating practices.
For further information, see AREMA Manual for Railway Engineering, Chapter 16 Economics of Railway
Engineering and Operations, and the commentary section on design for longitudinal forces (9.1.3.12). (References 67,
116, 145, 148, and 161)
The longitudinal force in Article 1.3.12 is based on E-80 loading. For structures with a live load rating different from
E-80, the longitudinal force used in rating is to be reduced or increased by the ratio of the rating for live load to E-80.
b. The longitudinal force due to locomotive traction is to be used in locations where train operations may include either of
the following conditions:
(1) Application of maximum tractive effort below 25 mph (40 km/h)
(2) Application of maximum dynamic braking effort with actual train speed less than 25 mph (40 km/h)

Steel Structures
c. In other locations, the longitudinal force due to locomotive traction may be reduced in proportion to the larger of the
actual locomotive tractive effort or the dynamic braking effort. The actual locomotive tractive effort or dynamic
braking effort used at a location can be obtained either from actual train operations, or estimated using the methods in
AREMA Manual for Railway Engineering, Chapter 16 Economics of Railway Engineering and Operations. The
maximum tractive effort and dynamic braking effort ratings of locomotives are typically listed in the operating
timetable, or may be obtained from the operating department of the Railroad.
f. The recommended practice also covers the extreme events of emergency braking, or starting a train from a stationary
position at maximum tractive effort, at locations where other longitudinal forces are expected to be low, with an
allowance of 1.5 times the allowable stresses for rating. For a Maximum Rating calculation, this will allow stresses
that exceed the yield point for this rare extreme event.
In the event that longitudinal forces are higher than the calculated capacity of the structure, operating restrictions for
the bridge need to be discussed with operating and mechanical personnel.
It is important to trace the load path that these forces will follow to the point at which they are taken out of the structure
and ensure that the load path is consistent with the compatibility of deflections and rotations.
9.7.3.3 STRESSES (2017)
9.7.3.3.1 Computation of Stresses R(2017)
b. The provisions for intermediate stiffener spacing in Article 1.7.8 are derived from the equations for elastic and inelastic
buckling of a flat web under shear stress, using suitable reduction factors. See Article 9.1.7.8. Those equations are
critical load solutions for thin flat plates based on small deflection theory and do not consider post-buckling conditions
in the web plate. The detailed analysis referred to in Article 7.3.3.1b is a more refined elastic/inelastic critical load
analysis of a flat plate subjected to shear and bending (Reference 70 and 131). The Engineer is advised to apply a
reduction factor to the computed critical load to account for web plate out-of-flatness and other imperfections.
These comments do not consider the effect of stiffeners to support the top flange.
c. It has been common practice not to rate gusset plates under the assumption they would rate as strong as the main
members. This sub-article was added to identify the gusset plates that need to be evaluated. Clearly under-designed
gusset plates and any other components that are under-designed relative to the rest of the structure should be evaluated.
Refer also to Article 9.1.5.4.
9.7.3.3.2 Fatigue
a. The intent of evaluating a structure for fatigue in this Article is to minimize the probability of failure as a result of
fatigue crack growth. This primarily affects the maximum service life for which the structure is designed. If a reduced
life is acceptable, higher loads are permissible providing the serviceability is not impaired throughout the shortened
useful life.
There are two ways to deal with fatigue. The first is to ensure that a structure is fail safe, and the second is to limit the
usable life to one that is shown to be safe for a certain period of use (Safe Life). Given that most details in older bridges
were developed before fatigue became a major concern, only a few structures could possibly be regarded as fail safe.
Even these are only fail safe within a risk management framework.
Where waivers of the need to make a safe life calculation are permitted in the Manual, it is felt that these structures are
at least as safe as the general level of safety provided by typical civil engineering structures.
In the more common case for railroad bridges, that of ensuring a safe usable life, there are again two major alternatives
to consider. One can limit the load to a very low value and obtain a long usable fatigue safe life, or one can use a higher

Commentary
load (usually the Normal Rating) but accept a reduced usable fatigue safe life. The latter alternative has been chosen
for this Manual, based on economics.
A multi-step method has been selected which is designed to first screen the overall bridge population for bridges with
details with potential fatigue problems, followed by more sophisticated evaluation methods where the Engineer deems
them to be needed and appropriate. The intent is to avoid detailed fatigue calculations when experience has shown that
a class of structures clearly has adequate fatigue life. In this case, calculating a fatigue rating is not appropriate,
although an estimation of remaining fatigue life may be needed.
The fatigue strengths used throughout this section are the latest available and are based on the results of full scale
testing on relevant bridge-sized components. The failure criterion used, where a safe life must be estimated, is that of a
2.5% probability of failure of a component based on simple calculations.
b. For lines carrying low volumes of traffic, fatigue is generally not a problem. In Article 7.3.3.2b for a bridge carrying
less than 5 million gross tons per annum throughout its existing and projected life, a fatigue check is waived for usual
mixed traffic. The term “usual mixed traffic” refers to normal North American equipment and is intended to exclude
solid unit train traffic and unusual heavy loads such as heavy molten metal cars or heavy transformers in frequent
service.
A special analysis is needed for non-freight traffic to establish the appropriate parameters for the relevant number of
cycles.
c. The first step in an evaluation of any detail is to check the detail against the design requirements for the Normal Rating
stress ranges. This Article also provides guidance on details not fully covered in the design section of this Manual and
for wrought iron riveted connections. 1
The purpose of Article 7.3.3.2c is to flag members with less than ideal fatigue strength. The various parts of
Article 7.3.3.2c, except Article 7.3.3.2c(4), are intended as a preliminary screening tool to see if there is a potential
fatigue problem. Article 7.3.3.2c(4) is intended to eliminate the need for calculation where adequate inspection within
the limited circumstances mentioned should find a fatigue crack before it becomes critical (a very necessary part of this
Fail Safe assumption).
3
Any structure that meets the requirements of Article 7.3.3.2c, except Article 7.3.3.2c(4), is deemed adequate from a
fatigue standpoint because it has met a very stringent criterion, that of calculated infinite, or at least very long, life.
This criterion is appropriate for newly designed structures.
(1) On multiple track structures, the incidence of more than one track being loaded frequently by heavy freight loads
is low. This section allows the use of probability evaluations to estimate the occurrence of more than one track
being loaded simultaneously. 4
(2) Welded structures do not have the inherent redundancy of riveted or bolted construction. Hence, the consequences
of fatigue crack growth are more serious for most welded connections and members than for riveted or bolted
structures with built-up sections. Severe corrosion may reduce the advantage of redundancy in riveted or bolted
members.
Experience with welded highway bridges that have experienced fatigue cracking has demonstrated that the
members usually fail before the crack is discovered (References 64, and 65). As a result, it appears prudent to use
the requirements of Article 1.3.13 when rating welded bridge members. High-strength bolted joints provide
improved fatigue resistance.
(3) The fatigue resistance of members with riveted or other mechanically fastened connections with low slip
resistance is defined by Detail Category D as a result of review of available test data (References 1, 57, 66, 118,
121, and 177). The most recent research indicates a variable amplitude stress range fatigue limit of 6 ksi, extending
to at least 100 million cycles (Reference 177). Referring to Figure 15-9-11, it is apparent that nearly all test data on

Steel Structures
riveted joints with normal levels of clamping force fall to the right of the line defined by Detail Category C
between 6 ksi and 9 ksi. The existing test data (References 57, 66, 118 and 177) show failures at high numbers of
cycles below the constant amplitude stress range fatigue limit for Detail Category C, 10 ksi, but above the variable
amplitude stress range fatigue limit value of 6 ksi. Hence, any evaluation using Detail Category C must extend on
to 6 ksi. For stress ranges above 9 ksi, the test results for riveted connections typical for railroad bridges fall to the
right of the line defining Detail Category D.
For riveted bridge components For optional evaluation of drilled or reamed bridge
For design only: components
N=2.183 x 109 Sr-3 Sr > 9 ksi (see 9. 7.3.3.2 Fatigue)
For evaluation: N=2.183 x 109 Sr-3 Sr > 9 ksi
N=2.183 x 109 Sr-3 Sr > 9 ksi N=4.446 x 109 Sr-3 9 ksi > Sr > 7.65 ksi
N=4.446 x 109 Sr-3 9 ksi > Sr > 6 ksi N=2.465 x 1015 Sr-9.5 7.65 ksi > Sr > 6 ksi
Fatigue limit: (Sr)fl = 6ksi Fatigue limit: (Sr)fl = 6 ksi
Figure 15-9-11. Riveted Bridge Components
It is reasonable to permit a higher fatigue stress range for Root-Mean-Cube (RMC) stress ranges below 9 ksi if the
connection or member in question has tight riveted joints. Where the rivets are tight and rivet holes are smooth,
having been correctly drilled or subpunched and reamed, a further refinement in the allowable stress range is
permissible. A line on the rivet S-N plot extending from Detail Category C at 7.65 ksi to 6 ksi at 100 million cycles
may be used in lieu of the horizontal line at 6 ksi (Reference 1 and 177). This discretion has been left to the
Engineer dependent on his verifying the tightness of the rivets or bolts and the adequacy of the clamping force.
This refinement does not apply to punched holes.
For riveted construction where the members are fabricated from multiple elements, the immediate consequences
of fatigue cracking may not be as serious as in welded structures. Riveted construction often has built-up members

Commentary
and connections, so that if one element fails there is normally sufficient capacity and redundancy for the force to
be redistributed. The members will usually survive long enough for the crack to be detected by routine inspection
thereby permitting corrective action before more serious damage develops. If no immediate repair action is to be
taken, the probable time between first detectable cracking and uncontrolled propagation should be taken into
account when setting up inspection frequency. Where the constant amplitude stress range exceeds 9 ksi, test results
indicate that not much time elapses between easily detectable cracking and member failure.
(4) Article 7.3.3.2c(4) permits waiver of the fatigue provisions when the Engineer can show that the structure has an
adequate level of redundancy, so that should cracking develop it can be accommodated. The requirement that
sufficient lateral resistance be provided by bracing or diaphragms to ensure that existing crack tips will not be
subjected to unaccounted secondary stresses is consistent with test results (Reference 66).
(5) Wrought iron riveted connections exhibit a fatigue strength represented by Detail Category D with a variable
amplitude stress range fatigue limit of 6 ksi (Reference 1 and 177).
(6) For eyebars and pin plates, the critical section is at the pin hole normal to the applied load. Several studies have
indicated that the stress concentration factor at such a location is in excess of 4 (Reference 57 and 151). Detail
Category E is intended to provide a conservative estimate of fatigue resistance at such connections. Particular
attention should be given to any forge seams or other unusual flaw-like conditions that may exist at the bore of the
eyebar normal to the applied load.
Suitable analytical and/or experimental studies may show that a lower stress concentration exists if pin fit and the
component geometry are favorable. If the stress concentration factor is less than 4, Detail Category D can be used
to assess fatigue resistance. Detailed analysis or full size testing may be used to demonstrate that an even more
favorable category is appropriate.
1
The inclusion of bending stresses is intended to apply primarily to hangers and similar members where pin
connections may develop large bending stresses due to configuration, corrosion, wear or other causes.
(Reference 31)
For advice on secondary stresses, see Article 1.3.15 and Commentary Article 9.1.3.15.
3
(7) Test results (Reference 66 and 118) indicate that severe corrosion may lead to the initiation of cracks. If the
thickness of a component is reduced by 50% or more, the member at that location is best categorized by Detail
Category E. Until more conclusive test results are available, no advice can be given in this Manual on sections
with less than 50% loss of thickness.
d & e.A structure that does not pass Article 7.3.3.2c may still be adequate, but only if further evaluation demonstrates that
this is the case. There are two generic ways to check this. 4
The first method is to ascertain as accurately as possible the actual damage done by traffic to date and to estimate the
remaining life based on future projected traffic. This requires the records of the operating railroad, if they are available,
and a calculation using the concepts outlined in Article 7.3.3.2d, e and f with a full spectrum rain flow analysis of
actual tested trains crossing the bridge, or a short cut method using the AAR bridge fatigue charts as an approximation.
When the actual stress cycles can be estimated from known traffic, the total variable stress cycles can be estimated and
the effective stress range calculated by the formula given in Article 7.3.3.2e.
The resulting coordinates can be compared with Figure 15-9-11 and Article 1.3.13 for the applicable fatigue detail.
The values of α for various spans and member classifications are tabulated in Table 15-9-1. The factor i is the ratio of
the number of occurrences of SRi to the total number of occurrences of cyclic stress Nv.
The second method is to refine the model of the structure by more sophisticated analytical means, or by field-testing
using the structure itself as the model. In the event that calculated stress ranges give a low estimated remaining safe

Steel Structures
fatigue life, it is suggested, if economically justified, to obtain stress range data by strain gaging under traffic that is at
the upper weight range of traffic expected on the structure. In this instance, if the actual strains are less than the
analytical model strains, either a rechecking of fatigue capacity using Article 7.3.3.2c or a more thorough analysis as
per Article Article 7.3.3.2d, e and f will result in a longer useful calculated fatigue life.
Clearly, in the most pressing case, these methods may be combined, i.e. real traffic data and the most accurate model
possible. Caution must be exercised in the application of these articles in order to avoid erroneous conclusions. For
example, the use of these articles to evaluate a floorbeam or stringer without being cognizant of the effect of potential
end fixity, or the application of these articles to the midsection of such a member while ignoring the real stress
variation at the end connections, could lead to wrong conclusions.
When the procedures above result in a deficient remaining life estimation, several alternatives are available. Depending
on the economics, consider:
Closing the structure or restricting traffic;
Repairing, strengthening or retrofitting the deficient details or replacing the structure;
Initiating frequent and very rigorous inspections, being very cognizant of fracture critical considerations;
Installing strain gages to establish actual stress ranges related to the traffic handled, to permit a more accurate
analysis;
Using more sophisticated techniques, such as acoustic emission verification and fracture mechanics.
RMC only works as a quick way of evaluating (ni/Ni) when the detail’s fatigue behavior is described by one curve
with the stress range elevated to the power -3.
f & g.The limits and the stress ranges outlined in Article 7.3.3.2f and 7.3.3.2g on fatigue details being sufficient to
eliminate the existence of the Constant Amplitude Fatigue Limit (CAFL) Stress Ranges are approximate and are based
upon a small number of tests (Reference 62).
h. This paragraph draws attention to details that have low fatigue capacity with particular reference to Fracture Critical
Members.
9.7.3.3.4 Allowable Stresses for Maximum Rating (2017)
a. Traditionally, bridge structures that have been designed and rated in accordance with AREA and AREMA procedures
have used yielding of the net section for tension calculations. Given that there are thousands of bridges already rated
based on these assumptions and with a desire that there not be a sudden change in the calculated rating of railroad
bridges, the Committee felt that yielding of the net section for tension calculations may continue to be used for
structures designed before 2006. This will tend to give conservative results for traditional steels. Ratings should
indicate the rating criteria used (e.g. AREMA Rating 2005) so as to clarify comparisons between ratings based on
different methodologies.
Nevertheless, it is imperative that steels with allowable Maximum Rating stresses based on Fy greater than 0.5 Fu be
evaluated differently, particularly because some High Performance Steels (HPS) have low ultimate to yield ratios.
Applying the ultimate tensile strength to the effective and/or net section and the yield strength to the gross section in
axial tension more correctly represents the behavior at failure. Because test results have been reported on the net
section for some fatigue studies, the fatigue limits recommended in other articles of this Chapter may not be consistent
with this provision.
Since there have been many failures in floorbeam hangers, and since an increase in allowable stress for high-strength
steels in such applications is not acceptable, the allowable stress for such members has been established as that

Commentary
permitted for members of A36 steel, and a greater apparent factor of safety has been adopted, in line with past
experience, for such members.
b. The allowable values represented in Table 15-7-2 for Shear in Rivets are intended to provide Maximum Rating
parameters that cover current and historic rivet steel specifications. As of 2011, the ASTM Specification for Carbon
Steel Rivets is A502 Grade 1. As of 2011, the ASTM Specification for Carbon Manganese Steel Rivets is A502 Grade
2. As of 2011, the ASTM Specification for Weathering Steel Rivets is A502 Grade 3.
9.7.4 REPAIR, STRENGTHENING AND RETROFITTING
9.7.4.2 PLATE GIRDERS OR ROLLED BEAMS (2012)
9.7.4.2.5 Effective Span
Abutment anaylsis should include, but not be limited to, loading conditions, footing pressures, pile loads, seismic response and
minimum edge distance.
9.7.5 MAINTENANCE
9.7.5.2 MAINTENANCE OF STRUCTURAL ELEMENTS (2015)
c. Ultrasonic impact treatment has been shown to alter residual stresses in welds and adjacent base metal and thereby
1
provide improved fatigue resistance to some weld details (Reference 105, 128, 152, 162). This treatment has proven to
be effective in reducing or delaying crack initiation in weld details attaching secondary members to webs of girders or
beam spans.
3
PART 8 MISCELLANEOUS
9.8.1 TURNTABLES
9.8.1.3 BASIC ALLOWABLE STRESSES AND DEFLECTIONS (2008) R(2014)
9.8.1.3.1 Structural Components

4
b. The second diagram in Figure 15-8-2b. consists of two 4-axle diesel locomotives and may be used to apply this article
by changing the 25 foot distance to ensure that all 8 axles are on the turntable.
9.8.2 METHODS OF SHORTENING EYEBARS TO EQUALIZE THE STRESS
9.8.2.1 GENERAL (2017)
There are two different procedures currently being used for shortening eyebars in order to equalize the stress in eyebar sets.
The first method, which utilizes clamp plates, rods and trammels to restrain the heated upset eyebar, was adopted as a
recommended practice in 1948. The second method, which uses no external hardware to restrain the heated eyebar, was first
used circa 2000, but it has been used more frequently since that time.

Steel Structures
9.8.2.2 SHORTENING EYEBARS WITH CLAMP PLATES, RODS AND TRAMMELS (2017)
This recommended practice for shortening eyebars to equalize their stress was originally published in the AREA Manual in
1948 following completion of a 1943 Committee 15 assignment on shortening of eyebars to equalize stress. This procedure
involved heating a short length of the bar, which was restrained between clamps, to 1600 to 1800 degrees Fahrenheit, low in
the range of forging temperatures, and drawing the clamps together to upset and shorten the bar. Since eyebar heads were
formed by forging, these temperatures were considered appropriate. Current practice relies more on restrained thermal
expansion of the heated area to produce upsetting followed by shortening as the bar cools. A detailed report comparing the
effects on strength of various methods used to tighten loose eyebars and recommending the procedure for flame shortening,
which was adopted as a recommended practice by the AREA in 1948, can be found in Reference 25. Fatigue tests were run on
three bars in each condition. A summary of data from these tests is tabulated below. Considering the variability of test results
and limited field experience when compared with shortening steel eyebars, extreme caution should be exercised if the method
is applied to wrought iron eyebars. In heat shortening wrought iron eyebars, there is a possibility of aggravating
delaminations, which may promote fatigue crack propagation.
EYEBAR FATIGUE TEST DATA FROM 1947 REPORT

Wrought Iron Steel
Not Flame Not Flame
Shortened Shortened Shortened Shortened
Max. Life/Mean Life 1.195 1.643 1.134 1.058
Min. Life/Mean Life 0.720 0.358 0.866 .900
Fatigue Strength
at 500,000 cycles 32.7 ksi 30.5 ksi 36.4 ksi 37.2 ksi
Fatigue Strength
at 1,000,000 cycles 28.4 ksi 26.4 ksi 31.5 ksi 32.4 ksi
Since the process has many features in common with heat straightening, a similiar temperature range was considered
appropriate for investigation. Experience indicated that temperatures below 1300 degrees Fahrenheit were not effective.
Since temperatures in the range of interest could be determined with sufficient accurancy using inexpensive temperature
sensing crayons, the recommended temperature range was changed and narrowed to reduce the risk of metallurgical damage.
9.8.2.3 SHORTENING EYEBARS WITH RING HEATING (2017)
Shortening eyebars using ring heating has been effectively used since circa 2000. With this method, rather than using clamp
plates, rods and trammels, the eyebar is heated in a 2-inch wide circumferential line (a ring) around its cross section. The
concentrated heated area is restrained by the adjacent cooler portion of eyebars; upsetting occurs locally in the eyebar and the
eyebar shortens as it cools. Experience shows that a temperature range of 900 degrees to 1,100 degrees Fahrenheit is effective.
A number of heat cycles in different locations on the eyebar may be required to obtain desired results. No eyebar should be
heated more than twice in the same location due to the diminishing effectiveness of the procedure in subsequent heat cycles.
Shortening eyebars using ring heating does not require the attachment of any hardware, such as clamp plates and rods. The
area that is heated is much smaller and the target heating temperature is lower compared to the traditional method of using
clamp plates, rods and trammels. Therefore, the time required to cool the eyebars is shorter.
Shortening eyebars using ring heating may not be as effective as the traditional method when shortening larger eyebars.
Additionally, the effects of this procedure on the fatigue performance of eyebars have not been documented.
9.8.2.4 DETERMINING APPROXIMATE DEAD LOAD STRESS (2017)
The formula provided for determining dead load stress (and the chart developed from this) is based on an exact relation
between tension stress and the fundamental flexural natural frequency of the eyebar about its minor axis, and assumes that the

Commentary
ends of the eyebar are ideal pin connections. When this formula was first introduced, it was recognized that while a much
higher degree of end fixity might be expected in at least some cases, similar exact solutions do not exist for any other end
conditions (Reference 25). A subsequent study (Reference 99) confirmed that actual eyebar behavior can closely emulate
ideal fixed-end conditions, and that use of the pin-ended formula in such situations can greatly overestimate the actual stress,
particularly for eyebars of lower slenderness; error in excess of 100% is not uncommon. Reference 99 provides supplemental
analysis tools that allow for more accurate estimates of dead load stress when essentially fixed-end conditions are apparent at
one or both ends of an eyebar, and considers how the effective eyebar length might be defined in these situations.
9.8.3 ANCHORAGE OF DECKS AND RAILS ON STEEL BRIDGES
9.8.3.1 FOREWORD (2010) R(2017)
a. Starting in 2003, as part of the Association of American Railroads’ Strategic Research Initiatives to reduce the stress
state of railroad bridges, the Transportation Technology Center, Inc. (TTCI) conducted a series of bridge tests,
developed an analytical model, and performed a parametric evaluation to investigate the interaction of continuous
welded rail (CWR) with long open-deck steel bridges (Reference 89).
The results of this investigation indicate that there are conflicting considerations regarding thermal effects of CWR on
long open-deck bridges.
Rail expansion joints (See Article 8.3.4) effectively accommodate rail thermal expansion and contraction; however,
their use generates high impact loads and may accelerate bridge degradation (Reference 2, 3, 72 and 89). Also they are
costly to install and require high maintenance. Without rail expansion joints, longitudinal rail restraint must be
incorporated to reduce gap width and derailment risk due to broken rails. Rail restraint might introduce high rail 1
longitudinal forces into the bridge in case of a broken rail.
Longitudinal restraint also causes longitudinal forces to develop in the rail during span expansion and contraction.
These forces add to the rail force developed from heating and cooling of the CWR. Additional compressive forces in
hot weather might increase the risk of track buckling at bridge approaches, particularly at abutments that support
expansion bearings. Additional tensile forces in cold weather might accelerate rail defect and crack growth rates and
increase the derailment risk in the case of a rail break. 3
An alternative to rail expansion joints in CWR is to allow the rail to be unanchored on bridges under a certain length.
The philosophy behind this approach is that the risk of rail break at cold temperatures, assuming there are no serious
rail flaws, should be less as there is little or no transfer of forces between rail and bridge. See Article 9.8.3.3.5.2. A
drawback is that, should a rail break, there may be little to constrain the resulting rail gap.
Although not specifically simulated in this investigation, damage to decks and fasteners due to large thermal
displacements between rail-tie and tie-deck interfaces has been reported in the field. This will likely be more evident 4
on riveted or bolted top surfaces or where there are other methods of holding ties longitudinally on structures where
ties do not easily slide on the top surface of the span. On long riveted or similarly constrained top surfaces of spans not
protected by expansion joints, fasteners should be selected that are capable of accommodating the expected rail-tie
displacement without damage to ties.
Due to these fundamental conflicts, it is unlikely that all of the design goals will be completely addressed. But a
balance is needed between a number of important considerations. Results emphasize the need to maintain good track
lateral resistance and proper rail neutral temperature on bridge approaches to minimize track buckling potential. On
approaches near expansion bearings track lateral resistance is critical. Methods to provide additional lateral resistance
should be considered – for example, additional width in the ballast shoulders, full height wing walls, sheet piling and
use of ties with improved lateral restraint.
The recommendations in Section 8.3 assume the following:
(1) Maximum hot weather temperature differential values for evaluation of forces due to span expansion and track
buckling risk:

Steel Structures
• A maximum rail ΔT of 45°F above the rail neutral temperature
• A maximum span ΔT of 45°F above the span installation temperature
(2) Cold weather values for the evaluation of rail break risk and effects as follows:
• A maximum rail ΔT of 100°F below the rail neutral temperature
• A maximum span ΔT of 70°F below the span installation temperature
For more extreme temperature variations that might occur in Northern regions of the US or in Canada, site
specific evaluations should be carried out.
One very cold weather case was studied using the same failure criteria.
For example, with rail ΔT of -130°F and span ΔT of -90°F, the thermal rail force alone would be above 300,000
pounds, which is considered a rail break risk for 136 lb. rail.
Fully anchored track on riveted-top structures is likely to be at risk of rail break on all span lengths, with forces
imparted into the bridge predicted to be about 120 percent of AREMA traction and braking forces for rail ΔT of
-130°F.
Controlling rail gap width of a broken rail at ΔT of -130°F to values equivalent to those of anchored track on ballast
away from bridges is highly unlikely for longer spans.
Addition of rail expansion joints would effectively eliminate any cold weather broken rail gap condition without
introducing the risk of track buckling or broken rails. However, costs of installation and maintenance for rail expansion
joints are high and significant bridge degradation is likely to occur due to increased impact loading for such joints
placed on a bridge. For this cold weather case, reducing the maximum span length that may remain unanchored and
without rail expansion joints to 200 feet in Article 8.3.4.2 would reduce the predicted broken rail gap to about 6 inches
(almost equivalent to 5 1/2 inches on anchored track on ballast away from bridges).
Rail gaps of this magnitude are not acceptable in open track or on bridges. Most railroads perform frequent rail flaw
detection in cold weather to find rails that have a high propensity for failure.
9.8.3.2 ANCHORAGE OF DECKS TO BRIDGE SPANS (2012) R(2017)
9.8.3.2.1 Open Deck Bridges
a. The maximum spacing of hook bolts was changed to 4’-8” in 2010 to reflect a connection of every 4th tie assuming 10
inch wide ties and 4 inch clear distance. The previous maximum of 4’-6” assumed 9-1/2 inch wide ties.
b. Bolted fastening systems for timber ties can loosen under train traffic in a relatively short time if loosening is not
prevented. Testing at FAST (Reference 47) has shown that systems employing some method to prevent loosening can
significantly extend the time between maintenance. The provisions of Article 8.3.2.1(b) are based on the results of this
testing. A variety of solutions are possible, some more permanent than others. Locking clips and locking nuts both
allow for future adjustments. New timber ties on riveted girders will typically require a tightening after a settling
period under train traffic. Solutions such as double nuts or thread fastening adhesive can make adjustments more
difficult. More permanent solutions might include tack welding of nuts or mashing of bolt threads; such solutions
might make adjustments impossible.

Commentary
9.8.3.3 ANCHORAGE OF RAIL (2011) R(2017)
9.8.3.3.5 Anchorage Requirements for Continuous Welded Rail (CWR) without Expansion Joints on Open
Deck Bridges
9.8.3.3.5.1 Continuous Welded Rail without Expansion Joints on Open Deck Bridges, Rail Not
Longitudinally Anchored
Some railroads deal with the conflicting problems of potential broken rails and higher forces induced in the rails and bridge by
allowing the rails to be unanchored on bridges up to a certain length. The philosophy behind this approach is that the risk of
rail break at cold temperatures, assuming there are no serious rail flaws, should be less as there is little or no transfer of forces
between rail and bridge.
Railroads with cold weather rail flaw detection and management programs may find this to be an acceptable option. A
drawback is that, should a rail break, there may be little to constrain the resulting rail gap.
9.8.3.3.5.2 Continuous Welded Rail without Expansion Joints on Open Deck Bridges, Rail Longitudinally
Anchored
Testing (Reference 87, 88, 89, 120, 144) indicates that unanchored CWR might allow excessive rail gap widths should a rail
break occur due to cold-induced tension. Anchoring rail as per this Article will reduce the severity of a rail gap due to a cold-
induced tension break but will not reduce the gap to a level that permits train operation at the temperature ranges studied
should a rail break occur.
1
Provisions of this Article recommend rail anchors at all ties anchored to bridge spans for spans 100 feet or less and at all ties
anchored to bridge spans only in the first 100 feet from the fixed end for longer spans. The referenced testing has indicated that
effective longitudinal resistance is dependent upon the interface between tie and structure, and the anchoring used. On spans
with a smooth interface between the tie and structure, whereas the rail fasteners may provide a strong bond between the rail
and the ties, longitudinal restraint is weakest at the tie-to-structure interface. On spans with rivets or bolts protruding from the
top of the bridge, the tie-to-structure interface is likely to be much stronger.
3
The referenced study indicated that when every 2nd tie is box-anchored on track on subgrade away from the bridge, and either:
• every 2nd tie is box-anchored on spans with protruding rivets or similar tie-to-structure interface, or
• every tie is box-anchored on spans with smooth tie-to-structure interface
an equivalent unacceptable rail gap from a broken rail results under the following three conditions: 4
• 300 foot long bridges under the cold scenario (rail ΔT = 100° F)
• 200 foot long bridges under the extreme cold scenario (rail ΔT = 130° F)
• up to 500 foot long bridges in climates that are warmer (rail ΔT = 70° F).
To meet the requirements of this article, rail anchors are placed at the same locations as the anchors between the tie and the
structure at every 4th tie (maximum spacing of 4’-8”) with riveted or similar tie-to-structure interface, or at a reduced spacing
on smooth interfaces. This is equivalent to anchoring at half the anchorage typically used on track on subgrade. To evaluate
reduced longitudinal restraint rail-to-tie fasteners, half the anchorage used on track on subgrade is approximately 40 lb/in/rail
on the bridge.
Unless overriding circumstances exist, anchoring more ties than recommended should also generally be avoided on riveted or
bolted tops or other methods of holding ties longitudinally on spans, as it might increase the risk of either hot-weather-induced
buckling on bridge approaches, or cold weather breaks.

Steel Structures
9.8.3.4 RAIL EXPANSION JOINTS (2017)
While use of rail expansion joints introduces increased cost and bridge degradation, their use can effectively control the risk of
bridge approach track buckling, excessive rail gap widths from cold weather rail breaks, and high forces due to relative
displacement between bridge and track.
Results of factorial testing carried out under very cold temperatures to determine actual span/rail behavior with various bearing
conditions have not been reported. Theory and extrapolation from testing at smaller temperature ranges seem to indicate a
need for rail expansion joints as mentioned above.
There are anecdotal instances of problems where expansion rails were not placed on:
• long bridges with relatively short spans,
• spans over 300 feet with provision for floor system expansion (Article 1.2.13) and
• bridges resulting in damage to deck timbers from standard rail anchors.
But, there is also anecdotal evidence that it is possible to eliminate rail expansion joints on some long spans without serious
consequences.
There are several possible explanations for this:
• It is possible that the bridge and rail neutral temperatures adjust somewhat with changes of season to reduce the
potential severity of the broken rail gaps and the associated forces.
• Rails may have the capacity to resist forces considerably greater than the 300,000 pounds considered a rail break risk
for 136 lb rail. Any weaknesses in rail need to be identified through more frequent rail flaw inspections, especially
during cold weather periods.
• There is a difference in behavior between tie-to-structure interfaces that are smooth and those that have more
resistance to sliding (e.g: protruding fasteners or ties held in place by angles, etc.).
• Use of zero longitudinal restraint rail clips eliminates most of the transfer of longitudinal forces between rail and
bridge structure.
• There is also a difference in behavior in cases where bearings from adjacent spans are placed to allow for opposing
movement (e.g.: two expansion bearings on the same pier) and those where the bearings do not allow opposing
movement.
• Non-functioning bearings may play a significant role.
• Stress in the bridge structure may be higher than expected, but the structure may still be able to accommodate this
condition without noticeable signs of deformation.
Further research is needed to fully explain the conflicting issues.
9.8.3.4.3 Number and Positioning of Rail Expansion Joints on Bridges with Continuous Welded Rail
b. In order to ensure the stability of the backwall while establishing the distance of the rail expansion joint behind the
abutment, consideration must be given to the following parameters:
• Significant dynamic and impact factors

Commentary
• Transition in subgrade stiffness
• Lack of mechanized maintenance
• Low bridge approaches
• Poor drainage
f. Expansion length of rail is limited to 1500 feet in this Manual, which is based on:
• Strength
• Tolerable Rail rupture gap
• Differential displacement between bridge and rail
• Rail stability (Buckling)
9.8.4 UNLOADING PITS
9.8.4.1 SCOPE AND PURPOSE (2015)
The original section on unloading pits was incorporated into the Manual in 1993, before the implementation of the Alternate
Live Load. In the 2013 edition, superstructure was upgraded to accommodate the Alternate Live Load. Following 1
recommended industry practices, the criteria for using an inverted rail as a supporting beam were removed and the designer
directed to use the structural support beams of Figure 15-8-6. The proposed rail sections of the table in the independent
unsupported rail section correlate with the running rail sections of Chapter 4, Part 1. The recommended unloading pit
maximum span was shortened to conform with FRA track tie spacing requirements. The welded diaphragm connection to the
main supporting beams of Figure 15-8-6 was revised to a bolted diaphragm connection to improve the fatigue category and
extend service life while allowing for a reduced beam section.
3
9.8.4.6 STRUCTURAL SUPPORTING BEAMS (2015)
In the 2013 edition, service load design method was used to analyze supporting beams of ASTM A36 structural steel for both
bending and shear, applying the Alternate Live Loading at a maximum speed of 10 mph with the associated live load impact of
28%. The beams in the table in Figure 15-8-6 were resized and the number of beam choices was optimized to better reflect
available choices. A double-angle bolted connection detail was selected for the diaphragms which essentially eliminated any 4
fatigue concerns and allowed for the use of a reduced section.
9.8.4.7 CONCRETE PIT (2015)
Concrete design and detailing were removed from Chapter 15 as they had become obsolete. The designer is directed to
Chapter 8.
9.8.5 WALKWAYS, HANDRAILS, AND FIXED LADDERS ON BRIDGES
9.8.5.2 CLEARANCES AND MINIMUM DIMENSIONS (2014)
9.8.5.2.3 Fixed Ladders
b. Regulations for fixed ladders in the USA for General Industry are contained in OSHA 29 CFR 1910.27, “Fixed
Ladders.” Federal regulations in the USA for fall protection systems refer to FRA Bridge Worker Safety Standards, 49

Steel Structures
CFR 214, Subpart B. Regulations for fixed ladders in Canada are contained in the Canada Labour Code Part II,
Sections 2.6 to 2.8 “Fixed Ladders.” Federal regulations in Canada for fall protection systems refer to the Canada
Labour Code Part II, Section 12.10. Some railroads may also be subject to additional State or Provincial regulations.
9.8.5.3 LOADS (2014)
9.8.5.3.3 Fixed Ladders
The 300 lb load is to ensure that workers carrying tools and equipment may be accommodated.
9.8.7 GUIDE TO THE PREPARATION OF A SPECIFICATION FOR THE CLEANING

AND COATING OF EXISTING STEEL RAILWAY BRIDGES
9.8.7.4 COATING SYSTEMS (2009)
9.8.7.4.1 General
b. The ratio of pigment to vehicle of a coating determines the level of coating gloss, the ease of application and other
properties. Coatings are most often referred to by the resin with which they are formulated. Examples of these are
alkyd, epoxy and urethane.
These various resins react in different ways to develop the dry coating film; for example, oxidation, solvent
evaporation or chemical reaction of multiple components called polymerization. These curing mechanisms, as well as
the other common coating characteristics, are discussed in the following sections.
Coatings for anti-corrosion service are segregated into three distinct types: barrier coatings, inhibitive primers and
sacrificial galvanic protection providers. The barrier coatings offer protection by film forming and creating a barrier to
minimize ion migration and to some extent moisture penetration to the steel substrate. Inhibitive primers reduce
electro-chemical corrosive action at the steel substrate by using sacrificial inhibitive pigmentation in the coating which
is effective in passivating the steel surface and deterring corrosion formation. Galvanic protection prevents corrosion
by using a material of lower electro-chemical potential such as metallic zinc or aluminum pigmentation which
sacrifices itself to protect the steel. This occurs in addition to the film’s barrier protection properties. Galvanic
protective coatings, specifically zinc-rich coatings, offer the highest levels of protection to properly prepared steel
substrate and are resistant to problematic undercutting corrosion. It should be noted that galvanizing can cause
hydrogen embrittlement. This is usually not a problem with very heavy, thick, low strength steel members.
c. Different generic coating types are often used in conjunction with each other as “systems” to provide maximum levels
of protection. However, due to the coating composition, some are not compatible with others. Therefore, development
of this “blend” of different coating types is critical to the long-term performance of the system. By using a systematic
approach to building a coating film, coatings that offer benefits as primers may be coupled with coatings that offer
other desirable characteristics such as moisture, chemical and ultraviolet resistance, plus color/gloss retention, etc.
Manufacturers also formulate coatings with different vehicles or pigment combinations, along with other complex
chemical modifications to maximize the protective qualities.
9.8.7.4.2 Coating Selection
Primary consideration must be given to the service environment which the coating system must endure. Railway bridge
structures are often found in mild exposure environments; however, chemicals used in conjunction with snow and ice removal,
the proximity of structures to industrial plants or factories, and even overspray of agricultural chemicals can dramatically
affect the coating system’s performance. Coating systems for railroad bridges over roadways must also resist abrasion from
splash and stones thrown against coated surfaces by moving traffic and must also have the chemistry within the system to
mitigate the crevice corrosion and pack rust which is usually present on these structures. Certain coating formulations may be
successfully applied with lesser degrees of surface preparation, while others require very clean surfaces. This is a factor which

Commentary
must be given careful attention when choosing a coating system. When cleaning steel on an existing structure where total
removal is required, abrasive blast cleaning to an SSPC-SP5 “White Metal Blast Cleaning”, SSPC-SP10 “Near-White Metal
Blast Cleaning” or SSPC-SP6 “Commercial Blast Cleaning” are the preferred methods of surface preparation. SP5 and SP10
cleaning standards may be difficult to achieve on existing structures under field conditions, especially for open deck structures,
intricate trusses and open box sections. Another consideration in selection of a coating system is the ability of the topcoat to
accept additional surface preparation and touch-up or overcoating. Many topcoats cure to form smooth, dense and hard films.
Hard, abrasion-resistant coatings, such as two-component urethanes, may require more rigorous surface preparation, such as
abrasive blast cleaning, to superficially roughen the surface and promote adhesion of subsequently applied coatings. On the
other hand, softer film topcoats, like alkyds or acrylics, often accept additional maintenance coats of paint with minimal
surface preparation (such as simple solvent cleaning or high pressure water washing).
The materials and methods used to clean and coat steel bridge structures are constantly changing. The following items, as a
minimum, should be considered for all coating specifications:
• Life expectancy and life cycle costs
• Successful protection of the structure and its critical elements (joints, connections, bearings, etc.)
• Compatibility with existing systems (where applicable)
• Ease of application and availability of materials
• Environmental conditions
• Aesthetics
1
• Overall coating strategy
9.8.7.4.3 Materials/Systems
a. Penetrants for treating crevice corrosion and pack rusted joints that cannot be cleaned are as follows: 3
(1) Epoxy Penetrating Sealers – Epoxy penetrating sealers are low molecular weight epoxies based on Chelated
Polymeric Oxirane technology. These high performance, two-component chemically-cured high solids epoxy
penetrating sealers are recommended for rusty steel when environmental, economic or safety concerns restrict
abrasive blast cleaning. The extraordinary penetrating properties of these sealers provide a means of reinforcing
rusty steel substrates, insuring adhesion of subsequent coatings. They are equally effective at penetrating,
reinforcing and sealing concrete and masonry surfaces in all industrial environments. They improve the 4
effectiveness and efficiency of the maintenance coating process by penetrating and sealing crevices, joints, back-
to-back angles and edges of old coatings, improving the service life of the maintenance coating system. These
sealers also serve to seal aged “White-Rusted” zinc galvanized surfaces for recoating. Epoxy penetrating sealers
are two-component products that cure by crosslink polymerization. These coatings provide excellent adhesion to
marginally prepared steel (SP2 minimum) and old coatings. Their lower viscosities allow epoxy sealers to
penetrate rust and wick into surface voids and around rivet heads. In addition, this wicking action penetrates
discontinuities in existing coatings which often times seals these areas and reduces undercutting and peeling. The
low viscosity also allows epoxy sealers to be applied by many techniques. This includes conventional air and
airless spray, brush and roller, flood or flow coating methods, and by low pressure hand pump sprayers (similar to
those used to spray concrete curing compounds, form release oils, or garden chemicals). Epoxy penetrating sealers
usually possess very high volume solids content typically over 80%, develop lower contractive curing stresses,
and meet the most stringent VOC regulations (often zero VOC). Corrosion inhibitors are generally used in their
formulations. Since epoxy penetrating sealers provide low film build (1-2 mils), the total amount of curing stress
and physical coating weight that the existing coatings must bear is also reduced. The drawbacks of these coatings
are that they require multiple component mixing, have short pot lives, cure hard and may crack on flexible
structures, must be topcoated to achieve maximum resistance, have high material cost, do not stay active if applied

Steel Structures
to crevice corroded or pack rusted joints and connections, have critical recoat times, and have application
temperature limitations. The sealers are usually used as primers to bind up surfaces to be overcoated and are then
topcoated with alkyd, acrylic, epoxy, or urethane coatings.
(2) Moisture Cured Urethane Penetrating Sealers – These penetrants are thin and designed to flow into the joints and
connections binding them up and sealing them up. For additional information see b(4).
(3) High Ratio Co-Polymerized Calcium Sulfonate Penetrant Sealers – These penetrants are active non drying
chemical treatments for crevice corroded and pack rusted joints and connections designed to stop corrosion by
neutralizing acid, displacing moisture and scavenging oxygen. For more information see b(8).
b. Coatings for recoating prepared steel are as follows:
The premier coatings for blast cleaned steel have historically been 3 coat zinc epoxy urethane systems (which require
this type of surface to perform properly). However, this does not mean that these coatings are the answer in all
situations, as they have limitations as well. The following describes the various coating types which are available,
gives a brief history of their development and usage, addresses surface preparation requirements, discusses touch-up
capabilities, reviews costs, and in some cases provides an estimate of the service life given the assumed exposure
conditions.
(1) Alkyds – Alkyds are a type of synthetic resin that cures by air oxidation. They are basically formed by a reaction
among an acid, an alcohol, and oil. Alkyds are formed and classified by the amount and type of oil present within
the formulation. “Long oils” contain greater quantities of oil and take longer to dry, while “short oils” have less oil
and shorter dry times. Medium length oil-alkyds are a good compromise of the two and are consequently the most
versatile and widely used. With the reduction in the amount of natural oil and an increase in the synthetic alkyd
resin, the resistive properties of the alkyds are superior to those of natural oils. The use of synthetic resin translates
into improved resistance to water, but has little or no effect on the resistance to chemicals and solvents.
Because of the presence of the oil in the alkyd, which aids in surface wetting, surface preparation requirements are
minimal. Therefore, the removal of all loose materials by hand or power tools is usually adequate for the use in
mild to moderate exposures. However, abrasive blast cleaning or water jetting to the same cleaning standard (i.e.
SSPC-SP6 or SSPC-SP12-WJ3) still provides superior surface cleanliness and may increase long-term coating
system performance. In most cases the high cost of such surface preparation would indicate the use of higher
performance coatings which would increase long term performance. Touch-up may be performed with a similar
material, or an oil-based coating if necessary. They can be easily applied by maintenance personnel.
(2) Modified Alkyds – The versatility of alkyds is further enhanced by combining them with any number of natural
and synthetic resins. By modifying the basic alkyd, additional corrosion protection may be gained while the ease
of application and surface tolerance is maintained. While the basic alkyd resins have been modified by combining
them with other materials, the modified product does not develop all of the characteristics of these materials. They
offer increased performance to the oil-based resins alone, but do not match the performance of the more advanced
coatings. These materials offer a potential solution for mild to moderate environments where additional protection
is necessary, but a cost-effective coating is desired. While there are many types of modified alkyds that have been
developed for specific uses, this discussion will include only three that have significance to steel from which all
coatings have been removed: vinyl alkyds, calcium sulfonate modified alkyds and silicone alkyds.
(a) Vinyl Alkyds – The Vinyl Alkyds offer decreased drying times, better adhesion and water resistance, and
improved exterior durability compared to the basic alkyd formulation. Because of the vinyl modification,
some formulations are also capable of being topcoated with high performance, stronger-solvent topcoats such
as epoxy or urethane. Vinyl alkyd modifications are generally used in readily recognized products referred to
as “universal metal primers”.
(b) Silicone Alkyds – Similar to the vinyl alkyd, the silicone alkyds as compared to unmodified alkyds offer an
increase in corrosion protection. The silicones also offer the capability of resisting somewhat higher

Commentary
temperatures while also improving gloss retention, color retention, and abrasion resistance. The increase in
resistance qualities appears to be directly related to the quantity of silicone used in the modification. As such,
the amount of silicone should be selected and specified. A 30% silicone content is a minimum amount of
silicone commonly specified to ensure superior performance.
(c) Calcium Sulfonate Modified Alkyds – similar to the silicone modified alkyd in that a small percentage of
calcium sulfonate is used to enhance the properties of the base alkyd resin. The calcium sulfonate is added to
the formula to give the alkyd better corrosion resistance, wetting properties, thixotroscopy (ability to resist
runs or sags) and as a pigment suspension agent. The amount of calcium sulfonate in modified alkyds may
range from 2 to 15 percent by weight with an improvement in performance with increasing calcium sulfonate
content. For best performance a percentage by weight of 14% to 15 % is recommended. It is also important to
insure that the ratio of calcium carbonate to active sulfonate is approximately 10 to 1. This ratio is required for
a balanced formula and is the ratio that has been used in the field proven materials. To reduce costs some
suppliers may supply what they call a calcium sulfonate alkyd but the formula is basically low cost, low
quality calcium carbonate filler with only a small percentage of active sulfonate added. Specifications should
clearly define the percentage of active sulfonate and quality control procedures should be put in place to
enforce the specification.
(3) Zinc-Rich Coatings – Zinc-rich coatings provide a high level of protection for blast cleaned steel, but are
expensive relative to other coatings. Zinc-rich coatings provide a combination of barrier and galvanic protection.
Zinc dust dispersed through various resins provides the galvanic and barrier protection as well as improved
abrasion resistance. Zinc-rich coatings offer significantly better performance than other types, through galvanic
action described earlier. This protection greatly reduces sub-film corrosion and cancerous undercutting corrosion.
Their limitations include somewhat higher cost, reliance on a high degree of surface preparation, skilled
1
applicators, and careful selection of intermediate and/or topcoats. Zinc-rich primers require the surface to be free
of flash rust for good performance. The industry standard is for surface temperature to be several degrees above
the dew point for zinc primer application. Zinc-rich coatings used alone also offer reliable one-coat protection in
normal weather conditions.
Zinc-rich coatings are available in organic and inorganic formulations. Inorganics are considered to provide
superior protection, but they are more sensitive to the surface preparation and applicator skills. Inorganic zinc-rich
3
coatings require surface preparation to Near-White Metal (SSPC-SP10) at a minimum, with White Metal (SSPC-
SP5) preferred. Field touch-up is performed with an organic material, such as a surface-tolerant epoxy, primarily
because inorganic zinc-rich coatings require spray application and are less user friendly. They can be used very
effectively in maintenance applications, but should be substituted with organic zinc on complex surfaces, e.g. steel
lacing, corroded crevices, pack rusted joints and connections. Inorganic zinc primers may be used in one, two or
three-coat systems. Usually, epoxies are used as intermediate coats and acrylic aliphatic urethanes as finish coats; 4
however, waterborne acrylic coatings have also been successfully used as topcoats for zinc-rich coatings.
Inorganic zinc or galvanizing are the preferred shop primers for replacement steel used to repair existing
structures. Care must be taken to ensure that all shop and field coatings specified are compatible.
Organic zinc coatings can be made from many different generic coating types, but the most prominent are epoxy
and moisture-cured urethanes. Epoxy zinc-rich coatings have primarily the same characteristics as epoxies;
excellent adhesion, abrasion resistance, good water resistance, and if modified increased flexibility. The zinc dust
dispersed in the coating gives galvanic (sacrificial) protection against corrosion and improves abrasion resistance.
Moisture-cured organic zinc-rich primers have the advantage of galvanic protection coupled with good adhesion,
abrasion resistance, and sunlight resistance when topcoated with a moisture-cured aliphatic topcoat. In addition
moisture-cured urethane organic zinc-rich primers have the ability to be applied in high humidity and colder
temperatures. Field application must be monitored carefully as too much moisture will cause carbon dioxide
gassing or poor adhesion. Coating manufacturers often tout the ability of their moisture-cured primer to adhere to
damp steel. Organic zinc coatings are generally topcoated with epoxies, acrylic urethanes, 100% acrylics, or
moisture-cured urethanes.

Steel Structures
(4) Moisture-Cured Urethanes – Moisture-cured urethane coatings react with atmospheric moisture (humidity) which
initiates the cure, creates carbon dioxide gas and provides a protective coating film. These single-component
products have excellent performance characteristics, including abrasion resistance, durability, and appearance.
Zinc-rich primer formulations made from moisture-cured resins give excellent protection against corrosion of
steel. Many moisture-cured urethane intermediate and finish coat formulations use micaceous iron oxide to
provide corrosion resistance. Moisture-cured urethane coatings are ideal for field application, since they may be
applied in periods of high humidity and moderate cold temperatures. Moisture-cured urethane coatings have
several unique disadvantages. They are moisture sensitive in the container, which can lead to gelling. If too much
moisture is present they will produce excessive carbon dioxide gas that could damage the film. When properly
cured they provide a hard and smooth coating film that may be difficult to overcoat in the future. Recoat windows,
the time during which an additional coat can be applied without additional surface preparation such as sanding or
light abrasive blasting, are narrow. They are more costly to purchase than other high-performance coatings but
may be more cost effective if conditions for application are right. Moisture-cured urethanes have only fair
flexibility, limited resistance to acid and chemicals, and notable yellowing when exposed to the ultraviolet rays of
sunlight. Moisture-cured urethanes require careful control of application thickness, particularly in windy, humid
conditions.
(5) Epoxy Coatings – Epoxy coatings have excellent adhesion to steel, excellent abrasion resistance, good water
resistance, and when modified relatively good flexibility. For bridge coatings, epoxy resins are used primarily for
zinc-rich primers, and for intermediate coats over inorganic or organic zinc-rich primers. Since epoxies are two-
component materials, they must be mixed in proper proportions to cure correctly. Other disadvantages of epoxy
coatings are that most materials have limited pot lives, specific recoat time intervals, and application temperature
and humidity restrictions. Epoxies are not usually used as finish coats because UV light attacks the structure and
they break down causing chalking.
(6) Epoxy Mastic Coatings – Epoxy mastic coatings cure by chemical reaction when a hardener is added to the resin.
Since the percentage of solids by volume is higher than that of regular epoxies, the amount of solvent used in the
coating formulation is low. Therefore, most epoxy mastics are VOC compliant and are less likely to overly soften,
wrinkle, or lift old coatings. They also may offer a higher film build per coat, which serves to improve on already
good abrasion and environmental resistance. Additionally, many formulas have low temperature catalysts or
additives which may extend the coating season into periods of cooler weather. Epoxy mastic coatings also readily
lend themselves to modifications which enhance their corrosion resistance and film strength. One such
modification is the addition of leafing or non-leafing aluminum into the coating, which serves to lower the epoxy
resin’s susceptibility to degradation by ultraviolet light and decreases moisture permeability of the film. For new
construction or exposed surfaces, aluminum flake pigmented epoxy mastic systems are the industry standard when
epoxy mastics are used. This addition increases the corrosion protection of the system and the mechanical strength
of the coating film.
Disadvantages of epoxy mastics are their higher cost and the epoxy resins’ inherent degradation by ultraviolet
light. Sunlight and weathering exposure commonly result in chalking and/or color fading of the exposed film. As a
result, if chalking and discoloration cannot be tolerated, they must be topcoated with better gloss and color
retentive finish coats, such as polyurethanes or acrylics. However, it is important to note that any chalking which
takes place has been found to have little or no effect on coating performance other than life expectancy due to the
film eroding away. Other disadvantages include slower drying time and strong odors. Other limitations of epoxy
mastics are that many have limited recoat times and multiple components which require mixing, include toxic
solvents and have limited pot life. These coatings are toxic and require special handling in the field.
A large variety of epoxy mastic formulations exist, with dramatic differences in performance between the best and
the worst. Proper specification is needed to achieve the desired results from this coating.
(7) Waterborne Acrylic Coatings – Waterborne acrylics are single-component coatings which cure by coalescence of
the resin particles that are dispersed in water. Variations of waterborne acrylics are used in both protective and
architectural coatings in the form of primers, intermediate coats, and finish coats. These materials have higher
moisture vapor transmission rates which allow moisture to readily pass through the coating film. Therefore, in

Commentary
coatings for use on steel, anti-corrosive pigments are added to inhibit rust formation. Acrylics offer excellent
exterior durability along with gloss and color retention similar to that achieved by urethane coatings. Waterborne
acrylics also have excellent flexibility, good drying times under low humidity conditions, relatively low odor, are
easy to apply, and readily accept future overcoats. Lower abrasion resistance properties, along with relatively
higher costs, are some of the disadvantages of waterborne acrylic coating materials. Limitations of acrylic coatings
are their fair corrosion resistance, application temperature limitations above 50°F, and relatively poor chemical
resistance compared to a two-component high performance coating system such as an epoxy. Waterborne acrylic
coatings are not resistant to high levels of moisture or prolonged condensation.
(8) High Ratio Co-Polymerized Calcium Sulfonates – High Ratio Co-Polymerized Calcium Sulfonate coatings are
different from Calcium Sulfonate Modified Alkyds. High Ratio Co-Polymerized Calcium Sulfonates are made up
of a co-polymerized reacted synthetic resin with a unique patented crystalline modification that cures by air
oxidation. These coatings provide protection by a combination of chemical and physical properties. The coatings
are excellent chemical treatments and film formers, and in both field and laboratory tests have demonstrated that
they are at the top of the performance envelope, when compared to traditional multi-coat high performance
coatings. The High Ratio Co-Polymerized Calcium Sulfonate coatings' major advantage is the active
Penetrant/Sealer and Primer/Topcoat, which have a fifteen-year history on structures in the field, and that stop the
progression of crevice corrosion and pack rust specifically in joints and connections. This activity, in the joints and
connections, is unique to the High Ratio Co-Polymerized Calcium Sulfonate chemistry and supplies the engineer
with a valuable tool for the preservation of aging complex structures, where crevice corrosion and pack rust are
present. In addition the coatings are very environmentally friendly with the system having a LC50 at 96 Hrs fish
kill at 41007 ppm (note typical epoxies and urethanes are 2-4 ppm). This test is used to assess the toxicity of
coatings if they are introduced into the fish habitat. The performance of the High Ratio Co-Polymerized Calcium
Sulfonate coatings is related directly to the percentage amount of synthetic crystalline based material and of active
sulfonate in the formulation. The ratio should be a minimum 90 to maximum 105 TBN (Total Base Number) and a
1
minimum 9.5 to 11% active sulfonate. There must be a minimum 9 to a maximum 11 to 1 ratio total base number
to active sulfonate. Calcium sulfonate coatings with lower active numbers will not perform as well and are not
equal to the High Ratio Co-Polymerized Calcium sulfonate type products, and should not be included in the same
specification. The formulation should contain no fillers or extenders. Some manufacturers fill their coatings with
low cost calcium carbonate fillers to lower the price with a negative impact on long term performance. Unlike
calcium sulfonate alkyds the alkyd or co-polymer used in conjunction with the High Ratio Co-Polymerized 3
Calcium Sulfonate should not comprise more than 25 to 27 % of the formulation. Formulations with more than
27% alkyd or co-polymer would not be considered equal to the High Ratio Co-Polymerized Calcium Sulfonate
formulations which have set the high performance standard since 1991. Increasing the amount of alkyd or co-
polymer is a way to reduce the cost, with the net effect of reducing the long term performance.
(9) Galvanizing – Hot-dip galvanized steel has been effectively used for more than 150 years. The value of hot-dip
galvanizing stems from the relative corrosion resistance of zinc, which, under most service conditions, is 4
considerably better than iron and steel. In addition to forming a physical barrier against corrosion, zinc, applied as
a hot-dip galvanized coating, cathodically protects exposed steel. Furthermore, galvanizing for protection of iron
and steel is favored because of its low cost, the ease of application, and the extended maintenance-free service that
it provides. Though the process may vary slightly from plant to plant, the fundamental steps in the galvanizing
process are:
Soil and grease removal – A hot alkaline solution removes dirt, oil, grease, shop oil, and soluble markings.
Pickling – Dilute solutions of either hydrochloric or sulfuric acid remove surface rust and mill scale to provide a
chemically clean metallic surface.
Fluxing – Steel is immersed in liquid flux (usually a zinc ammonium chloride solution) to remove oxides and to
prevent oxidation prior to dipping into the molten zinc bath. In the dry galvanizing process, the item is separately
dipped in a liquid flux bath, removed, allowed to dry, and then galvanized. In the wet galvanizing process, the flux
floats atop the molten zinc and the item passes through the flux immediately prior to galvanizing.

Steel Structures
Galvanizing – The article is immersed in a bath of molten zinc at between 815º-850º F (435º-455º C). During
galvanizing, the zinc metallurgically bonds to the steel, creating a series of highly abrasion-resistant zinc-iron
alloy layers, commonly topped by a layer of impact-resistant pure zinc.
Finishing – After the steel is withdrawn from the galvanizing bath, excess zinc is removed by draining, vibrating
or for small items - centrifuging. The galvanized item is then air-cooled or quenched in liquid. Galvanized steel
that is to be topcoated for cosmetic considerations must be air-cooled without quenching to avoid adherence
problems.
Inspection – Coating-thickness and surface-condition inspections complete the process.
(10) Metalizing – Metalizing is a thermal spray process that requires surface preparation by abrasive blasting followed
by metal spraying which can then be sealed and thereafter topcoated. There is a higher initial cost for metalizing
but new application technologies and life cycle costing show that it is about half the cost of coating with high
performance three coat systems. The three spray wires used for atmospheric or immersion service are pure
aluminum, pure zinc or an 85/15 alloy of these two metals. (The alloy is approximately 85% zinc and 15%
aluminum by weight.) A metalized coating may be bare sprayed metal, sprayed-metal-plus-sealer or sprayed-
metal-plus-sealer-plus-topcoat. Coating thickness may vary according to application from .004" to thicker coats of
zinc in the range of .012" - .014" for seawater splash zones. Metalizing is considered a cold process in that the
aluminum or zinc is deposited onto steel by spraying rather than by dipping the steel into a bath of molten zinc as
with galvanizing. The steel remains relatively cool at about 250º-300ºF. There is virtually no risk of heat distortion
or weld damage by metalizing. There are no VOC's (volatile organic compounds) in the metalized coating. There
is no cure time or temperature limit for metalizing, so metalizing may be applied throughout the year. The sealed-
sprayed-metallic coating is often the most economical and is the preferred system of the three metalized coating
options as it offers the longest service life. The use of a coating directly over an unsealed sprayed-metal coating
should be avoided. The disadvantage to the system is that the blast profile is very specific. The profile must be a
minimum of 4 to 4.5 mils and angular in nature. Careful inspection is required to insure it is achieved.
(11) Polyurea Coatings – Polyurea-based thick film coatings encompass a diverse group of products. A pure polyurea is
the combination of isocyanates with a long chain amine, excluding the hydroxyl reactive sites. For reference, pure
polyurethane coatings are formulated using an isocyanate combined with hydroxyl-containing polyols. Polyurea
coatings can be formulated as hybrids by combining isocyanates with a mixture of polyols and long chain amines,
resulting in a coating that bears the performance characteristics of a polyurethane and a polyurea coating. Polyurea
coatings can be either aromatic or aliphatic, and can be formulated with catalysts, pigments, fillers and other
performance-enhancing additives. Pure polyurea coatings offer the highest degree of chemical resistance, but
hybrids offer improved wetting (the cure time is retarded) and other desirable performance characteristics. The
relative production cost is lowest for a polyurethane, increases for polyurea hybrids, and is the highest for pure
polyurea coatings. These new technology polyurea coatings and their hybrids offer the industry an
environmentally compliant, high performance option (with very attractive film forming properties) for corrosion
prevention and asset protection. However, like all industrial protective coatings they have performance limitations
and minimum surface preparation requirements. Use of these materials outside of the recommended service
environments or over marginally prepared surfaces can result in catastrophic failure and costly rework.
c. Coatings for Overcoating Existing Coatings and Stable Substrates
Many coating types are reformulated specifically for use as overcoating materials. At a minimum, the following
generic coating types would usually be recommended for over coating the existing coatings on railway structures.
Compatibility testing should be done between the coating to be overcoated and the coating to be applied to insure it
will not delaminate or otherwise adversely affect the adhesion properties of the existing coating.
(1) Alkyds – See b(1).
(2) Modified Alkyds – See b(2).

Commentary
(3) Epoxy Mastic Coatings – Epoxy mastic coatings offer many advantages as overcoats and are widely specified for
use as an overcoating material. Because epoxy mastics are formulated to have good wetting properties, they
possess excellent adhesion to marginally prepared contaminant free surfaces (SSPC-SP2 minimum). Testing
should always be done to insure compatibility with the existing coating. When properly formulated, the coatings
will maintain very low stress, making them good overcoat candidates for aged alkyds. For additional information
see b(6).
(4) Moisture-Cured Urethane Coatings – See b(4)
(5) Low Molecular Weight Epoxy Penetrants – See a(1)
(6) Waterborne Acrylic Coatings – Waterborne Acrylic Coatings are good overcoating materials because they have
lower shrinkage stresses as they cure and therefore apply little contractive stress on existing coatings. Since these
coatings use water as a solvent, they are VOC compliant and do not over-soften or lift existing films. They are
typically used in overcoating as finish coats over epoxy mastics or epoxy penetrant sealers. For additional
information refer to b(7)
(7) High Ratio Co-Polymerized Calcium Sulfonates – See b(8)
(8) Urethane Systems – Chemically cured acrylic urethane coatings are not typically used as overcoating primers, but
do offer excellent characteristics as finish coats with superior gloss and color retention, and UV resistance over
some of the materials previously discussed (Zinc primers, epoxy mid coats, epoxy mastics, epoxy penetrating
sealers, and moisture cured urethanes). These coatings offer excellent water and corrosion resistance. They also
allow lower application temperatures, and can be modified to be high solids, high build, or 100% solids coatings,
thus VOC compliant. Disadvantages with urethane coatings are that they have limited flexibility and are two-
1
component materials with a limited pot life. They are also moisture sensitive during application and may haze or
blush (develop a cloudy milky looking appearance) if applied during periods of high relative humidity. Another
disadvantage is that the coating film produced is slick and hard, which may necessitate substantial surface
preparation prior to future overcoating operations. This disadvantage may also prove to be advantageous in that
graffiti can easily be removed from high gloss urethane coated bridges by wiping with solvent.
3
WELDING INDEX (2004)
This Welding Index makes reference to some of the articles in the Manual pertaining to Welding involved in design,
fabrication, repair and rating of steel structures. This index does not include every reference to welding within the Manual, but
can serve as a ready guide for designers.
4
Subject Article Reference
Allowable stresses–base metal 1.3.13; 1.4.1; 9.1.4 and Tables 15-1-9, 15-1-10 and 15-7-1
1.3.13; 1.4.2; 6.5.36.10b; 9.1.4 and Tables 15-1-9, 15-1-14 and
Allowable stresses–weld metal
15-7-1
Attachments 1.10.4
Bridge types 1.2.3
1.2.2; 1.10.2; 1.10.6; 1.14.1; 3.3.1a; 3.3.5; 3.5.5b and c; 9.1.2.2; 9.1.4.2;
Bridge welding code, AWS D1.5
9.1.10.1; 9.1.14.1; 9.3.1.6
Butt joints 1.7.2.2a; 1.10.1; 7.4.4; 8.1.4.12c; 9.1.10.1
Closed boxes 1.5.15
Combination of welds with fasteners 1.5.12b; 9.1.5.12

Steel Structures

Compression in welded box-type flexural
1.4.1
members
Compression in welded built-up flexural members 1.4.1; 1.6.1; 1.6.2; 1.6.3; 1.6.4.2g; 9.1.7.1
Connections 1.5.9
Connections, field welded 1.5.10; 1.5.12b
Cover plates, fillet welded 1.7.2.2b
Deck plate–girder flanges 1.7.4b
Drainage pockets 1.5.6
Eccentric connections 1.5.7b
Existing bridges 7.4.1.5; 7.4.2.2; 7.4.2.3a and b; 7.4.2.4; 7.4.3.2; 7.4.4.1e
Fabrication tolerances 3.1.7.1d and e; 3.1.7.2
Fatigue 1.3.13; 7.3.3.2c and f
Field welding 1.5.10; 4.19; 9.1.5.10
Fillet welds 1.7.4b; 1.10.3; 3.3.3; 3.5.5b and c; 7.4.2.3a and b; 9.1.10.4
Fire damage 8.6.5c
Flange to web welds 1.7.4b; 3.3.3; 3.5.5c; 9.1.7.4; 9.3.3.3
Flange splices 1.7.5
Fracture critical members Section 1.14
Full penetration groove welds 1.5.9b; 1.7.5; 1.7.6c; 3.1.10a; 3.5.5b; 6.5.36.10c; 9.1.7.4
Groove welds 1.10.1; 3.5.5b
Impact tests 1.14.5 and Table 15-1-15
Inspection 3.5.5; 7.2.7; 7.2.8; 9.3.5.5
Intermittent fillet welds 1.10.2; 7.4.2.3b
Lacing bars, fillet welded 1.6.4.2g
Lateral bracing 1.11.2d
Longitudinal load–welded rail 1.3.12
Machine welding 9.3.3.3
Materials 1.2.1 and Tables 15-1-1 and 15-1-2
1.2.3; 1.7.1; 1.7.2.2; 1.7.4b; 1.7.5; 1.7.6; 1.7.8a and b; 7.4.2; 9.1.7.1;
Plate girder
9.1.7.4
Plug welds 1.10.2
Prohibited welds 1.10.2
Sealing welds 1.5.5; 1.5.13c
Seam welding 6.7.5.35c
Sheaves 6.5.36.10
Shop painting joints 3.4.1b
Slot welds 1.10.2
Splices 1.5.9b; 1.7.5c and c; 6.6.6
Spot welding 6.7.5.35c

Commentary

Stay plates 1.6.3d
Stiffener plate 1.7.7a and d; 1.7.8b; 1.10.4; 3.1.10; 7.4.2.2
Tack welding 3.3.4; 3.3.5; 9.3.3.4
Welded construction 1.7.2.2; Section 1.10; Section 3.3; 6.2.11.9; 7.4.1.5
Welder and welding operator qualifications 3.3.5
Welding: connection angle flexing (O.S.L.) leg 1.8.3a
Welding: electrodes 1.2.1
Welding: machinery weldments 6.2.11.9; 6.5.36.10
Welding: preparation of material 3.3.2
Welding: repair of flame cut edges 3.1.6
Welding rope splices 6.6.6
Welding: requirements 1.2.2; 1.14.4; 3.3.1; 7.4.1.5

Steel Structures
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Section/Article Description Page

9.1.1 Proposals and Drawings . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-6 3

9.1.2 General Requirements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-7

9.1.3 Loads, Forces and Stresses . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-12

© 2017, American Railway Engineering and Maintenance-of-Way Association 15-9-1

TABLE OF CONTENTS (CONT)

Section/Article Description Page

9.1.3.13 Fatigue (2016). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-18

9.1.4 Basic Allowable Stresses . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-26

9.1.5 General Rules. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-30

9.1.6 Members Stressed Primarily in Axial Tension or Compression . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-32

9.1.7 Members Stressed Primarily in Bending. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-35

9.1.8 Floor Members and Floorbeam Hangers . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-39

9.1.9 Construction with Fasteners. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-39

9.1.10 Welded Construction. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-40

9.1.11 Bracing . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-42

© 2017, American Railway Engineering and Maintenance-of-Way Association

15-9-2 AREMA Manual for Railway Engineering

TABLE OF CONTENTS (CONT)

Section/Article Description Page

9.1.12 Pins and Pin-Connected Members . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-43

9.1.13 Continuous and Cantilever Steel Structures . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-43

9.1.14 Fracture Critical Members . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-44

9.3.1 General . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-47

9.3.2 Bolted Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-49

9.3.3 Welded Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-50

9.3.6 Shipment and Pay Weight . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-51

9.4.6 Plans (2017) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-51

Part 5 Bearing Design and Construction

9.5.1 Bearing Design. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-51

© 2017, American Railway Engineering and Maintenance-of-Way Association

AREMA Manual for Railway Engineering 15-9-3

TABLE OF CONTENTS (CONT)

Section/Article Description Page

9.5.1.2 General Requirements (2010) R(2017). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-51

9.5.2 Basic Allowable Stresses . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-55

9.5.3 Steel Bearing Component Design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-56

9.5.4 Bronze or Copper-Alloy Sliding Expansion Bearing Design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-56

9.5.5 PTFE Sliding Bearing Surface Design. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-56

9.5.7 Multi-Rotational Bearing Design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57

9.5.10 Bronze or Copper-Alloy Sliding Expansion Bearing Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57

9.5.11 PTFE Sliding Bearing Surface Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57

Part 6 Movable Bridges

9.6.1 Proposals and General Requirements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-57

9.6.2 General Features of Design. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-59

9.6.3 Loads, Forces and Stresses . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-59

9.6.4 Basic Allowable Stresses and Hydraulic Pressures. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-59

9.6.5 General Details. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-59

© 2017, American Railway Engineering and Maintenance-of-Way Association

15-9-4 AREMA Manual for Railway Engineering

TABLE OF CONTENTS (CONT)

Section/Article Description Page

9.6.5.36 Special Provisions for Vertical Lift Bridges (1997) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-60

9.6.6 Wire Ropes and Sockets . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-60

9.6.9 Erection. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-62

Part 7 Existing Bridges

9.7.2 Inspection . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-63

9.7.3 Rating . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15-9-63 1